New Canalization of the Nederrijn and Lek Appendix

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NEW CANALIZATION OF THE NEDERRIJN AND LEK APPENDIX Design of a weir equipped with fibre reinforced polymer gates which is designed using a structured design methodology based on Systems Engineering

25 January 2013 : Henry Tuin

Appendices New canalisation of the Nederrijn and Lek

Colophon

Reference: Tuin H. G., 2013. New canalization of the Nederrijn and Lek – Design of a weir equipped with fibre reinforced polymer gates which is designed using a structured design methodology based on Systems Engineering (Master Thesis), Delft: Technical university of Delft.

Key words Hydraulic structures, weir design, dam regime design, Systems Engineering, canalization of rivers, fibre reinforced polymer hydraulic gates, Nederrijn, Lek, corridor approach, river engineering.

Author: Name: ing. H.G. Tuin Study number: 1354493 Address: Meulmansweg 25-C 3441 AT Woerden Mobile phone number: + 31 (0) 641 177 158 E-mail address: [email protected] Study: Civil Engineering; TUDelft Graduation field: Hydraulic Structures

Study: Technical University of Delft Faculty of Civil Engineering and Geosciences Section of Hydraulic Engineering Specialisation Hydraulic Structures CIE 5060-09 Master Thesis

Graduation committee: Prof. drs. ir. J.K. Vrijling TU Delft, Hydraulic Engineering, chairman Dr. ir. H.G. Voortman ARCADIS, Principal consultant Water Division, daily supervisor Ir. A. van der Toorn TU Delft, Hydraulic Engineering, daily supervisor Dr. M.H. Kolstein TU Delft, Structural Engineering, supervisor for fibre reinforced polymers

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Appendices New canalisation of the Nederrijn and Lek

Contents

Colophon ...... i

A. Literature study ...... 1 A.1 Introduction to the Nederrijn and Lek ...... 1 A.2 History ...... 3 A.2.1 The Rhine branches ...... 3 A.2.2 Canalization of the Rhine branches ...... 6 A.2.3 The canalization project of the Nederrijn (1960) ...... 8 A.2.4 Dam regime ...... 10 A.3 Present water system ...... 11 A.3.1 Water management in the Netherlands ...... 11 A.3.1.1 The flow of water through the Netherlands ...... 11 A.3.1.2 Present dam regime of the Nederrijn and Lek ...... 13 A.3.1.3 Water levels for the Nederrijn and Lek ...... 13 A.3.1.4 Safety levels of the levees in the Netherlands ...... 14 A.3.1.5 Change of the maximum design discharges ...... 14 A.3.1.6 Governing high water levels ...... 16 A.3.2 Sediment transport ...... 16 A.3.2.1 Bed transport ...... 17 A.3.2.2 Suspended sediment transport ...... 18 A.3.3 Salinization of the delta ...... 19 A.3.4 Autonomous bed erosion...... 20 A.4 Future situation ...... 21 A.4.1 Second delta committee ...... 22 A.5 Description of the weirs in the Nederrijn ...... 24 A.5.1 Present weirs ...... 24 A.5.1.1 Variant analysis ...... 25 A.5.1.2 Weir design ...... 26 A.5.2 Changes in the design ...... 30 A.5.2.1 Waterpower installation at weir Amerongen (1984) ...... 30 A.5.2.2 Improving fish migration by building fish ladders (1990) ...... 31 A.5.2.3 Undesired vibrations of the visor gates of weir Hagestein (1993) ...... 31 A.5.2.4 Waterpower installation at weir Hagestein (2005) ...... 31 A.5.3 State of the weirs ...... 32 A.5.3.1 Functions of the weirs ...... 33 A.5.3.2 Failure scenarios ...... 35 A.6 Weir designs ...... 39 A.6.1 The infuence of a weir on the water levels ...... 39 A.6.2 An overview of weirs ...... 44

B. Information and methods ...... 51 B.1 Systems engineering theory ...... 51 B.2 Multi criteria analysis method ...... 52 B.3 Model description for configuration variants ...... 54 B.3.1 Input data and boundary conditions of the present situation ...... 54 B.3.2 Visualization of the configuration ...... 59

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B.4 Fibre reinforced polymers (FRP) ...... 61 B.4.1 Reinforcement ...... 61 B.4.1.1 Glass fibre reinforcement ...... 61 B.4.1.2 Polyaramid fibre reinforcement ...... 61 B.4.1.3 Carbon fibre reinforcement ...... 62 B.4.1.4 Application of reinforcement ...... 62 B.4.2 Resins ...... 62 B.4.3 Cores ...... 63 B.4.4 InfraCore® panels patented by FiberCore ...... 63 B.4.5 Fabrication of fiber reinforced polymer plates ...... 64 B.4.5.1 Open mould processes ...... 64 B.4.5.2 Closed mould processes ...... 65 B.4.5.3 Continuous process ...... 66 B.4.6 Design properties of laminates ...... 66 B.4.6.1 Stiffnes properties ...... 66 B.4.6.2 Safety factors ...... 68 B.4.6.3 Moment of inertia of sandwich panels ...... 70 B.5 Laminate for gate design ...... 70 B.5.1 Laminate used for the bending moment ...... 71 B.5.2 Laminate used for the shear force ...... 71

C. Area analyses ...... 73 C.1 Area analysis of the Rhine brances (design level 1) ...... 73 C.2 Area analysis of the reach Nederrijn-Lek (design level 2) ...... 73 C.2.1 General description of the present situation ...... 73 C.2.2 Details of the present situation...... 74 C.2.2.1 Waterway dimensions ...... 74 C.2.2.2 River bed level ...... 74 C.2.2.3 Water levels of the Nederrijn and Lek ...... 76 C.2.2.4 Discharges ...... 77 C.2.2.5 Height restrictions ...... 78 C.2.2.6 Pumping stations and inlets ...... 79 C.2.2.7 Channels ...... 79 C.2.2.8 Navigation ...... 83 C.2.2.9 Levees ...... 85 C.2.2.10 Harbours ...... 85 C.2.2.11 Recreation ...... 87 C.2.2.12 Nature ...... 87 C.2.2.13 Agriculture ...... 88 C.3 Area analysis of the reach Wijk bij Duurstede till weir Hagestein (design hpase 3) ...... 89 C.3.1 Waterway dimensions ...... 89 C.3.2 Water levels ...... 89 C.3.3 Bed level ...... 90 C.3.4 Discharges ...... 90 C.3.5 Height restrictions ...... 91 C.3.6 Pumping stations and inlets ...... 91 C.3.7 Channels...... 91 C.3.8 Harbours ...... 91 C.3.9 Recreation ...... 92

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C.3.10 Nature...... 92 C.3.11 Agriculture...... 93 C.4 Area analysis of the weir location (design level 4) ...... 93 C.4.1 Weir location ...... 93 C.4.2 Global weir dimensions ...... 93 C.4.3 Road connections ...... 93 C.4.4 Waterway connections ...... 94 C.4.5 Soil parameters ...... 95

D. Stakeholder analysis ...... 98 D.1 Government (Rijksoverheid, 2012) ...... 98 D.2 Rijkswaterstaat (Rijkswaterstaat, 2009) ...... 98 D.3 Provinces ...... 99 D.4 Water boards ...... 99 D.5 Municipalities ...... 99 D.6 Inhabitants ...... 100 D.7 Koninklijke Schuttevaer (Koninklijke Schuttevaer, 2012) ...... 100 D.8 HISWA (HISWA, 2012) ...... 100 D.9 Utrechts Landschap (Utrechts Landschap, 2012) ...... 100 D.10 Mooi Gelderland (Mooi Gelderland, 2012) ...... 101 D.11 Hart voor natuur (Hart voor natuur, 2012) ...... 101 D.12 Enterprises ...... 101 D.13 Energy suppliers ...... 101 D.14 LTO Nederland (LTO Nederland, 2012) ...... 101 D.15 Table of actors ...... 102

E. Requirements analyses ...... 105 E.1 Requirement analysis of the reach selection (design level 1) ...... 105 E.1.1 Customer expectations (step 1) ...... 105 E.1.2 Constraints (step 2 & 3) ...... 105 E.1.3 Operational scenarios (step 4) ...... 106 E.1.4 Measures of effectiveness (step 5) ...... 106 E.1.5 System boundaries (step 6) ...... 106 E.1.6 Interfaces (step 7) ...... 106 E.1.7 Utilization environments (step 8) ...... 107 E.1.8 Life cylcle process concepts (step 9) ...... 107 E.1.9 Functional requirements (step 10) ...... 107 E.1.10 Performance requirements (step 11) ...... 108 E.1.11 Modes of operation (step 12) ...... 108 E.1.12 Technical performance measures (step 13) ...... 108 E.1.13 Physical characteristics (step 14) ...... 108 E.1.14 Human factors (step 15) ...... 108 E.2 Requirement analyisis of the configuration design (design level 2) ...... 108 E.2.1 Customer expectations (step 1) ...... 109 E.2.2 Constraints (step 2 & 3) ...... 109 E.2.3 Operational scenarios (step 4) ...... 110 E.2.3.1 Zero scenario ...... 111 E.2.3.2 Economic growth ...... 111 E.2.3.3 Economic decline ...... 112

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E.2.3.4 Wetter climate conditions ...... 112 E.2.3.5 Drier climate conditions ...... 112 E.2.4 Measures of effectiveness (step 5) ...... 113 E.2.5 System boundaries (step 6) ...... 113 E.2.6 Interfaces (step 7) ...... 113 E.2.7 Utilization Environments (step 8) ...... 114 E.2.8 Life cycle process concepts (step 9) ...... 114 E.2.9 Functional requirements (step 10) ...... 114 E.2.10 Performance requirements (step 11) ...... 115 E.2.11 Modes of operation (step 12) ...... 116 E.2.12 Technical performance measures (step 13) ...... 116 E.2.13 Physical characteristics (step 14) ...... 117 E.2.14 Human factors (step 15) ...... 117 E.3 Requirements analysis of the weir location and weir configuration (design level 3) ...... 117 E.3.1 Customer expectations (step 1) ...... 117 E.3.2 Constraints (step 2 & 3) ...... 117 E.3.3 Operational scenarios (step 4) ...... 118 E.3.4 Measures of effectiveness (step 5) ...... 118 E.3.5 System boundaries (step 6) ...... 118 E.3.6 Interfaces (step 7) ...... 118 E.3.7 Utilization environments (step 8) ...... 119 E.3.8 Life cycle process concepts (step 9) ...... 119 E.3.9 Functional requirements (step 10) ...... 119 E.3.10 Performance requirements (step 11) ...... 119 E.3.11 Modes of operation (step 12) ...... 120 E.3.12 Technical performance measures (step 13) ...... 120 E.3.13 Physical characteristics (step 14) ...... 120 E.3.14 Human factors (step 15) ...... 120 E.4 Requirements for weir design (design level 4) ...... 120 E.4.1 Customer expectations (step 1) ...... 121 E.4.2 Constraints (step 2 & 3) ...... 121 E.4.3 Operational scenarios (step 4) ...... 121 E.4.4 Measures of effectiveness (step 5) ...... 121 E.4.5 System boundaries (step 6) ...... 122 E.4.6 Interfaces (step 7) ...... 122 E.4.7 Utilization environments (step 8) ...... 122 E.4.8 Life cycle progress concepts (step 9) ...... 123 E.4.9 Functional requirements (step 10) ...... 123 E.4.10 Performance requirements (step 11) ...... 124 E.4.11 Modes of operaton (step 12) ...... 125 E.4.12 Technical performance measures (step 13) ...... 125 E.4.13 Physical characteristics (step 14) ...... 126 E.4.14 Human factors (step 15) ...... 126

F. Lists of requirements ...... 127 F.1 Reach requirements (design level 1) ...... 127 F.1.1 Functional requirements ...... 127 F.1.2 Aspect requirements ...... 127 F.1.3 Internal interface requirements ...... 128

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F.1.4 External interface requirements ...... 128 F.1.5 Realisation requirements ...... 128 F.2 Configuration requirements (design level 2) ...... 128 F.2.1 Functional requirements ...... 128 F.2.2 Aspect requirements ...... 129 F.2.3 Internal interface requirements ...... 129 F.2.4 External interface requirements ...... 129 F.2.5 Realisation requirements ...... 130 F.3 Requirements for weir location and configuration (design level 3) ...... 130 F.3.1 Functional requirements ...... 130 F.3.2 Aspect requirements ...... 130 F.3.3 Internal interface requirements ...... 130 F.3.4 External interface requirements ...... 130 F.3.5 Realisation requirements ...... 131 F.4 Requirements for weir design (design level 4) ...... 131 F.4.1 Functional requirements ...... 131 F.4.2 Aspect requirements ...... 132 F.4.3 Internal interface requirements ...... 132 F.4.4 External interface requirements ...... 132 F.4.5 Realisation requirements ...... 132

G. Functional analyses ...... 133 G.1 Functional analysis of reach selection (design level 1) ...... 133 G.1.1 System functions ...... 133 G.1.2 System objects and allocation of requirements ...... 133 G.2 Functional analysis of configuration design (design level 2) ...... 133 G.2.1 System functions ...... 134 G.2.2 System objects and allocation of requirements ...... 134 G.3 Functional analyisis of weir location and configuration (design level 3) ...... 135 G.3.1 System functions ...... 135 G.3.2 System objects and allocation of requirements ...... 136 G.4 Functional analysis of weir design (design level 4) ...... 137 G.4.1 System objects and allocation of requirements ...... 137

H. Design criteria ...... 139 H.1 Design criteria for reach selection (design level 1) ...... 139 H.2 Design criteria for configuration design (design level 2) ...... 139 H.3 Design criteria for weir location and configuration (design level 3) ...... 141

I. Redesign of the waterways ...... 142 I.1 Waal ...... 142 I.1.1 Lower discharges ...... 142 I.1.2 Equal discharge ...... 142 I.1.3 Higher discharges ...... 142 I.2 IJssel ...... 143 I.2.1 Lower discharge ...... 143 I.2.2 Equal discharge ...... 143 I.2.3 Higher discharge ...... 143 I.3 Nederrijn and Lek ...... 144

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I.3.1 Lower discharges ...... 144 I.3.2 Equal discharge ...... 144 I.3.3 Higher discharges ...... 144

J. New configuration of the Nederrijn-Lek ...... 145 J.1 First brainstorm session ...... 145 J.2 Second brainstorm session ...... 145 J.2.1 Preliminary selection of variants ...... 148 J.3 Elaboration of configuration variants ...... 151 J.3.1 Elimination of variants ...... 151 J.4 Elaboration of configuration variants ...... 152 J.4.1 “2w;Driel&Lekkanaal;com.ship” ...... 153 J.4.2 “2w;Driel&Culemborg;com.ship” ...... 155 J.4.3 “2w;Driel&Lekkanaal;recr” ...... 158 J.4.4 “2w;Driel&Culemborg;recr” ...... 160 J.4.5 “2w;Driel&Culemborg;com.ship&recr” ...... 163 J.4.6 “3w;Driel&Amerongen&Hagestein;com.ship” ...... 163 J.4.7 “3w;Driel&A-R Kanaal&Hagestein;com.ship” ...... 165 J.4.8 Overview of the amount of adaptatoin work per variant ...... 168 J.5 Multi criteria analysis for configuration design ...... 168 J.5.1 Evaluation criteria ...... 168 J.5.2 Weight reference ...... 169 J.5.3 Rating of variants ...... 170 J.5.4 Costs ...... 173 J.5.5 Score (performance/costs) ...... 176 J.6 Conclusion ...... 177 J.6.1 Description of the remaining variants...... 177

K. Verification of the horizontal water level assumption ...... 179 K.1 Minimum discharge ...... 180 K.2 Length and water levels of a dammed section...... 180

L. Cost estimation of weirs ...... 182

M. Weir location ...... 184 M.1 Location 1; weir near the Amsterdam Rijnkanaal ...... 184 M.2 Location 2; weir at the floodplains of Beusichem (Figure M-1) ...... 184 M.3 Location 3; weir in the waterway in between Culemborg and Beusichem ...... 185 M.4 Location 4; weir at the floodplains in between Culemborg and Beusichem ...... 186 M.5 Location 5; weir located in the sand pit ...... 187 M.6 Location 6; weir located at the floodplains of Culemborg ...... 187 M.7 Location 7; weir located in the river at Culemborg...... 188 M.8 Conclusion ...... 188

N. Preliminary weir design ...... 189 N.1 General weir design...... 189 N.2 Gates ...... 191 N.2.1 Horizontal roller gate ...... 191 N.2.2 Sector gate (horizontal axis) ...... 191

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N.2.3 Segment/radial gate ...... 191 N.2.4 Submerged segment/radial gate ...... 192 N.2.5 Vertical lifting gate ...... 192 N.2.6 Vertical lowering gate ...... 193 N.2.7 Flap gate (mechanically operated) ...... 193 N.2.8 Flap gate (hydraulically operated) ...... 194 N.2.9 Drum gate ...... 194 N.2.10 Inflatables ...... 195 N.2.11 Visor gate ...... 195 N.3 Variants ...... 195 N.4 Multi criteria analysis of the variants ...... 197 N.4.1 Weight reference ...... 197 N.4.2 Rating of variants ...... 197 N.4.3 Costs ...... 199 N.4.4 Score ...... 200 N.5 Conclusion ...... 201

O. Hydraulic models of the weir ...... 202 O.1 General model for the Nederrijn and Lek water levels ...... 204 O.1.1 Model description for the water levels in the Nederrijn and Lek ...... 204 O.1.2 Verification of the model for a (partial) dammed river ...... 207 O.1.3 Verification of the model for an open river ...... 208 O.1.4 Comparison with the present dam regimes ...... 209 O.1.5 Result of the hydraulic model for the Nederrijn and Lek ...... 211 O.1.6 Suitable dam regime for weir Culemborg ...... 213 O.2 Local model for the flow at the weir ...... 215 O.2.1 Model description for the water level and speed at contraction of the weir ...... 215 O.2.2 Application of the flow model for determination of the width of the weir ...... 216 O.3 Model for an overflow gate ...... 219 O.3.1 Model description for a weir with overflow gates ...... 219 O.3.1.1 Tail controlled flow ...... 221 O.3.1.2 Structure controlled flow ...... 221 O.3.1.3 Discharge coefficients ...... 222 O.3.1.4 Transition of tail and structure controlled flow ...... 222 O.3.2 Results of the hydraulic model of a weir with overflow ...... 224 O.4 Model description for a weir with underflow gates ...... 227 O.4.1 Hydraulic model based on energy and momentum equation ...... 228 O.4.2 Hydraulic model based on the RWS manual ...... 229 O.4.3 Results of the hydraulic model of a weir with underflow gates ...... 230 O.5 Accurate discharge control ...... 232 O.5.1 Underflow gate ...... 232 O.5.2 Overflow gate ...... 232 O.5.3 Conclusion ...... 233 O.6 Mathcad sheets ...... 234

P. Load definitions ...... 235 P.1 Loads at the gate ...... 235 P.1.1 Permanent loads ...... 235 P.1.2 Variable loads ...... 236

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P.1.3 Special loads ...... 236 P.2 Load factors ...... 237 P.3 Load combinations ...... 238

Q. Waves and hydrostatic pressure ...... 239 Q.1 Wave height ...... 239 Q.2 Wave pressure ...... 242 Q.3 Hydrostatic pressure ...... 243 Q.4 Load at the gates ...... 244 Q.5 Mathcad sheets ...... 245

R. Gate weight and gate costs ...... 246 R.1 Estimation of gate weight (girder on two supports) ...... 246 R.1.1 Calculation method of the gate weight ...... 247 R.1.1.1 Calculation for maximum stresses ...... 247 R.1.1.2 Calculation for maximum deflection ...... 248 R.1.2 Steel gate weight ...... 249 R.1.3 FRP gate weight ...... 250 R.1.4 Comparison of a steel and FRP gate ...... 254 R.2 Mathcad sheets ...... 255

S. Weir foundation ...... 256 S.1 Limit states...... 256 S.2 Fully dammed operation ...... 258 S.2.1 Continuous foundation slab ...... 258 S.2.2 Separate foundations for pylons and sills ...... 259 S.2.3 Comparison with the present weirs ...... 260 S.3 Conclusion ...... 261 S.4 Mathcad sheets ...... 261

T. Global weir gate design ...... 262 T.1 Rotatability of the gate ...... 262 T.2 Maximum gate deflections ...... 262 T.3 Gate geometry ...... 263 T.4 Gate cross sections ...... 264 T.5 Loads at the gate ...... 267 T.6 Mechanical scheme of the gate ...... 268 T.6.1 Mechanical scheme of the gate ...... 268 T.6.2 Deflections of the gate ...... 269 T.6.2.1 Euler-Bernoulli & Timoshenko beams ...... 269 T.6.2.2 Deflections of Timoshenko beams ...... 272 T.6.3 Verification of the behaviour of the 4th mechanic scheme ...... 276 T.6.3.1 Conclusion ...... 279 T.7 Mathcad sheets ...... 279

U. Gate design ...... 280 U.1 Global design of the gate ...... 280 U.1.1 Solid plates ...... 280 U.1.2 Sandwich panels ...... 287

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U.1.2.1 InfraCore® panels ...... 287 U.1.2.2 Adjustment to the infracore® panel cross section ...... 289 U.1.3 Dimensions of the webs for shear ...... 289 U.1.4 Check for torsion ...... 292 U.1.5 Design check of the panels ...... 294 U.1.5.1 Check for moment capacity ...... 296 U.1.5.2 Check for shear capacity ...... 296 U.1.5.3 Check for maximum deflections ...... 297 U.1.5.4 Check for combined stress capacity ...... 298 U.2 Design checks ...... 299 U.2.1 Yielding or fracture of the face in tension or compression (failure mode a) ...... 300 U.2.2 Core shear failure (failure mode b) ...... 300 U.2.3 Face wrinkling (failure mode c and d) ...... 300 U.2.4 General buckling (failure mode e) ...... 300 U.2.5 Shear crimping (failure mode f) ...... 301 U.2.6 Face dimpling (failure mode g) ...... 301 U.2.6.1 Face dimpling of a corruaged core ...... 301 U.2.6.2 Face dimpling of a honeycomb core ...... 303 U.2.6.3 Conclusion ...... 303 U.2.7 Core indentation (failure mode h) ...... 304 U.3 Plate calculations ...... 304 U.3.1 Maximum plate deflections ...... 305 U.3.2 Plate buckling ...... 307 U.3.2.1 Check for shear crimpling ...... 309 U.4 Connections ...... 309 U.4.1 Connection of the webs to the front and back wall ...... 310 U.4.2 Connections of the FRP gate to the rotation disks ...... 310 U.5 Natural frequencies ...... 311 U.5.1 Added water mass ...... 311 U.5.2 Vibrations of a one degree of freedom system (1 DOF system) ...... 312 U.5.3 Vibrations of a bending beam ...... 314 U.5.4 Verification for other hydraulic gates ...... 316 U.5.5 Conclusion ...... 316 U.6 MathCad sheets ...... 316

References ...... 318

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A. Literature study

The literature study is split up in several distinct subjects. A.1 starts with an introduction of the subject of the thesis. This helps an uninitiated person to understand the context and the goal of this literature study. An insight of the history of the Dutch Rhine branches is given in A.2. The history which is of interest in this research starts around 1800. The river management and engineering problems were in these years quite severe and people encountered many floods. ‘Serious’ engineering works started around 1850 in order to control the rivers. First a description of the rivers in the Netherlands is given. The functions of the structures which are built in the Nederrijn are described at the end of this section. The next paragraph (A.3) treats the behaviour of the present river system. The water distribution, sediment distribution, flood peaks, the present dam regime, the behaviour of the Nederrijn, and the navigation at the Nederrijn and Lek are elaborated in this section. The future situation of the Netherlands is described in A.4. A short overview of the predictions of the royal Dutch meteorological institute (KNMI) and the recommendations which are relevant for this research is given. A.5 gives a description of the weirs which are built in the Nederrijn and Lek. First the design synthesis of the present weir is described, second the changes over time of the designs of the weirs, and at last the results of the availability and reliability assessment executed by ARCADIS is described.

A.1 INTRODUCTION TO THE NEDERRIJN AND LEK

The demands population of the Netherlands is increasing each year. With the increasing population also the demands of the people living in the Netherlands are increasing. The need for (fresh) water for agriculture, shipping, desalination of the delta, and drinking water has increased drastically over the last century. To realise a better water management system several structures are built in the Netherlands. The best known structures which have been built are located in the deltaic area near the coast. The main functions of these structures are defending the hinterland from flooding and creating fresh water lakes for agriculture, desalination, and drinking water. Besides these structures, three less known structures have been realised along the Nederrijn. The weir complex near the village of Driel, shown in Figure A-1 is the most important structure. Weir Driel has to regulate the discharges over the Nederrijn and IJssel by the visor gates.

Figure A-1 Weir complex at Driel during high water from the downstream side (Arjen, 2007)

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Two other weir complexes, which have nearly the same design, are situated near the village of Hagestein and Amerongen. These weir complexes keep the water level high enough for navigation in the Nederrijn during periods of low discharge. The mean water depth of the Nederrijn created by these weirs is around 2.80 metres, which is nearly the same as the water depth in the Waal (Ministerie van V&W, 1957). In this manner, vessels are able to sail through the Nederrijn and Lek without discharge and are able to sail from a dammed river section via a lock into the other dammed river section. Nowadays not every ship class is able to sail in the Nederrijn; some restrictions in ship classes are set. An overview of ship classes which are allowed to sail on the Nederrijn and Lek is presented in A.3. The water which flows through the Nederrijn originates from the Rhine and other smaller rivers in Switzerland, France, and Germany. The Rhine enters the Netherlands near the village of Lobith as shown in Figure A-2. The river bifurcates at Pannerdensche kop (encircled with number 1 in the figure) into the river Waal and the Pannerdensch Kanaal after at couple of kilometres downstream from the border. The Waal runs towards the city of Rotterdam and the deltaic area of Zeeland and changes name into the Boven-Merwede. The Boven-Merwede bifurcates into the Beneden-Merwedes and Nieuwe Merwede at circle 3. The Pannerdensch Kanaal changes name into Nederrijn at the old connection of the old Rhine which is located at the present pumping station Kandia; indicated by the red line in between circle 1 and circle 2 of Figure A-2. The IJssel branches of from the Nederrijn at the IJsselkop (encircled at number 2) and flows towards the north. The Nederrijn flows towards the west and changes name into the Lek at the crossing of the Amsterdam-Rijnkanaal. The Lek joins the other reaches of the river again. The water is discharged by the Nieuwe Waterweg which is the only open connection to the Noordzee. The southern part of the river system, the Nieuwe Merwede, flows into the Haringvliet, which is connected to the North Sea by sluices. Also the Maas which is part of another river system flows into the Haringvliet.

Figure A-2 Location of the three main bifurcations in the Netherlands: 1) Pannerdensche Kop, 2) IJsselkop and 3) Merwedes. The weirs on the Nederrijn are shown in orange; from right to left: Driel, Amerongen, and Hagestein (Rijkswaterstaat, 2006)

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The visor gates are fully closed for discharges lower than 1400 m3/s at Lobith (Rijkswaterstaat, 2006). A minimum discharge of 25 m3/s for the Nederrijn has been set in order to refresh the waterway and preventing a reduced water quality in the Nederrijn (Til, 1961). The visor gates of weir Driel are party closed for discharges higher than 1400m3/s and lower than 2300m3/s. Navigation passes the weirs by locks which are built parallel to the complexes. The gates are fully lifted when the discharge at Lobith exceeds 2300 m3/s. In this case the water is able to freely flow from upstream to the downstream end of the river without nearly any hindrance of the structures. The vessels are able to sail underneath the visor gates if their height is equal or lower than 9.10 metres for the highest known water level (Rijkswaterstaat, n.d.).

A.2 HISTORY

The history of the water system of the Netherlands which is described in this section starts around 1800. A lot of small water boards were already present before this year. Every water board arranged his own water protection measures and had his own ‘law’ and funding. A major project which took place before 1800 is the building of the Pannerdensch Kanaal in 1707. A stable distribution of water over the Rhine branches and the IJssel was created by excavating this channel and the Bijlands Kanaal. This project was the start of the ‘modern’ river management. The formation of the Dutch rivers and the development of the project plans of the canalization of the Nederrijn are fully described in this appendix.

A.2.1 THE RHINE BRANCHES

This section is based on: (Woud, 2006), (Heezik, 2006)

Before 1850 people thought that they were not powerful enough to control the rivers, because rivers were ‘uncontrollable systems’ and they had to ‘live their own life’ with their extreme behaviour. Furthermore they thought that they were not able to calculate and predict the behaviour of the river. Therefore, they went into a defensive river engineering strategy. They excavated navigational channels along the rivers, because they were able to regulate these channels. Furthermore they realised a lot of spillways along the river to relief the rivers during high waters. A well-known spillway is the Beerse overlaat, presented in Figure A-3. The area which was flooded by the spillway is indicated in blue. The rivers had insufficient capacity to ‘transport’ the water and ice towards the sea during high river runoff which resulted in planned (overflow of a spillway) and unplanned floods (breached levees). Major floods occurred in 1809, 1820, 1855 and 1861. The cause of these major floods was the bad shape of the river system and the lack of river mouths. Some shortcomings of the river system causing these floods were: too high summer levees, a lot of (small) islands situated in the summer beds, large sandbars in the river, and trees which were planted in the floodplains which caused high roughness of the river bed. Furthermore, the Maas and the Waal joined each other near Loevestein. The water of the Maas was not able to freely flow towards sea due to the high water levels of the Waal during spring. Levees along the Maas became too wet and failed as a result of these high and long lasting water levels. Moreover, the connection of the combined Maas-Waal water from Loevestein towards the sea was very poor; the Biesbosch sucked too much water into the Hollands Diep which resulted in low water levels of the Boven-Merwede which hampered navigation. The levees were also in a bad shape; they were too low, too narrow, and too steep. The Noorder Lekdijk near Wijk bij Duurstede was the most unreliable levee. It was founded on a weak soil layer and a layer of quick sand and could fail at any moment. Half of the province of Utrecht and large parts of Holland would be flooded if this levee would have failed.

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Committees were founded to prevent new major floods. The first committee was the ‘Comité central du Waterstaat.’ They advised to dam off the Pannerdensch Kanaal and to excavate a new IJssel to relief the Lek. In this way, they wanted to prevent the Noorder Lekdijk from failing. The drawback of this solution is an increase of flow in the Waal and the IJssel. Therefore, a new river branch would be excavated between Heusden and the Hollands Diep to counteract this drawback. The second committee, installed by King Willem I in 1825, advised to enlarge the Baartwijkse overlaat near Heusden, which is indicated in Figure A-3, and to make a river map. A third committee was installed in 1828 and their report was ready in 1849. This committee advised the excavation of a new Merwede.

Figure A-3 The old Rhine branches and the spillways (Woud, 2006)

Ferrand and Van der Kun, two inspectors of the Waterstaat, compared the recommendations of the committees. From this comparison they concluded that the river should be able to transport the water and ice by itself and not by using spillways. A normalised width of the rivers was set to fulfil this. The river would erode the islands and sandbars away by implementing this normalised width. Furthermore, they advised to excavate a ‘Nieuwe Merwede’ for improving, water transport capacity of the river system, transport capacity of ice, and the navigation of Rotterdam and Dordrecht. Navigation was of minor importance despite the political pressure of Germany. Introducing a safe river system in the Netherland was the aim of their work. The benefits for navigation were just ‘side effects.’

Difference between normalisation and canalization of rivers Two methods are available for increasing the depth of a river namely normalisation and canalization. The width of the river is restricted by groins in order to force the river to deepen itself by scour for normalisation. The total wet cross section remains equal but the width is smaller and the depth is larger with respect to the old situation. The depth of the river can be increased by canalization when the depth remains too small after normalisation. Water is retained by weirs in order to generate higher water levels and larger depths in the waterway.

Normalisation and excavation of the rivers The river retaining works started in 1850. The first project was the excavation of the ‘Nieuwe Merwede’ which is indicated in Figure A-4. People excavated channels which would be enlarged by scour changing the river into a sufficient waterway. The river section upstream of the Nieuwe Merwede was called ‘Boven Merwede’ and the river section downstream of the Nieuwe Merwede was called ‘Beneden Merwede’. Building of the Nieuwe Merwede was harder than expected due to thick clay layers. Dredging equipment had to be used to remove the clay. The Nieuwe Merwede was finished after 40 years of working. The capacity of the Nieuwe Merwede was large enough to transport the discharge of the Maas and Waal

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Appendices New canalisation of the Nederrijn and Lek

towards sea. But the water depth of the Boven Merwede remained too low for navigation; the Boven Merwede had to be normalised to increase the water depth. But first the Maas had to be disconnected in to be able to normalise the Boven Merwede. Therefore, a separate river mouth called the Bergse Maas was excavated. The Boven Merwede could be normalised after the realisation of the Bergsche Maas.

Figure A-4 River branches in the western river delta

Ferrand and Van der Kun indicated a lot of islands and sandbars in the Waal which had to be removed during the normalisation project. The river was forced to follow a narrower river section by restricting the river with groins. The increased flow eroded the sandbars and islands away which is shown in Figure A-5. Furthermore, they closed off the Sint Andries channel with a lock to split the Maas and the Waal. This first phase of normalisation was not enough; the river had to be restricted even more. Larger vessels were sailing on the river and the vessels needed a larger draught. The normalisation was finished in 1916. A three metre deep ‘river’ from the border of Germany to the city of Dordrecht was created with a sufficient transport capacity for water and ice.

Figure A-5 Normalisation of the Waal (Heezik, 2006)

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The improvements of the Nederrijn and Lek took place between 1850 and 1896. Two bends were cut off and 62 kilometres of groins were constructed. The depth of the main shipping lane was more than 2 metres after normalisation. One of the last normalisation works of the Rhine branches is the waterway from Dordrecht towards Rotterdam. No discharge was available, so it was not possible to normalise this waterway with the above described method. Therefore, they dredged a 3.5 metre deep waterway, called the Dordtse Kil, to realise sufficient water depth for navigation. Furthermore, the waterway the Noord (running from Krimpen towards Rotterdam) was improved during the construction of the Nieuwe Waterweg. P. Caland, the project leader of the Nieuwe Waterweg project, tried to establish the Nieuwe Waterweg by natural scour. First a small channel was excavated from sea towards Rotterdam near the village of Hoek van Holland. They tried to maintain a certain depth with use of natural scour, but decided to use dredging equipment after 14 years of effort. The Nieuwe Waterweg was finished in 1872; a channel between the Noordzee and the harbour of Rotterdam had been realised. All the major rivers were successfully normalised except the IJssel. Rijkswaterstaat [RWS] tried unsuccessfully to normalize the IJssel for 15 years long. The only solution for improving the navigability of the IJssel was canalization in their opinion. Details about the canalization of the IJssel are described in paragraph A.2.2 and A.2.3.

A.2.2 CANALIZATION OF THE RHINE BRANCHES

The significance of the improvement of the depth of the IJssel appears from the recorded water depth of the years 1947, 1949, 1953, and 1959. These years showed some very limited water levels. The permitted draught of vessels was lower than 1.5 metre in September and October 1959. The lowest draught which was recorded was 0.80 metres for this year as presented in Figure A-6. Statistics from 1901-1950 showed a water depth of less than 1.5 metres during 30 days in an average year. Vessels encountered a lot of hindrance due to the low water levels every 4 years for several months (Til, 1961). So, limited depths were no exception before canalization

Figure A-6 Allowable draught for ships between 400 and 2000 tons on the Boven IJssel during the second half-year of 1959 (Til, 1961)

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Appendices New canalisation of the Nederrijn and Lek

‘Studiedienst Bovenrivieren’ made a project plan for improving the water depth of the IJssel by canalizing the IJssel in 1938. This canalization should improve the depth of the IJssel during periods of low waters. The weirs at the IJssel had to retain the water in order to guarantee a sufficient depth when the inflow at Lobith was insufficient. The vessels would pass the weirs by locks which would be located parallel to the weirs. It would possible for navigation to sail from the Waal towards the IJsselmeer with limited hindrance during periods of low discharge after realisation of the canalization. Besides this main goals, there were also secondary goals which had to be fulfilled, which were:  Getting more fresh water into the Lek for drinking water purposes in the Western part of the Netherlands.  Getting more fresh water into the Lek to counteract the salt intrusion from sea.

Fresh water would be blocked by the weirs from the IJsselmeer, so less water could temporally be stored due to the blockage of the new weirs during droughts. To counteract this drawback; the weirs at the IJssel would be opened if the water level of the IJsselmeer dropped below -0.25m NAP to supply the IJsselmeer. This plan was handed in at the ‘Tweede Kamer’ for approval, but was not approved due to outbreak of the Second World War (Rijkswaterstaat; Directie bovenrivieren, 1979). During the Second World War another plan for improving the water depth of the IJssel was developed. This plan was focused on the canalization of the Nederrijn instead of the canalization of the IJssel. The main goals of the canalization of the Nederrijn were (Blokland, 1957):  Improving the navigability on the IJssel, Lek, and Pannerdensch Kanaal.  Improving the fresh water supply of the Northern part of the Netherlands by diverting more fresh water into the IJssel.

Prof. dr. ir. J. Th. Thijsse, (Professor of Hydraulics of the Technical University of Delft and director of ‘het Waterloopkundig Laboratorium’) suggested to execute this plan but encountered much opposition. Salinization would cause a serious problem when a certain quantity of water was being blocked from the South-Western part of the Netherlands. Ir. J. van Veen stated to close off some estuaries to counteract this side effect. In this manner, fresh water lakes would be created and the discharge through the remaining river mouth (Nieuwe Waterweg) would increase which would counteract the salinization. Executing this plan was not possible due to political-social circumstances. However, money was available and the political-social circumstances had been improved after the floods of 1953. The ‘delta plan’ started and some estuaries in Zeeland were closed. The drawback of salinization caused by the canalization of the Nederrijn was solved by the structures which were built in the context of the ‘delta project’ (Heezik, 2006).

The Dutch Parliament decided to implement the canalization of the Nederrijn. The benefits of the canalization of the Nederrijn (undisturbed flow of fresh water into the IJsselmeer) were larger than the benefits of the canalization of the IJssel (counteracting the salt intrusion in the western part of the Netherlands) (Ministerie van V&W, 1957). Moreover, inland navigation which navigates towards or from the harbour of Rotterdam would not be hampered. Because they navigate on the Waal (parallel to the Nederrijn), which is a much better waterway. Realising weirs at the IJssel would cause more hindrance for navigation with respect to the canalization of the Nederrijn. Only one route is available towards the North so vessels had to use the locks in the IJssel during periods of low discharge. The usage of these locks would cause delay and would raise the shipping costs which are a major drawback. In fact, canalization of the Nederrijn is the only option to fulfil both goals simultaneously (increasing the water depth on the Boven-IJssel and increasing the flow of fresh water towards the northern part of the Netherlands). Therefore, they chose for building the weirs in the Nederrijn to increase the rate of flow of the IJssel and to increase the navigability of the Nederrijn, Lek, IJssel, and Pannerdensch Kanaal. (Til, 1961).

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A.2.3 THE CANALIZATION PROJECT OF THE NEDERRIJN (1960)

The upstream weir which had to be built should divert water into the IJssel during periods of low discharge. The water should be able to flow through the Nederrijn for periods with normal or large discharges. Influence of the upstream weir to the discharge of the Waal and other river branches was not allowed. This could result in lower water levels of the Waal which would hamper the navigation and the flow of fresh water towards the southern part of the Netherlands. To prevent this from happening, some restrictions, side effects, and boundary conditions were set for the canalization project (Rijkswaterstaat, n.d.): 1. A decrease of the discharge of the Waal was not allowed during periods of low discharge. A decrease would result in a decrease in depth. This would hamper the navigation in the Waal. 2. The dimensions and configuration of the Pannerdensch Kanaal, the Nederrijn-Lek, IJssel and the bifurcation points of Pannerden and the IJsselkop had to be changed due to the first boundary condition. 3. The weir at Driel would cause a higher water level along the IJssel which would improve the navigation on this river. Therefore, more vessels would sail on the IJssel after realisation of the weirs. The river has to be adapted for this increase in navigation. 4. The river bed of Pannerdensch Kanaal, the Nederrijn, and the IJssel had be lowered to counteract the backwater effect caused by weir Driel. 5. The highest known discharge of the Nederrijn and IJssel are not allowed to increase. An increase of discharge would be caused by the lowering of the river bed of the Pannerdensch Kanaal. 6. The weir should be out of service for high waters. 7. Navigation on the Nederrijn and Lek had be possible during low water. Other complexes than weir Driel had to provide a water depth of 2.80 metre. 8. A minimum discharge of 50 m3/s had be present during low discharges. This minimum discharge is needed for preventing pollution of the river water. Extra water must be provided from the Amsterdam-Rijnkanaal if the minimum discharge is not met due to water extraction along the Nederrijn. 9. The intersection of the Amsterdam-Rijnkanaal should not be negatively affected for the new situation. 10. Planned changes had be feasible in the long term. Unacceptable changes of the bottom profile are not allowed. 11. Fish migration had to be possible. 12. The structures had to meet the navigational requirements and navigation had to encounter a minimum of hindrance during and after the execution of the structures. 13. Damage caused by the backwater effect must be as limited as possible.

Weir Driel is one of the control instruments for the distribution of water in the Netherlands. This weir is able to divert water into the IJssel (a minimum increase of 27% and a maximum of 44% compared to the previous situation) (Heezik, 2006). It was hard to determine the location of the most upstream weir. It was beneficial to place the upstream weir as far as possible from the bifurcation. This would limit the amount of weirs in the Nederrijn to three. On the other hand, it was hard to control the discharge distribution of the IJssel and Nederrijn if the weir had been placed too far away from the bifurcation. This would cause (economical) damage caused on the floodplains by the backwater effect. By executing river models RWS decided to realise the upstream weir at the floodplains of the Nederrijn near the village of Driel. The distance between the weir and the bifurcation point Pannerdensche Kop is 23 kilometres which is measured along the river axis. This short distance caused back water effects at the Pannerdensche Kop, so not only water was diverted towards the IJssel but also towards the Waal. The river bed of the Nederrijn,

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Appendices New canalisation of the Nederrijn and Lek

IJssel, and Pannerdensch Kanaal had to be lowered to counteract this undesired effect. Furthermore some bends had been cut off of the IJssel near Doesburg and Rheden in order to increase the water demand of the IJssel. (Ministerie van V&W, 1990). This resulted in a decrease of river length of 8 kilometres. So more water was attracted to flow into the IJssel and not into the Waal (Ham, 1999). A second weirs was needed to create sufficient water depth in the Nederrijn for navigation. The designers chose for the location in between Amerongen and Wijk bij Duurstede after detailed model tests. This weir is able to dam the river without severe damage caused by the backwater effect. Also the 9th boundary condition (the intersection of the Amsterdam-Rijnkanaal may not be negatively affected in the new situation) has been fulfilled with this choice of location. The third and last weir has been projected at the upstream side of the Lekkanaal. The longitudinal bed profiles are presented in Figure A-7. The longitudinal distance along the river from the IJsselkop (km914) to Driel (km 891) is 23 kilometres; the longitudinal distance from weir Driel to weir Amerongen (km 922) is 31 kilometres and the distance from weir Amerongen to weir Hagestein (km 947) is 23 kilometres (Ministerie van V&W, 1967). The weirs were built within 10 years; the first weir was completed in 1960, the second in 1965, and third in 1970. The opening of the canalization of the Nederrijn took place in 1970.For realising sufficient draught, triangles had to be dredged from the bottom profile as presented in Figure A-8. Maintenance dredging is necessary for the bed level regulations at these triangles.

Figure A-7 Longitudinal profile of the Nederrijn (Til, 1961)

: ARCADIS & TUDelft 9

Figure A-8 Designed bed levels (Gaay & Blokland, 1970)

A.2.4 DAM REGIME

Three dam regimes were elaborated for the canalization, namely: dam regime “250,” dam regime “350,” and dam regime “400.” The discharge through the Nederrijn and Lek would always be higher than 50 m3/s in order to guarantee a sufficient water quality. Dam regime “250” was a lower limit determined by the needs of the navigation. The discharge through the Pannerdensch Kanaal minus the minimum discharge of weir Driel was fully diverted into the IJssel when the total discharge of the Pannerdensch Kanaal was lower than 300 m3/s. The discharge of weir Driel would increase to keep the discharge in the IJssel stable at 250 m3/s when the total discharge of the Pannerdensch Kanaal exceeds 300 m3/s. The minimum discharge of 250 m3/s would not be met 37 days a year according Figure A-9. The discharge through Driel would be 50 m3/s until the discharge of 350 m3/s at the IJssel was realised if dam regime “350” was in operation. This higher discharge of the IJssel would be necessary when more water was needed in the IJsselmeer. This dam-regime would not cause any damage on the floodplains of the IJssel. Dam regime “350” will not be met during 120 days a year according to Figure A-7. Dam regime “400” would be applied when even more water was needed in the IJsselmeer in a short period of time. This dame regime would cause damage on the long term, so dam regime “400” is a maximum dam regime applicable for emergency situations (Rijkswaterstaat, 1957)& (Til, 1961). The dam regimes “250,” and “350” are presented on Figure A-9. The water levels at Driel, Amerongen, and Hagestein are shown in this graphs. Furthermore the discharges of the IJssel and Nederrijn are presented in these graphs.

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Appendices New canalisation of the Nederrijn and Lek

Figure A-9 Dam regimes and discharge on the IJssel, Nederrijn, and Boven-Rijn (discharge at Lobith) (Vriend, et al., 2011)

A.3 PRESENT WATER SYSTEM

The water system of the Netherlands is described in this chapter. Firstly, the distribution of water is described A.3.1, secondly the distribution of sediments in A.3.2, thirdly the present dam regime, fourthly the salinization of the delta, and at last the autonomous bed decline A.3.4.

A.3.1 WATER MANAGEMENT IN THE NETHERLANDS

A.3.1.1 THE FLOW OF WATER THROUGH THE NETHERLANDS

The water which flows through the Nederrijn originates from precipitation and the melting of snow and ice in Switzerland, France, Luxemburg, and Germany. The water from this catchment area flows through small river branches and some major river branches like the Main, Moezel, and Ruhr into the Rhine. High river runoffs are caused by: a long period of precipitation, a saturated or frozen soil, and melting water. In this case, water cannot be stored in the subsoil and flows directly into the river resulting in a high discharge of the Rhine. A flood wave originated in Southern Germany ‘travels’ five days for reaching Lobith; from Lobith towards sea takes two more days (Rijkswaterstaat, 2007). The Rhine enters the Netherlands near the village of Lobith. After a couple of kilometres the river bifurcates at Pannerdensche kop into the river Waal and the Pannerdensch Kanaal. The Waal runs towards the city of Rotterdam and the deltaic area of Zeeland and bifurcates into the Merwedes. The Pannerdensch

: ARCADIS & TUDelft 11

Kanaal changes name into Nederrijn at Pumping station Kandia which is built in the old river bed of the dammed ‘Oude Rijn.’ The IJssel branches of from the Nederrijn at the IJsselkop. The Nederrijn changes name again into the Lek at the crossing of the Amsterdam-Rijnkanaal. Further downstream the Lek confluences with the Noord and the combined river flows towards sea. The IJssel flows northward into the IJsselmeer where the water is stored. The Rhine water is able to ‘leave’ the Netherlands at five locations, which are: two discharge sluices located in the Afsluitdijk, a discharge sluice located near IJmuiden, the Nieuwe Waterweg which is the only open connection with sea, and the Haringvliet sluices. The sluices are being used to manage the amount of discharge to sea. The gates of the sluices are closed during periods of low discharge to keep the fresh water in the Netherlands for shipping, agriculture, desalination, and drinking water purposes. Sluices are opened to let the water flow into sea when the Rhine discharge becomes higher. These sluices act as the ‘valves’ of the Netherlands. Besides these external valves also an important internal valve is present, which is weir complex Driel. This weir is able to control the discharge of the IJssel, Nederrijn, and Lek till a discharge of 2300 m3/s. The main valves are indicated with red dots in Figure A-10. Also two less important valves are present at the Nederrijn which are weir Amerongen and weir Hagestein. These valves regulate the water levels in between weir Driel and weir Amerongen and in between weir Amerongen and weir Hagestein. These water levels have to be maintained at specified levels in order to guarantee a certain draught for commercial shipping and sufficient water levels for intakes, harbours, etc. Also other, less important, valves are present in the Netherlands. These have been indicated in this figure with green and blue dots (Rijkswaterstaat, 2011) and are beyond the scope of the graduation research.

Figure A-10 The valves of the Netherlands (red dots are a main valves, blue dots are secondary valves, and green dots are supply valves for regional waters)

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Appendices New canalisation of the Nederrijn and Lek

A.3.1.2 PRESENT DAM REGIME OF THE NEDERRIJN AND LEK

Before the water has reached the ‘external valves’ of the Netherlands it has been distributed over the IJssel, Nederrijn and the Waal. Two distinct situations at the Nederrijn are present, during dry periods and wet periods. The Nederrijn is a fully dammed river for dry periods (discharge at Lobith is lower than 1400m3/s) and a free flowing river during wet periods (discharge is higher than 3000 m3/s). A minimum discharge in the Nederrijn of 25 m3/s has been introduced to prevent a reduced water quality. This discharge is regulated by cylinder valves installed in the middle pier of the weirs. The minimum discharge is regulated until the discharge of the IJssel equals 285 m3/s. The visor gates are gradually lifted in order to keep the discharge of the IJssel stable and to increase the discharge of the Nederrijn. The visor gates at Driel are fully lifted when a discharge at Lobith of 2300 m3/s has been met. The gates of Amerongen and Hagestein are fully lifted when the discharge at Lobith exceeds 3000 m3/s. The Nederrijn is a free flowing river from this discharge. This discharge is realised for a water level of +8,90m NAP at the IJsselkop (ARCADIS, 2010). The minimum discharge of 285 m3/s is not always met; the discharge of the IJssel is lower when the water supply is not sufficient. First the minimum discharge of 25 m3/s must be present in the Nederrijn before more water is diverted into the IJssel. Low discharges than 285m3/s in the IJssel are shown in Figure A-11. A distinction has been made in the period before the canalization and the period after the canalization of the Nederrijn. From this graph can be concluded that the average water levels of the IJssel are increased with one metre after canalization of the Nederrijn; the lowest recorded water level after canalization is 1.50 metres which is 1 metre higher than the lowest recorded water level before the canalization.

Figure A-11 Lowest discharges of the IJssel, before and after the canalization of the Nederrijn

A.3.1.3 WATER LEVELS FOR THE NEDERRIJN AND LEK

The water levels and discharges are recorded at several measuring stations. Relevant data is available at the IJsselkop, Pannerdensche kop, several villages along the Nederrijn, and the upstream and downstream side of Amerongen, Hagestein, and Driel. The following specific data is present in tables:  the highest known discharge and the highest water level  the discharge and water level with a exceedance frequency of: o 1/1250 per year o 1/100 per year o 1/10 per year o 1/2 per year

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o 1 per year  average water discharge  average summer discharge  lowest allowable discharge (Dutch: Overeengekomen Lage Afvoer)  lowest known discharge.

The corresponding values are only published in this report when they are needed. For now they are available at the ARCADIS database and on the website of Rijkswaterstaat (Rijkswaterstaat, 2012).

A.3.1.4 SAFETY LEVELS OF THE LEVEES IN THE NETHERLANDS

To defend ourselves against the high water, legislation has been made for the probability of flooding for specific areas. The Netherlands is split up in several levee ring areas. Every levee ring has a specific probability of exceedance per year which were determined after the flood of 1953. The hydraulic structures belonging to a specific levee ring have to retain corresponding water level. This probability varies from 1/10.000 per year for the orange part of the map shown in Figure A-12, 1/4000 per year for the beige part, 1/2000 per year for the yellow part, and 1/250 for the purple part.

Figure A-12 Safety levels of the Netherlands per levee ring area

A.3.1.5 CHANGE OF THE MAXIMUM DESIGN DISCHARGES

The design of the flood defences and river improvements are based on a discharge with a probability of exceedance of 1/1250; this is the so called governing discharge. The governing discharged used to be 15.000 m3/s but the floods of 1993 and 1995 had a major impact on the statistics. The governing discharge of Lobith was raised to 16.000 m3/s in order to comply with the probability of exceedance of 1/1250 per year.

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Appendices New canalisation of the Nederrijn and Lek

Furthermore, experts expect an increase of temperature for the coming century. The KNMI made three different scenarios, namely a minimum, median, and maximum scenario which are presented in Table A-1. The climate scenarios are further described in A.4.

Table A-1 Increase in temperatures and the effect on the discharge according to the KNMI

Temperature Temperature Discharge in 2100 Scenario increase 2050 [0] increase 2100 [0] [m3/s] minimum +1 +1 16.800 median +1 +2 17.600 maximum +2 - +3 +4 - +6 19.200

The increase in temperature results in more precipitation during winter and causes a higher runoff. The Rhine acts more and more like a precipitation river and less like a combined melting water and precipitation river. The increased discharge at Lobith is shown in the last column of Table A-1. It remains doubtful whether these runoffs are able to enter the Netherlands. From the report ‘Ministerium für Umwelt und Naturschutz, Landwirtschaft und Verbraucherschutz des Landes Nordrhein-Westfalen et al. (2004) can be concluded that the rainfalls in the catchment areas are able to generate a runoff peak of 19.000 m3/s. But this runoff peak is not able to arrive in the Netherlands because it first causes floods along the Ober- and Niederrhein in Germany. Water is stored and the flood peak arriving in the Netherlands is lower compared to the flood peak in Germany. So, large amounts of water would be temporally stored in Germany, and the flood peaks which would enter the Netherlands would ‘only’ be 15.500 m3/s. The German authorities want to reduce the risk of flooding and introduced a levee reinforcement program which will be finished in 2015. The maximum runoff capacity of the river section upstream from Düsseldorf will be increased to 15.000 m3/s and of the river section Düsseldorf-Lobith 17.500 - 18.000 m3/s. The governing discharge capacity of Germany is a kind of guarantee for the maximum discharge in the Netherlands. The levees in Germany would fail and the height of the flood peak will be reduced if a flood peak occurs which is higher than 18.000 m3/s in the future (Hermeling, 2004). The project Room for the River (Dutch: Ruimte voor de Rivier) has been set up to implement the future discharges for the present river system and to improve the spatial quality of the river area. This project has to increase the maximum discharge from 15.000 m3/s up to 16.000 m3/s of the Rhine branches by implementing river-widening solutions. Levee improvements are only executed when no other solutions are available or cost effective. Solutions are already implemented for the 18.000 m3/s scenario if it is possible and cost-effective. Furthermore inner dike areas are reserved for possible river widening works for the 18.000 m3/s scenario. The project Room for the River has to be finished in 2015. The projects which are or will be executed in and around the Nederrijn-Lek are presented in Figure A-13. The projects executed along the Nederrijn are mainly floodplain excavations. The projects along the Lek are mainly levee improvements which are the best solution due to the constricted river profile. Discharges at Lobith higher than the 16.000 m3/s scenario have to be transported in the Waal and the IJssel and not in the Nederrijn due to the physical constraints along the Lek. More extensive river improvements and levee reinforcement are not possible due to the lack of space and the high costs of river improvement works. So the discharge of the Nederrijn and Lek has reached its maximum for the 16.000 m3/s scenario. The extra water which flows through the river in the 18.000 m3/s scenario has to flow through the Waal and the IJssel. Bifurcations have to be modified to relief the Nederrijn and to divert the water towards these branches. The present maximum discharge through the Nederrijn is 3.165 m3/s and will be 3.376 m3/s after the completion of the project Room for the River (Ministerie van V&W, 2007) & (Rijkswaterstaat, 2009).

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Figure A-13 Project Room for the River

A.3.1.6 GOVERNING HIGH WATER LEVELS

The design discharge of 16.000 m3/s at Lobith is used to determine the design water levels and the distribution of flow at the bifurcations. These water levels were determined with use of hydraulic models. The results per dike ring area are presented in ‘Hydraulische Randvoorwaarden 2006.’ The levee rings 15 (Lopiker- en Krimpenerwaard), 16 (Alblasserwaard en de Vijfheerenlanden) 43 (Betuwe/Tieler- en Culemobrgerwaard), 44 (Kromme Rijn), 45 (Gelderse Valei), and 47 (Arnhemse- en Velpsebroek) are important for this research (Rijkswaterstaat, 2007). The values of the design heights of these dike rings are not presented in this report. The distribution of water between the Nederrijn and IJssel can be controlled by a non-active controllable hydraulic structure called the Hondsbroekse Pleij which is completed at January the 30th of 2012. A high water channel was excavated and a hydraulic structure was built in this channel. The structure is a concrete dam with 30 openings with four concrete gates per opening which are placed on top of each other. A hydraulic crane is able to remove gates to regulate the discharge through the channel before the flood peak has arrived.

A.3.2 SEDIMENT TRANSPORT

The sediments entering the Netherlands at Lobith are divided over the Rhine branches. A distinction has to be made between the transport during low and high discharges. Only the upper grains of the bed are being transported during low discharge. The thickness of the transport layer is just one or a couple of grain diameters. The thickness of the transport layer is about one metre during a period of high discharge (measured during the high water of 1995). The flat bed changes into a bottom with consist of ripples and dunes. These ripples and dunes move through the river and transport the sediments. The transport layer is also called the active layer and the thickness is defined as the half of the height of a dune from top to trough. The sediment present in the active layer represents morphological properties of the river. The silts are part of the suspended sediment load. It is transported through the river branches and settles down at the downstream end of the river. The bed transport characteristics of the Rhine branches are elaborated in A.3.2.1 and the distribution of the silts in A.3.2.2.

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A.3.2.1 BED TRANSPORT

The size and the quantity of sediments is not uniform over the full cross section of the river. Some parts of the cross section transports coarse sediments and other parts transport the fine sediment. The cause of this distribution of sediments over the cross section is the meandering characteristics of the river. The flow of water at the bottom is directed towards the inner bend of the river due to the spiral flow. This flow transports the sediments towards the inner bend of the river. The impact of this effect depends on the sediment characteristics. Fine sediments are suspended over the water column and are not affected by the spiral flow. Coarse sediments are too heavy to be transported ‘uphill’ towards the inner bend. They roll down towards to the outer bend which is deeper than the inner bend. The spiral flow is strong enough to transport the medium grained sediments ‘uphill’ so the medium grained sediments are transported in the inner bend. So, the sediments are being sorted over the cross section. Coarse sediments are transported in the outer bend and fine sediments are being transported in the inner bend. This phenomenon has an impact on the distribution of sediments over different branches when the bifurcation of a river is located in a bend like the bifurcation of the Pannerdensche Kop and the IJsselkop. The connection of the Pannerdensch Kanaal at the Pannerdensche Kop is located at the outer bend and the connection of the IJssel at the IJsselkop is also located in the outer bend of the river. So the coarse sediments are first transported into the Pannerdensch Kanaal and then transported into the IJssel. The size of the sediments in the Waal and Nederrijn are smaller (a diameter between 1mm and 2mm) compared to the diameter of the sediments in the Pannerdensch Kanaal and IJssel (a diameter between 5mm and 8mm) as presented in Figure A-14.

Figure A-14 Sediment size [mm] at bifurcation Pannerdensche Kop (left) and IJsselkop (right)

Another cause of the distribution of sediments at bifurcation is an asymmetrical approach condition like the configuration of the Pannerdensch Kop which is shown in Figure A-15. The distribution of bed transport is not proportionally with the distribution of water. 88% (±5%) of the bed transport is transported into the Waal and 12% (±5%) is transported into the Pannerdensch Kanaal. This difference can be explained with use of the right image of Figure A-15. The river dunes (which represent the transport the sediment) continue their way into the Waal and not into the Pannerdensch Kanaal due to the asymmetrical approach. So more sediment is transported in the Waal and less sediments in the Pannerdensch Kanaal (Brinke, et al., 2001).

: ARCADIS & TUDelft 17

Figure A-15 Bifurcation point of Pannerden (left) and the dunes during high water at the bifurcation (right)

The division of sediments over the Nederrijn and IJssel is more complicated due to the presence of the weirs. A distinction is made between a free flowing Nederrijn during high discharge and a dammed Nederrijn.

The weirs of the Nederrijn are fully lifted during a high discharge, so the water is able to flow freely into the Nederrijn. The river runoff develops dunes in the bed of the Pannerdensch Kanaal. The dunes continue their way into the open Nederrijn due to the asymmetrical shape of the bifurcation. 89% of the bed transport of the Pannerdensch Kanaal enters the Nederrijn and 11% enters the IJssel. This division is not equal to the distribution of water over the Nederrijn and IJssel. The situation is different during low discharge. The bed transport which enters the Pannerdensch Kanaal is transported into the IJssel. The amount of sediment is low, just a very small percentage of the transport during high water. The distribution during high en low water is shown in Figure A-16 (Rijkswaterstaat, 2006). (Brinke, 2004).

Figure A-16 Sediment distribution during the high waters of January 2004 and November 2002 and low water of September 2004

A.3.2.2 SUSPENDED SEDIMENT TRANSPORT

The amount of silts is about three times as much as the amount of coarse sediment which enters the Netherlands but is of minor importance with respect to the bed transport. The silt is transported from Germany towards the river mouth located near Rotterdam and settles down due to the decreased velocity and the increased salinity level. The amount of transported silts is compared with the amount of transported sand per year in Figure A-17. The dimension ton has been used because the density of silt is highly dependent on the moisture content.

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Figure A-17 A) Sand transport B) and silt transport

The silt distribution over the cross section is not affected by the spiral flow and is equally distributed over the cross section. So, the distribution of silts over the Waal and Pannerdensch Kanaal are equal to the distribution of the water over the river branches. The weirs of the Nederrijn influence the distribution of the silt over the Nederrijn and IJssel. The distribution of silt over the IJssel and Nederrijn are equal to the distribution of the water when the Nederrijn is a free flowing river. A small amount of transport was measured for a dammed Nederrijn due to the minimum flushing discharge. A part of the suspended sediments floats into the Nederrijn and settles down due to the low flow velocity. The characteristics of this sedimentation of silts at the weirs are shown in Figure A-18. The fraction of fines is larger on the upstream side of the weirs due to the limited velocity. A fraction of the silt is suspended again when the weirs are opened. Also the configuration of the more coarse sediment is interesting. The fraction of gravel in the Nederrijn is larger at the upstream part of the Nederrijn compared to the downstream side. The fraction of sand increases at the downstream side and represents the major part of the sediments which are transported in the Nederrijn (Brinke, 2004) & (Bolwidt, et al., 2006).

Figure A-18 Sedimentation behind the weirs and the bed profile of the Nederrijn

A.3.3 SALINIZATION OF THE DELTA

Salinization of the western part of the Netherlands is a problem for dry periods. Salt water from sea is able to enter the Netherlands through the Nieuwe Waterweg. The combination of the quantity of discharge of the river and the sea level determine the characteristics of the salt intrusion. The salt intrusion reaches the Willemsbrug in Rotterdam (30 kilometres inland measured from Hoek van Holland) during an average discharge and an average tide. The Haringvlietsluizen are important for counteracting the salinization of the delta. The sluices create a fresh water lake in the Haringvliet and Hollandsche Diep. Furthermore, they are closing off an estuary and forcing the water of the Rhine and Maas to flow towards the Nieuwe Waterweg for counteracting the salt intrusion. A minimum discharge of 1500 m3/s has to flow through the Nieuwe Waterweg to counteract the salinization of the Hollandsche IJssel. This minimum discharge is not

: ARCADIS & TUDelft 19

always present; salt water could even enter the Haringvliet and the Hollandsche Diep from behind for low discharges. This happened in the autumn of 2003 and 2005. (Rijkswaterstaat, 2011).

A.3.4 AUTONOMOUS BED EROSION

The river bed levels are subjected to an autonomous decrease. Especially the river bed level of the upper part of the branches decreases. The development of the bed level is shown in Figure A-19. Three different periods are shown. The first graph gives an impression of the autonomous bed decrease before the canalization of the Nederrijn-Lek, the second graph represents of the period of construction of weir Hagestein and weir Amerongen, and the third graph represents the period from the construction of weir Driel till 1990. The bed decrease from 1933-1970 is caused by the normalisation works and the dredging operations. The rivers were and still are not in equilibrium after the normalisation works which are described in A.2. Erosion takes place for reaching a new equilibrium state resulting in a decrease of the bed level. The bottom decrease located at the weirs is caused by the dredging works performed for construction and local high flow velocities caused by the weirs. The IJsselkop is recognizable in Figure A-19 as a sill in the river profile located at km 10. This is caused by the extraction of water from the Nederrijn by the IJssel. The rate of flow at the bifurcation point decreases which results in sedimentation. The rate of erosion of the Pannerdensch Kanaal and the upstream part of the Nederrijn is higher compared to the rate of erosion of the Waal. This difference results in a change of the distribution of discharge at the Pannerdensche Kop. The effect of the bed decrease with respect to the design water levels is 1 to 2 mm per year (Rijkswaterstaat, 2006).

Figure A-19 Autonomous bed erosion

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The autonomous bed decline has an impact on the functioning of weir Driel. The damming efficiency/effect of weir Driel decreases with the decline of the bed level. The weir has to be closed more often and longer compared to the originally designed solution in order to be able to influence the water level at the IJsselkop (ARCADIS, 2010) which is caused by the decline of the equilibrium depth of the river. Sand and was dredged from the shallow parts of the river and sold before 1990. The amount of sand which was extracted from the river was larger compared to the supply of sediments, which had also an effect on the bed decline. A better maintenance dredging method was introduced after 1990 to counteract the bed decline caused by dredging. The sediments which are dredged have to be relocated at deeper parts of the river. The relocated sand is again transported by the river towards the shallow parts and has to be dredged again in order to guarantee a sufficient water depth. The dredged bed material is again transported towards the deeper parts of the river and the cycle starts again (Brinke, 2004). In this way the bed decrease caused by dredging is counteracted (Miniserie van V&W, 2007).

A.4 FUTURE SITUATION

The climate change is accelerated by the impact of men. The average world temperature has increased with 0,7 0C since 1950. Local changes are influenced by the factors like air temperature, moisture balance in the bottom and air, and the air circulation patterns. The Dutch meteorological institute (KNMI) made four different scenarios for the future climate. They made a distinction between a change in temperature (an average scenario: ‘G’ and a warmer scenario: ‘W’) and a distinction between a normal air circulation pattern and a changed air circulation pattern (‘+’ scenarios). The increase or decrease in precipitation, the increase of sea level, and river runoff are determined per scenario. The reference year of these calculations is 1990, so an increase of 5% of precipitation is an increase of 5% with respect to this reference year. This ‘reference year’ was determined with use of data from 1976 until 2005. An overview of the distinct scenarios is given in Figure A-20. (Rijkswaterstaat, 2009).

Figure A-20 Overview of the distinct climate scenarios of the KNMI

The general trends based on the KNMI scenarios are:  Summers (June, July and August) will be dryer. o The water levels of the rivers will decrease. o The river discharge will decrease. o More salt intrusion due to the small river discharge will occur.  Winters (December, January, and February) will be wetter o The water levels in the polders and rivers will increase. o The river discharge will increase.  Melting of the ice will increase.

: ARCADIS & TUDelft 21

o The sea level will increase due to melted land ice. . More salt intrusion will occur due to a higher sea level.

The second delta committee performed more research into the raise of sea level. They concluded that the sea level will rise between 0.20 metre and 0.40 metre until 2050; which is in accordance with the research of the KNMI. The sea level rise until 2100 is higher with respect to the prediction of the KNMI. The delta committee stated that the sea level rises with 0.65 metres to 1.30 metres. The land subsidence of the Netherlands is taken into account in this research. The average Rhine discharge will increase up to 12% in the winter and will decrease to 23% in the summer. This change is caused by the changed precipitation patterns in the catchment area of the Rhine with respect to the reference year 1990. The combination of sea level rise and a decreased summer discharge causes more salinization in the delta. The ‘salt water wedge’ will penetrate further into the rivers and fresh water intake points along the rivers could not be used anymore during dry periods. Furthermore the amount of fresh water for counteracting the internal salinization will decrease. So, less water will be available for flushing the polders and maintaining a sufficient water level (Rijkswaterstaat, 2011). The expected fraction of time per year of a specific discharge at Lobith is shown in Figure A-21 (Technical University of Delft, 2011). The impacts of the ‘+’ scenarios are the most severe; the water levels will more often be lower when the air circulation pattern changes. The impacts of climate change of the middle part of the Netherlands are shown in Figure A-22

Figure A-21 Time fractions of certain low discharges at Lobith for the G, G+, W, and W+ scenario

Figure A-22 Impact of climate change

A.4.1 SECOND DELTA COMMITTEE

The second delta committee (Dutch: tweede delta commissie) made a list of twelve recommendations in order to keep the Netherlands safe and ready for the future. The government of the Netherlands confirmed the recommendations of the committee and decided to use this advice. The advice of the

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committee forms a basis for further elaborations and has been used for the ‘Nationaal Waterplan,’ which describes a water management vision and policy based on the Waterwet and Wet ruimtelijke ordening. The principle of the Nationaal Waterplan is: ‘move with the natural processes, perform counter-pressure when necessary and use the opportunities for increasing the prosperity’ (Rijkswaterstaat, 2011). A selection of the recommendations which are applicable for this research are listed below (Deltacommissie, 2008).

Recommendation 8; the South western delta, Krammer-Volkerak Zoommeer The Krammer-Volkerak, Zoommeer, the Grevelingen, and possibly the Oosterschelde are needed as a temporary storage of Rhine water when a high Rhine runoff and a storm surge occur at the same time. The Drechtsteden and the city of Rotterdam are preserved in this extreme situation by storing the water in these lakes and estuaries. Furthermore, the Krammer-Volkerak has to be reconnected to the Oosterschelde in order to solve the water quality problems in this area.

Recommendation 9; the river area The project Room for the River (Dutch: Planologische KernBeslissing ‘Ruimte voor de Rivier’) has to be executed in order to increase the capacity of the rivers. This project has already been described in A.3. Besides the works on the Rhine a separate plan has been made for the Maas. The maximum discharge capacity of the Rhine at Lobith in 2050 will be 16.000 m3/s and 18.000 m3/s in 2100. The maximum discharge of the Maas will be 3.800 m3/s in 2050 and 4.600 m3/s in 2100. The Rhine capacity should be enlarged to 16.000 m3/s in the first phase. Extra space should be reserved in advance for the works which should be implemented during the second phase. The first phase of Room for the River has to be finished in 2015.

Recommendation 10; the Rijnmond The Rijnmond area is a vulnerable part of the Netherlands. Many people, the harbour of Rotterdam, and a lot of industries are located along the river mouth. Nowadays the Rijnmond is protected from the sea by the Maeslantbarrier. The gates of the barrier are closed during a storm surge and are protecting the hinterland from flooding. But this is not sufficient during a storm surge and a high discharge simultaneously for now and for the future. The Maeslantbarrier is able to operate with a sea level rise of 50 centimetres, so no problems should occur before 2050. However, the barrier has to be closed more often and the closing frequency of the barrier increases further after 2050. The amount of closures per year and the increasing probability of high river runoff results in a higher probability of flooding of the Rijnmond. The second delta committee advises to implement a ‘closable but open’ solution for the Rijnmond (Dutch: ‘afsluitbaar open Rijnmond’). The Rijnmond will be closed off from the surrounding waterways by the Maeslantbarrier, Hartelbarrier, Haringvlietsluizen, a new barrier in the Spui, Oude Maas, Dordtse Kil, and the Merwede. An area which behaves like a levee ring will be created during high water. The river water will be diverted into the Krammer-Volkerak Zoommeer and Grevelingen. Furthermore, a solution for the Lek has to be found; (Temporally) closing off the Lek could be another solution when the storage capacity of the Rijnmond is not sufficient. A possible configuration of a ‘closed but open’ Rijnmond is shown in Figure A-23.

: ARCADIS & TUDelft 23

Figure A-23 ‘Closed but open’ Rijnmond

A.5 DESCRIPTION OF THE WEIRS IN THE NEDERRIJN

The weirs in the Nederrijn are described as valves which have to pass a certain discharge and have to regulate a certain upstream water level in section A.3.1. The design of the weirs is described in this appendix. First the design of the present structures is described in section A.5.1.Section A.5.2 describes the changes which are made on the original design. At last section A.5.3 describes the state of the present weirs based on the conclusions of the RINK-SSE assessment performed by ARCADIS.

A.5.1 PRESENT WEIRS

This paragraph describes the design of the present weirs. First the ‘variant analysis’ of the general structure and a ‘variant analysis’ of the gates are presented. Secondly the structure is described from global to a detailed design. The design of weir complex Hagestein is used in this paragraph as reference since every structure is nearly the same (Rijkswaterstaat, 1955). The differences of the designs of Driel and Amerongen are mentioned in the footnotes.

Figure A-24 Scale model of complex Hagestein (Rijkswaterstaat, 1955)

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A.5.1.1 VARIANT ANALYSIS

The weirs have two functions which are: ‘damming’ the river in periods of drought and providing a passage for navigation in the Nederrijn during periods of high discharge. They were not allowed to form an obstacle for navigation. For these reasons, the board of RWS stated the following:  There is not a preference between solutions with one opening with a width of 80 metres compared to a solution with two openings with a width of 45 to 50 metres.  The opening should span from bank to bank in order to generate benefits for navigation when one opening would be applied. The width of the gate would be 110 metres.  Applying one pier or two piers close to each other for the accurate discharge control is allowed with respect to ice drift during winter.

From these summation designers elaborated two different designs namely:  A weir consisting of one opening of 80 metres, an accurate discharge control opening, and a fish passage in the abutments.  A weir with two openings of 40 metres wide divided by a middle pier. The accurate discharge control would be located in the middle pier and the fish passages would be located in the abutments.

They chose for the second variant, because:  From nautical terms, there is no preference for one or two openings.  The best location for the accurate discharge control is in the middle of the river. The discharges in the middle of the river are the largest and more water could flow easily into the control system. Large unwanted velocities would be developed near the bank if the control systems would be located in the abutments.  Making two openings is much cheaper than one big opening. A gate with a width of 80 metres weighs about 1000 tons, and the total weight of two gates of 40 metres is about 650 tons. This difference will lead to a reduction of costs of f 525.000,-- (price level 1955).  The risks when using one gate are higher compared with the solution of two gates.

These arguments have resulted into option ‘C’ shown in Figure A-25. The fish passages are located in the abutments, and the accurate control system in the middle pier located in the middle of the river.

Figure A-25 Variant analysis of the weirs

: ARCADIS & TUDelft 25

The designers compared three different gates, namely: a plane lifting gate, segment gates, and a half circular segment gate. They concluded that a circular segment gate (also called a visor gate) would be the best solution. The ‘benefits’ of this gate were financially and from the hydrodynamics much larger compared to the other solutions. Some advantages of this option are summed up in the following enumeration:  The water is retained at the hollow side of the gate. This generates tensional forces in the arc which is better than moments and compression forces.  It is possible to reduce the dimensions of the gates due to the generated tensional forces. The gate is therefore 100 tons lighter than a normal gate. This would safe 700 tons of steel when seven gates (two gates per complex and one spare gate) are applied.  The hoisting machines would also be less expensive due to the reduction of weight.  The water tight plates on the gate are also shaped in an arc. Not only a horizontal radius of the gate but also a vertical radius was designed. The hydraulic load also generates tension forces in the plates on the gate (Eggink, 1954).  The gates have to be lifted over an angle of only 60° in order to realise the vertical clearance for navigation.  The top of the lifting towers can be shortened for 8 metres when a circular gate would be applied instead of applying a plane lifting gate. This safes concrete and money.  Eddies caused by the flow underneath the gates does not occur due to the perpendicular flow direction of the water underneath the gate. A redistribution of flow over the whole river width is obtained rapidly

A.5.1.2 WEIR DESIGN

Soil conditions Weir Hagestein and Amerongen are founded on a strong layer of sand. Some small peat and clay layers were present which had to be replaced by well compacted sand. However, the soil condition at weir Driel was bad; a thick layer of small grained glacial deposits were found at this location. This layer had to be excavated; the total volume of the excavation works was 500.000 m3. The thin soil layer remaining in the building pit after excavation was reinforced with a layer of 50 centimetres lead slag. This slag penetrated the remaining layer of silt and formed a firm layer for the 230.000 m3 of refill sand. The force of the piers and abutments on the soil layers are equal to force of the excavated soil, so big soil deformations of the remaining silt layer were not expected (Rijkswaterstaat, n.d.).

Weir complex description The top of the abutment on the upstream side of weir Hagestein is located at +6.50m NAP. This is 37 centimetres higher than the highest recorded water level of +6.13m NAP (1926). The top of the downstream side of the foundation is 2 metres lower with respect to the upstream side in order to remove and install a new visor gate. Two hoisting towers are located at the abutments. The engine room is accessible by the stairs on each side of the abutment. Three channels were constructed at the abutments of weir Amerongen and weir Hagestein: one channel for eel, one channel for the other fish species, and flushing drain for the hinge of the visor gate. Glass eel are able to swim against a velocity which is lower than 0.2m/s and are attracted by a flow located near the banks; so a separate channel was needed.1 The other fish species needed a counter flow of 1 to 1.5 m/s directed towards the middle of the river. This stream of water attracts the fishes from the river towards the channel. The flushing drain was necessary

1 The channel for glass eel was not necessary in Driel because the glass eel evolves into young eel between weir Amerongen and Driel. Young eel is strong enough to use the general fish passage, so a separate channel was unnecessary (Rijkswaterstaat, n.d.).

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for keeping the hinge of the gate free from sediments and preventing a local backwater effect caused by a partly lifted gate. The channels for eel, other fish species, and the flushing channel are shown at the right abutment of Figure A-24 and on the cross section of the abutment shown in Figure A-26.

Figure A-26 Cross section of the abutment

Two other hoisting towers are located at the middle pier. The engine rooms are interconnected and accessible by stairs. The top of the base structure of the pier at the upstream side is 1.50 metre higher with respect to the highest known water level. The top of the foundation of the pier at the downstream side is equal to the top of the abutments at the downstream side which makes it easy to remove and install a new visor gate. The bottom of the foundation is located at -14.60m NAP. Openings have been realised in each side of the pier for the flushing drains to keep the hinge free from sediment which are presented in Figure A-27. Furthermore, an accurate discharge control system has been installed. This system is able to fine tune the rate of flow with use of a ‘cylinder valve.’ In the first phase of the accurate discharge control, only the cylinder valve will be used. The valve will be opened in steps till the maximum discharge of 90 m3/s has been reached. The visor gates will be used when the maximum discharge of the cylinder valve has been reached and when a higher discharge is needed. The gates are lifted in steps of 13 centimetres. Each step represents an increase of about 50 m3/s per visor gate. First the gates are lifted for 13 centimetres and the cylinder valve is closed. A discharge of about 100 m3/s flows underneath the visor gates. The cylinder valve is used again. It is opened in steps when a discharge higher than 100 m3/s has to be obtained. The cylinder valve closes and the visor gates are opened for another 13 centimetres when the maximum discharge of the cylinder valve has been reached again. This process goes on until the desired discharge has been set. Furthermore a hydro power station is located in the middle pier of Hagestein. 2. This hydropower station has a capacity of 65 m3/s, which makes a combined discharge of 150 m3/s possible for the accurate discharge system when the cylinder valve is included (Boersma, 1987).

2 Hydro power station in the weir of Driel and Amerongen were not profitable with the available techniques at that time (Hof, 1961). The dimensions of the middle pier of Hagestein differ from Amerongen and Driel as a consequence of this hydropower installation. The middle piers of Amerongen and Driel are 1.50 smaller, and 7 meters shorter with respect to weir Hagestein

: ARCADIS & TUDelft 27

Figure A-27 Horizontal cross section of the middle pier

A lift shaft is also included in the design of the weir. The shaft connects the upper part of the structure with a tunnel constructed in the river bed. This tunnel connects the middle pier with the abutments. People can use this tunnel to walk from the banks towards the middle pier when the visor gates are open. An overview of the cross section of the middle pier has been given in Figure A-27 and Figure A-28. The lift shaft is illustrated on the right hand side of Figure A-28. The inflow and outflow of the cylinder valve and the hydropower station are also visible at Figure A-24.

Figure A-28 Horizontal and vertical cross section of the middle pier

The visor gates are located in between the abutments and the middle pier. The ‘horizontal’ span of these gates is 48 metres and the total weight equals 210 tons. The visor gate is divided into two segments by a shear connection. The gate has been connected to the pier and abutment by hinges which makes the gate a statically determined structural system. Therefore, uneven stresses caused by differences in temperature are limited. Furthermore the shear connection makes it is possible to fold up the gate for maintenance and

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renovation. The main girders of the visor gate are placed 6.4 metres from each other with vertical connections located in between. The height of the gate is 9 metres and the water tight plates are 8 mm thick. The main forces in the gates are tension forces due to the cylindrical shape of the gate and the plates due to the horizontal and vertical curvature of the plates. The horizontal radius of the gate and the vertical radius of the plates are equal. In this way the tension forces created by the hydrostatic pressure are equally divided in the horizontal and vertical direction of the plates. The hinges are located 25 centimetres above the ‘normal’ water level at the downstream side of the weir. This makes it easy to perform maintenance during operation. Rubber profiles are used to keep the hinge free from the upstream water. These profiles close off the gap in between the gate and the abutment using the upstream water pressure. The visor gate is hoisted up by cables which are connected in line of the centre of gravity of the visor gate. The cables are guided by rollers on the arches towards the engine room on top of the arc/hoisting tower. Two engines of 10 HP (equal to .4 kW) are used to lift the visor gate. The hoisting velocity is 2.5 cm/s and the hoisting process takes 2.5 hours. The water levels will gradually change over time and a sudden water level drop in the river is not present due to this low velocity (Peere, 1961), (Eggink, 1954). An overview of the gates, hoisting equipment, and engines is given in Figure A-29.

Figure A-29 Visor gate, hoisting equipment, and engine

Seepage underneath the structures of weir Amerongen, Hagestein and Driel is counteracted with use of an already present horizontal layer of clay. This layer of clay is a few metres thick and is able to withstand the water pressure caused by the weir if it is reinforced with a layer of concrete blocks. Using this layer is a cheaper solution compared with a vertical water sealing screen due to a layer of coarse sand which is present over a great depth. Some perforations were made and were filled up with course sand to prevent bursting up of the soil due to high ground water pressures caused by a sudden raise of the gates. The total seepage length is 300 metres and the maximum water head over the weir is 5.10 metres. The sill of the weir consists of concrete blocks which are reinforced with a 3 centimetre thick layer of special reinforced concrete (Dutch: ‘staalpantserbeton’). They have been aligned in line with the radius of the gate. These blocks are also used for the stilling basins. An impression of the dimensions of the sill has been given in Figure A-30.

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Figure A-30 Installation of the sill (Hof, 1961)

A.5.2 CHANGES IN THE DESIGN

Some changes are made in the design from the completion until now. This chapter describes the major adaptation of the design. The adaptations are described chronologically.

A.5.2.1 WATERPOWER INSTALLATION AT WEIR AMERONGEN (1984)

A.2 stated that a profitable hydropower station was only possible at Hagestein due to the larger head compared to the other locations. But the fuel prices increased over the years which made the generation of waterpower energy at weir Amerongen profitable. The first ideas for new waterpower installations in the Nederrijn started at the beginning of 1983. The feasibility studies were completed in February 1984. These studies stated that a water power installation was feasible near the village of Maurik next to weir Amerongen. A net output of 32GWh was possible with use of an optimisation of the dam regime. Without this optimisation the yield would be 29 GWh. (A waterpower installation next to weir Driel was not profitable due to the lower water head and the limited days for which the is closed.) Four Kaplan turbines with a horizontal axis were applied at the waterpower station of weir Amerongen. The maximum efficiency of this water power installation is 94.15% and generated with a net head of 3.18 metres and an optimal discharge of 50 m3/s. Construction started in December 1985 and finished in September 1988 (Boersma, 1987). The structure is shown in Figure A-31. This waterpower installation is still in operation.

Figure A-31 Water power station of Maurik (Klunne, 2012)

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A.5.2.2 IMPROVING FISH MIGRATION BY BUILDING FISH LADDERS (1990)

The Nederrijn is part of the fish migration route from the Noordzee towards Germany. Fish passages were realised in the abutments to enable fish to pass the weirs. A description of these channels has already been given in A.2. These fish passages are out of service since 1990 and were disassembled between 1998 and 2002 (ARCADIS, 2010). New fish passages have been realised next to the weirs to enable the fish to cross the weir. The passage of Amerongen consists of 24 basins with a difference in height of 16 centimetres per basin. Small V shaped weirs divides the basins (Waterloopkundig Laboratorium, 1995). These fish passages improved the fish migration.

A.5.2.3 UNDESIRED VIBRATIONS OF THE VISOR GATES OF WEIR HAGESTEIN (1993)

Tests were performed to measure the stresses and vibrations of the gates during operation after the construction of the visor gates. The result of these test indicated that the stresses of the visor gates were ‘far below the calculated stresses’ and the vibrations of the visor gates were within the set limits (Rijkswaterstaat, n.d.). However the vibrations of the gates increased and the gates started to vibrate from every position after a couple of years. The weir operator had to hurry to get in time at the weirs to take the visor gates out of their vibrations. An unwanted load condition was introduces by the vibrations which is fatigue. RWS started an investigation to determine the cause of the vibrations in 1993. The results showed that the visor gates vibrated in a higher vibration mode. The circle shaped seal of the visor gate caused the vibrations. Dynamic vortex shedding due to an undefined release point caused the gate to vibrate. For solving this problem a seal had to be installed with a defined release point which limits the vortices (Daniel, 2005). The difference of the old and new rubber seals are shown in Figure A-32.

Figure A-32 The old seal a) and the new seal b)

The seals could only be changed during high water because the visor gates are lifted for this situation. Water levels needed to be high enough for a week in order to finish the works, which is an exceptional situation. The seal of the southern visor gate was changed during February 2003; the first melting water came down the river which resulted in water levels of +11.40 m NAP at Lobith. The seal of the northern gate was replaced during the second part of February 2005. The unwanted vibrations stopped after replacement of the seals (Daniel, et al., 2005).

A.5.2.4 WATERPOWER INSTALLATION AT WEIR HAGESTEIN (2005)

The waterpower station was closed in 2005 for maintenance but NUON made a sustainability report (2005) which indicated that the production was lower than expected and the interest in ‘green electricity’ was

: ARCADIS & TUDelft 31

decreasing. As a result the waterpower installation was shut down until now. Some firms and the municipalities of Houten and Vianen showed interest for reopening the waterpower plant, but RWS decided to delay the reopening of the plant in 2010 (Nederlandse vereniging voor Energie uit Water, 2010). In the meantime RWS performed research about the condition of the waterpower plant and have to decide whether they were going to tender the plant or exploiting the plant by themselves. This decision will be taken after further investigation in the course of 2013 (Atsma, 2011).

A.5.3 STATE OF THE WEIRS

Rijkswaterstaat is the manager of about 500 hydraulic structures and the responsible organisation for the functioning of these structures. A lot of the structures have (nearly) exceeded their technical life time. The reliability and availability of the hydraulic structures decreases over time, but an overview of the actual reliability and availability of these structures is no available. RWS awarded ARCADIS the project RINK_SSC (Risico-Inventarisatie Natte Kunstwerken –Sluis-StuwComplexen) to determine the reliability and availability of the complexes of Driel, Amerongen, and Hagestein. The following main functions and failure definitions were used during the reliability and availability assessment (ARCADIS, 2010):  Water management function; the following events results in failure: o Water level too low: the water level is 15 centimetres lower compared to the ‘normal water levels.’ o Water level too high: the water level is 50 centimetres higher compared to the ‘normal water levels.’  Navigational function; the following events will result in failure: o Passage of the locks is not possible for an hour and passage of the weirs is not possible. o Water level too low: the water level is 15 centimetres lower compared to the ‘normal water levels.’

The following RA-requirements were defined from a consult with Rijkswaterstaat:  Weir: no requirements were defined. Measures are defined in the Probabilistic maintenance plan (Dutch: Probabilistisch InstandHoudingsPlan; PIHP) for the parts which have the largest contribution of the unforeseen non availability.  Lock: The unforeseen unavailability needs to be lower than 12 hours per year.

The ‘normal water levels’ shown in Figure A-33, are determined using the river modelling program SOBEK. The water levels at the IJsselkop, water levels at the upstream side of the weir Driel (‘Driel boven’), water levels at the downstream side of weir Driel (‘Driel Beneden’), ‘Amerongen boven,’ ‘Amerongen beneden,’ ‘Hagestein boven,’ and ‘Hagestein beneden’ are modelled and presented in Figure A-33. The water levels at these seven locations are a function of the water level at Lobith. The water level of ‘Driel boven’ follows the water level at the IJsselkop til the water level of +8.90 m NAP has been reached. The visor gates and the cylinder valves are gradually opened and the water level at ‘Driel beneden,’ increases. Also the visor gates of Amerongen and Hagestein are partly opened; the water levels of ‘Amerongen beneden’ and ‘Hagestein beneden’ are increasing and the water levels of ‘Amerongen boven’ and ‘Hagestein boven’ are decreasing. The visor gates of Driel are fully opened when the water level at Lobith is +10.0 m NAP; the water level of ‘Driel boven’ is equal to the water level of ‘Driel beneden.’ The visor gates of Amerongen and Hagestein are fully opened when a water level of +11.50m NAP at Lobith has been reached. The water levels of ‘Amerongen boven’ and ‘Hagestein boven’ are equal to the water levels of ‘Amerongen beneden’ and ‘Hagestein beneden’ when the water level at Lobith is +11.75m NAP. The Nederrijn is from here on a free flowing river.

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Figure A-33 Normal water levels (ARCADIS, 2010)

ARCADIS analysed the structures in order to determine the availability and reliability. The outcomes of this analysis are described in this paragraph. First the main functions of the weirs are presented. Secondly the reliability and availability of the structures are described and the methodology of the determination of the probabilities and the outcomes of the analysis are presented. Finally the conclusions made by ARCADIS are given. The data which has been used, originates from the following documents: the ‘Integrale rapportage - complex Driel’ (ARCADIS, 2010), ‘Integrale rapportage – complex Amerongen’ (ARCADIS, 2010), and ‘Integrale rapportage - complex Hagestein’ (ARCADIS, 2010).

A.5.3.1 FUNCTIONS OF THE WEIRS

The original purposes of the weirs did not really change since the completion of the canalization of the Nederrijn. In A.2 several dam-regimes which were designed for the canalization project are described. After 40 years of experience the dam-regimes were fine tuned into the present regime S285. This dam- regime covers a minimum discharge of 25 m3/s in the Nederrijn and 285 m3/s in the IJssel. The cylinder valve and the visor gates are gradually opened for supplying a 285 m3/s discharge in the IJssel when a discharge higher than 310 m3/s is present in the Pannerdensch Kanaal. The main functions of the weir which are determined in the context of RINK-SSC are presented in Table A-2. These are the same functions as described in the previous chapter but extended with some specific values of discharge and water levels. The basic functions of Amerongen and Hagestein are maintaining the present situation of the designed normal water levels generated by the weirs. The weirs of Driel, Amerongen, and Hagestein have also secondary functions. These secondary functions are also presented in Table A-3. The described secondary functions are part of the basic functions, navigational functions, and usage functions which are defined in the ‘Beheer- en Ontwikkelingsplan Rijkswateren 2010-2015.’

: ARCADIS & TUDelft 33

Table A-2 Main functions of the complexes

Basic functions of ‘Safety’ and Complex Navigational function ‘Sufficient water’ Driel Maintaining the discharge of Maintaining the navigational depth by damming the IJssel at a minimum of 285 the waterway IJsselkop-Driel to provide a water m3/s by maintaining the water level of +8,30m NAP and maintaining the level at the IJsselkop at +8,30m navigational depth of the Boven-IJssel by NAP or higher. maintaining a discharge of 285 m3/s. Amerongen Maintaining the upstream Maintaining the navigational depth of the water system and ground Nederrijn by damming the waterway Driel- water level along the Nederrijn Amerongen to provide a water level of +6,0m NAP by maintaining a water level of or higher. +6,0m NAP or higher. Hagestein Maintaining the upstream Maintaining the navigational depth of the water system and ground Nederrijn by damming the waterway Amerongen- water level along the Nederrijn Hagestein to provide a water level of +3,0m NAP by maintaining the water level or higher. of +3,0m NAP or higher.

Table A-3 Secondary functions of the complexes

Effect Secondary functions Realisation of the function Insufficient refreshment Maintaining a controlled Cylinder valve in the middle pier of of the dammed Nederrijn. minimum discharge of 25 m3/s every weir. of the Nederrijn. Chatter of the lock doors Realising a minimal head by Using the dam regime. The prins of the prins Bernhardsluis controlling the water level of Bernhardsluis will be opened when the due to insufficient water the waterway Amerongen- water level at the Waal is lower than head difference between Hagestein. +3,15m NAP. An open connection the Amsterdam- between the Amsterdam-Rijnkanaal Rijnkanaal and the Waal. and Nederrijn will be present in this situation. Too high water levels at Lifting the visor gates over an Part of the dam-regime. All weirs are the Nederrijn and IJssel angle of 60°. opened when the water level at the during high Rhine Bovenrrijn has exceeded +11,4m NAP. discharges. Driel will be opened when a water level of +10,0m NAP has been reached. Blocking transport of ice. Lifting the gates during ice Part of the dam-regime. drift. Blocking navigation by to Enable navigation. Providing locks next to the weirs. the closed weirs. Concentrated water level Generate 'green' energy. Realising a turbine at Hagestein and difference caused by Amerongen. closed weirs. Idem Damage will be caused due to Providing a manual (which is yet not undesired operation of the available). visor gates and high water level differences.

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A.5.3.2 FAILURE SCENARIOS

Possible failures of the design and failures of the systems are indicated with use of a Failure Mode Effect and Criticality Analyse (FMECA). The qualitative analysis gives insight in the scenario’s which could lead to failure. The following aspects were determined per component of the system:  What could go wrong?  What would the cause be?  What are the consequences of the mistake for the component and for the functions?  How often will the cause happen and how severe will it be? The criticality will be determined on basis of this aspect.

The results of the FMECA are:  the system which ensures a synchronous operation of the lifting device (all weirs)  hoisting cables (all weirs  driving unit of the cylinder valve (all weirs)  shear connections (piano hinge) of the visor gate (Amerongen)  the bearings of the visor gate (Amerongen)

Furthermore external events have been indicated. The analysis indicated the following external events:  extreme wind  ice drift  wind  lighting  air crash  fire  internal floods  (navigational) transport accidents in the surroundings  vandalism.

New risks were identified during the reliability and availability assessment. These risks are divided into the subjects: procedures and tests, maintainability and maintenance, changes in loads, organisation, law, and changes in policy. The following load aspects have been changed over time:  The amount and size of vessels navitaging on the Nederrijn  Weir Driel has to be used more often and longer due to the autonomous bed decrease. A decreased bed level causes a decreased controllability of the water levels at the IJsselkop  The norms of structures have been changed, which results in the following consequences: o Changed safety factors of the loads and materials which results in other unfavourable load conditions o Different safety factors for water level difference have been used in the original design. o The assessment of fatigue has been changed. The other subjects will not be described in this literature study.

Making a fault tree is the next step of the reliability and availability analysis. Risks which have been indicated in FMECA and the external risks are included in the fault tree. The top event of the fault tree is: “water management function fails”. This top event has been split up into three categories, namely: the deviation of the water level is too large, structural failure of the weir (excluding failure of the bed and bank protection), and failure caused by an external event. The failure functional failure resulting form a too large water level deviation is presented in the sequential enumeration.  Water management function; the following events leads to failure:

: ARCADIS & TUDelft 35

o Water level too low: the water level is 15 centimetres lower compared to the ‘normal situation.’ o Water level too high: the water level is 50 centimetres higher compared to the ‘normal situation.’  Navigational function; the following events will cause failure: o Passage of the locks is not possible for an hour and passage of the weirs is not possible. o Water level too low: the water level is 15 centimetres lower compared to the ‘normal situation.’

Failure due to a too large deviation of water levels would be caused for three operational situations. The operational situations are described in the following enumeration:  Situation one: the discharge is only controlled by the cylinder valve (<90m3/s).  Situation two: the discharge is controlled with the visor gates within the control range (90-435 m3/s). The following events could result in failure: o Non-functioning of one visor gate. o Non-functioning of two visor gates.  Situation three: free flowing situation of the river.

The three different situations are shown in Figure A-34 as a function of time and discharge. The different situations were modelled with SOBEK to determine the deviation of the water levels resulting from each failure situation for each complex separately.

Figure A-34 Different operating situations: 1) only the cylinder valve, 2) cylinder valve and the visor gates, and 3) a free flowing river

The following events contribute to the event “a too big deviation in water level:”  The cylinder valve does not close (during low discharges).  One visor gate does not function (within the control range).  Both visor gates do not function (within the control range).  One visor gate does not function (during a free flowing river situation).

A quasi stationary calculation was performed for the given events. The discharge at the Boven Rijn (Lobith) was increased in steps of 300 m3/s until a discharge of 3500 m3/s was reached. The weirs are not active anymore at this discharge. From heron the modelling steps were increased to 1000 m3/s. The calculations were executed until an equilibrium of the river was reached. The calculated water levels were compared with the normal water levels shown in Figure A-33. The water level deviations for every

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circumstance were determined using this method. The inertia of the river was also taken into account, because failure of a visor gate will not lead to instant water level changes; it will take time before the water levels become too high. The weir or cylinder valve could be repaired in between the ‘event moment’ and ‘failure moment.’ The duration between the ‘event moment’ and the ‘failure moment’ has been determined with historical data. The maximum and average increase of water level at the Nederrijn as a function of the water level at Lobith was determined. An estimate could be made of the timespan between the failure event and failure moment by combining the discharge relation between the Nederrijn and Lobith with the determined water level increase at Lobith. With this data an estimate is made of the maximum threshold of reparation time after failure. The quantification of the failure tree of the weir gives an indication of the frequency failure. Three different aspects were taken into account during the modelling and quantification, namely:  Time fractions of the three different situations: The failure depends on the operating situation of the weir, so time fractions of each situation had to be known in order to determine the probability of failure. The time fractions were determined using hydraulic modelling. The results are shown in Table A-4.

Table A-4 Time fractions of the weirs

Weir Weir Weir Situation Hagestein Amerongen Driel 1 28% 28% 28% 2 60% 60% 43% 3 12% 12% 29%  Thresholds of the recovery time: The system fails not directly after a failure of an object. It takes time before unwanted water level deviations occurs. These thresholds are also determined for every different situation. No failure occurs when the reparation of an failed object is successful exceeded within the maximum recovery time. The duration of a reparation has been determined by ‘probability of failure specialists’ and ‘discipline specialists.’ The probability density function of the durations of the repairs was combined with the probability density function of the thresholds of recovery time which were determined by hydraulic modelling. This combined probability density function was used in the assessment for the determination of the availability and reliability.  Non noticeable failure of the visor gates: the visor gates could fail during stand-by phase. Failure will not be noticed until the next use of the visor gates.

The uncorrected failure frequencies were determined by ‘probability of failure specialists’ using generic data. These failure frequencies were adjusted following the results of inspections and structural calculations. Unity checks were executed for the constructive parts. The object passes the check and is reliable when the outcome of the unity check is ⁄ . Further investigation is necessary if the outcome of the check is ⁄ ; the probability of failure of the object has to be determined with a probability of failure calculation. An object has to be renewed when the outcome of the check was ⁄ . In this case the object has exceeded its technical life time and needs to be replaced. Otherwise, these objects would dominate the outcome of the assessment which results in a unusable outcome of the assessment. Calculations are made of the renewed situation with the probability of failure of the renovated parts. No adjustments are made for the electrical components and the control systems because a distinction can only be made between well-functioning components and components which have exceeded their technical life time. Furthermore no adjustments have been made for components which could fail constructive and on a non-active manner. For the last manner of failing a high failure frequency was already applied.

: ARCADIS & TUDelft 37

The results of the assessment for every weir are shown in Table A-5. A probabilistic maintenance plan (PIHP) could be introduced to improve the availability and reliability over time. This plan presents a set of actions and the expected unforeseen non-availability for the next 20 years.

Table A-5 unforeseen non availabilities for different situations

Description Weir Hagestein Weir Amerongen Weir Driel unforeseen non [-] [hour/year] [-] [hour/year] [-] [hour/year] availability Present condition "Poor" "Poor" "Poor" After replacement / revision of the objects 0,18% 15 0,53% 46 0,32% 28 which exceeded their technical life time After implementing the 0,03% 10 0,05% 8 0,03% 12 PIHP actions The actions which can be undertaken from the PIHP reduce the non-availability. The relation between the cumulative cost and the reduction in the non-availability is shown in Figure A-35. The first costs are the most effective. A big reduction of non-availability could be achieved with a smaller amount of money compared to the same reduction in non-availability in the ‘tail’ of the graph. RWS has to find a balance between the costs and the benefits in order to make a proper decision about the amount of money they will spend.

Figure A-35 Effects of the PIHP actions

It is clear that a renovation of the weirs is necessary to guarantee a sufficient performing of the basic functions. The elements showed in the following enumeration did not pass the unity check

⁄ and needs to be replaced:  electric engines, brake clutches, brake thrusts, reduction gears of the visor gates, cylinder valve, and flushing drains (all weirs)  cables and rotation wheel of the visor gate (all weirs)  chains, brakes, transmissions, and guidance of the propulsion of the visor gate (all weirs)  brakes, transmissions and guidance of the propulsion of the flushing drains (all weirs)  electrical installation and operating system (all weirs)  low voltage installation (all weirs)  guidance and brakes of the cylinder valve (Amerongen)  steel structure of the cylinder valve (Amerongen)

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A.6 WEIR DESIGNS

A.6.1 THE INFUENCE OF A WEIR ON THE WATER LEVELS

Controlled rivers are often regulated with dams and weirs in order to increase the water level upstream and/or regulate the discharge. The goals of the regulation of a river are:  Controlling the water levels at inlet works (irrigation etc.).  Creating a water level difference for hydropower stations.  Counteracting bed erosion.  Increasing the water depth for navigation.  Controlling the discharge distribution over different branches (distribution of water of the IJssel and Nederrijn).

Two different kinds of structures could be built for these goals, namely: a fixed weir and a movable weir. Both structures have boundary conditions determined by the goal and the feasibility of the project. The two main boundary conditions are:  Fixed water levels; the water levels needs to be constant at certain locations along the river.  Water levels which are too many times too high or too low with respect to the required water levels.

The minimum and maximum discharge is determined with the probability distribution for the ‘normal’ discharges. The water level needs to be high enough at certain locations or river sections (for navigation, intake points etc.) for the minimum discharge. On the other hand, water levels need to be limited at certain locations or river sections during the maximum allowable discharge. A fixed weir needs to be designed in such a way that the discharge and back water effect are in accordance with the minimum and maximum requirements. The water level effects are shown in Figure A-37. The back water effect is recognizable in the water levels of this figure. Water levels will go towards their equilibrium depths which belong to the free flowing river. The equilibrium depths for a sub critical state are a function of the specific discharge, the Chézy roughness, and the slope angle. This depth can be calculated with use of Equation A-1. ⁄

( )

Equation A-1 Equilibrium or normal depth according to Chézy

In which:

he = normal depth [m] q = specific discharge [m2/s] C = coefficient of Chézy [m1/2/s]

ib = slope of the bed surface [-]

A critical depth is present in the river iwhenthe flow is critical (Froude is 1). This depth can be calculated with Equation A-2. ⁄

( )

Equation A-2 Equilibrium of a critical depth

In which:

hc = critical depth [m] q = specific discharge [m2/s]

: ARCADIS & TUDelft 39

g = gravitational constant [m/s2]

The back water effect can be approximated with the formula of Bresse for small Froude numbers.

⁄ ( )

Equation A-3 The approximation of Bresse

In which:

he = normal equilibrium depth [m]

h0 = boundary condition depth [m] x = x coordinate of the river [m]

x0 = x coordinate of the boundary condition [m]

L1/2 = half-length given by Equation A-4. ⁄

⁄ ( )

Equation A-4 The 'half-length'

The relative bottom slope determines the difference between a subcritical flow and a supercritical slope. The square of the Froude number (Equation A-5) of a steady flow represents the distinction between the both kinds of flows.

Equation A-5 The calculation of Fre2

In which: Fr = the Froude number [-]

ib = the bottom slope [-]

cf = dimensionless coefficient of resistance [-]

A steep slope (S-type) is present if the square of the Froude number is larger than 1. A mild slope (M-type) is present if the square of the Froude number is smaller than 1. Steep slopes are located in streams located

in the mountains. The bed slopes of the Dutch rivers are generally mild (ib=10-4, cf=4 x 10-3, and Fr2=0.025). Three different energy lines (Dutch: verhanglijnen) could be present in the Dutch river systems for a non- steady situation. Namely a M1 curve, an M2 curve, and a M3 curve:  Type 1: the depth is larger than the equilibrium depth and the critically depth. The depth is increasing in the downstream direction. A horizontal water level is present in at the downstream side.  Type 2: the depth is in between the equilibrium depth and the critical depth. The level is decreasing from the equilibrium depth towards the critical depth.  Type 3: the depth is smaller than the equilibrium depth and the critical depth. Furthermore the flow is supercritical (Fr>1). The water depth is increasing from the upstream to the downstream side.

The energy lines for a mild slope are given in Figure A-36. The equilibrium depth he and the critical depth

hc are also shown (Battjes & Labeur, 2009).

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Appendices New canalisation of the Nederrijn and Lek

Figure A-36 M-type energy lines, equilibrium depth (de) and critical depth (dg)

The new weir has a major impact on the morphology of the river. The influences of the river can be split up in three different sections (Vriend, et al., 2011):  upstream: outside the influence of the weir The original discharge Q is present in this river section. The flow causes a sediment transport S. The total undisturbed sediment transport in shown in Equation A-6.

∫

Equation A-6 Undisturbed sediment transport

 upstream: within the influence of a weir The flow velocity of the river decreases resulting from the weir. This causes a decrease in the

annual sediment transport V0 and causes sedimentation in this reach.  downstream of the weir. o Water with little sediment passes the weir when the reservoir catches nearly all the sediments. The water which flows at the downstream side of the weir is able to transport sediments and causes erosion. o Sedimentation occurs due to the discharge regulation if sediment is able to pass the weir.

Figure A-37 Water level regulation of a fixed weir

A series of weirs are needed if one weir is not sufficient for regulating the minimum discharge. The amount of weirs could be calculated with a water level slope calculation (Dutch: verhanglijn berekening). The water level at point B in Figure A-38 can be calculated with Equation A-1 and Equation A-3 if the river

: ARCADIS & TUDelft 41

is sub-critical. The water levels are horizontal and the water level at point A is equal to point B if the discharge is equal to zero. The minimum depth at point P can be guaranteed for a certain amount of days per year by determining a optimal spacing between the weirs.

Figure A-38 Configuration of weirs on a river

The dimensions of a movable weir are determined by the minimum design discharge and the minimum required water levels. The back water effect which is caused by weir is limited when the gate is fully removed during high water; the river is a free flowing river in this situation and the situation is comparable with a fully dammed river when the gates are fully closed. The ‘path’ from a dammed river until a free flowing river is called a dam regime. A general dam regime is shown in Figure A-39. The water head at the weir is decreasing due to a partly opened gate. At a certain point (at 230 days in Figure A-39) the gates are fully opened and a free flowing river is present. The water head at the weirs could be used for generating electricity with use of a hydropower plant.

Figure A-39 An indication of a dam regime

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Sedimentation and erosion also occurs with a solution consisting of movable weirs. Sedimentation occurs at the upstream side of the weir when the gates are closed. But erosion also occurs on the upstream and the downstream side of the weir for a (partly) open situation:  Erosion at the downstream side occurs due to ‘clear’ water with an available transport capacity which passes the weir during a partly closed situation. The water which passes the weir does not transport sediments so it has a high available transport capacity. The clear water picks up sediments ant the downstream side which causes erosion.  Scour holes in front and behind the weir are caused by the strong turbulence generated by the sub-structure in an open situation.

The bottom profile shaped by the sedimentation process upstream and downstream of the weir is presented in Figure A-40. Two distinct situations are sketched in this figure; one for a closed situation and one during for an open situation. The trends of autonomous bottom erosion in the Nederrijn shown in Figure A-19 are in accordance with the described sedimentation and erosion pattern caused by a movable weir. Sedimentation between Hagestein and Amerongen occurred, which is visible in the third cross section of the Nederrijn. Erosion took place before and after the weirs which is visible in the same cross section of this figure. The weirs influence the water levels at the upstream side. A fixed weir or a (fully) closed weir causes a rise in the water levels. This water level increase causes ‘weir damage’ like an increase of groundwater levels, blockage of present drainage, levees which are too low for the new situation, and pumping station which do not have sufficient capacity. These side effects have to be taken into account during designing of a canalization of a river. The ‘weir damage’ could be reduced by limiting the water head at the weir and the distance between two successive weirs, but the consequence is an increase of the amount of weirs which have to be built. So, a balance has to be found between:  the number of structures  Hindrance costs of navigation during construction and in the future situation.  ‘Dam damage’ and works which have to be implemented to limit the damage.  exploitation costs and maintenance costs

The solution with the lowest costs and the highest benefits will probably be constructed (d'Angremond, 1998).

Figure A-40 Bottom profile before and after a weir

Two different kinds of weirs gates are available namely an overflow and an underflow gate. An overflow weir has a better control of the water levels upstream, the downstream water levels does not influence

: ARCADIS & TUDelft 43

upstream water levels, good discharge of debris, but a bad discharge of sediments. The properties of an underflow weir are: well controlled discharge downstream, larger water level variation upstream, good discharge of sediments, but a bad discharge of debris (Gijt & Toorn, 2011). The choice of a over or underflow gate is site specific.

A.6.2 AN OVERVIEW OF WEIRS

Several weirs and barriers have been built in the Netherlands. These structures are described in this paragraph and could serve as a reference work for the design of a new weir for the Nederrijn. Furthermore, some innovative designs of barriers and weirs are described in this paragraph. The border between barrier and weir is not really clear. As example; a structure like the Ramspol barrier could also serve as a weir in a river. First the weirs in the Maas are described. Seven weirs are constructed in the Maas. The weirs of Sambeek, Belfeld, Roermond, and Linne do have nearly the same design. The only difference between the weirs is the amount of openings of the Poirée part and the Stoney part Therefore, only the weir of Sambeek is considered. The weir of Lith and Grave located in the province of Noord-Brabant are built according other designs and are described in this chapter.

Weir Sambeek (Maas River; Province of Noord-Brabant) The weir of Sambeek was constructed in 1929 and is part of the canalization of the Maas. The weir consists of two parts, namely a Stoney part and a Poirée part which is shown in Figure A-41. The Poirée part consists of 13 openings. Three separate gates can be placed on top of each other to close of the Poirée part of the weir. The yokes are placed on the sill of the weir. A special designed gantry crane is able to remove or place the gates and is able to remove the rail way track/bridge. The yokes are lowered when all the gates and the bridge has been removed. The Poirée part is used for the crude regulation of the discharge during high river runoff. The weir is fully closed if the discharge is lower than 200 m3/s. In this case the discharge is regulated by the Stoney part. Gates of the Poirée part are removed in order to be able to regulate the water levels with the Stoney part when the discharge is higher than 200 m3/s and smaller than 1070 m3/s. All the gates are removed and the yokes are lowered when the discharge is higher than 1070 m3/s (Brabants Historisch Informatie Centrum, 2012) The Stoney part consists of two openings of 17 metre wide. The gate has several wheels either side and is held in position by two vertical plates. The wheels are positioned at the downstream side of the pier, so they are being pushed by the water head to the concrete. The gate is lifted by steel cables connected to the gate. Accurate regulation of discharge is possible by gradually lifting the gates The Stoney gate which has largest span of 40 metres is constructed near the village of Landenburg in Germany. The Stoney gate which has the greatest height (17 metres) is constructed in near the village of Cize in France. These kind of gates are not often built anymore due to the high costs and the greater need of maintenance compared to other gates (Erbisti, 2004).

Figure A-41 The weir of Sambeek (Google maps)

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Weir Lith (Maas River; province of Noord-Brabant) The weir of Lith, shown in Figure A-42, was constructed in 1934. The construction of this weir was part of the Maas canalization and normalisation works described in A.2.1. The discharge of the Maas is (highly) variable. The discharge could be less than 5 m3/s in summer, and could rise above 3.000 m3/s during winter. The purpose of the weirs in the Maas is to guarantee a sufficient water level during summer for wicht a discharge of 5m3/s is present. The weir of Lith was completed in 1936. The weir consists of three distinct vertical cilincrical roller gates placed between hoisting towers. A small ‘flap gate’ has been constructed on top of each gate. The flap gates could gradually being raised by chains and winches in order to regulate the discharge over the weir. The maximum discharge per flap gate is 333 m3/s and the maximum discharge which can be regulated by the flap gates is 999 m3/s. The vertical roller gates are fully lifted by the hoisting equipment located in the hoisting towers when the discharge exceeds 1000 m3/s. The bottom of the vertical roller gate in lifted position is +13.60m NAP. The target water level of weir Lith is +4.90m NAP, so significant clearance is available for navigation. A lock is available for navigation for low discharges. (Brabants Historisch Informatie Centrum, 2012). The gates are lifted by the hoisting equipment located in the pylons. The steel cylinders are applied with a toothed gear and are fixed on a toothed track on each side. These gates are generally used for weirs with a low water level difference and a wide opening. A cross section of a gate and the toothed gear is presented in Figure A-42. Wide openings between piers are possible due to the high stiffness of the cylindrical shape of the gate. Vertical roller gates till a span of 45 metres have been built (Erbisti, 2004).

Figure A-42 Vertical roller gates; Weir Lith (left) Cross section of a roller gate (right)

Weir Grave (Maas River; province of Noord-Brabant) The weir of Grave is combined with a bridge over the Maas, which is shown in Figure A-43. The construction was completed in 1929 and was part of the improvement works of the Maas. The weir structure has two openings which are located underneath the bridge. Twenty yokes are applied per opening. Every yoke has three small removable gates which are placed on each other as shown in the right figure of Figure A-43. The gates can be removed or added in order to regulate the discharge of the river; 20 m3/s can be regulated per gate. The water level at the upstream side of the weir is regulated between +7.30 m and +7.60 m NAP by placing or removing a certain amount of gates. The distinct gates which are removed are placed on the yoke above the water line. This system is called an inverted Poirée system. All the separate gates are removed when the discharge exceeds 1000 m3/s. The yokes rotate in the direction of the flow when they are removed from the waterway. They are secured underneath the bridge by a specially designed gantry crane. A sudden removal is possible if a sudden high discharge or ice drift is present at the river due to the direction of the movement of the gate. Damming the river is more difficult; it is hard to place the yokes against the sill due to the flow of the river which has to be overcome The yokes

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are pulled towards the sill with a system of pulleys and chains, operated from the gantry crane. Navigation is able to pass the weir by the locks (Brabants Historisch Informatie Centrum, 2012)

Figure A-43 Inverted Poirée weir; Weir Grave (Google maps) (left) and a cross section of this system (right)

Weir Borgharen (Maas River; province of Limburg) The weir of Borgharen is the most upstream weir located in the Maas in the Netherlands. It has to retain water and divert the water into the Julianakanaal in periods of drought. The weir is subdivided into four openings; one opening of 30 metres wide for shipping and three openings of 23 metres for water discharge. 30 metres of width is sufficient for the ships which sails on the non-canalised Maas. The hoisting towers of the ‘navigation gate’ are higher with respect to the hoisting towers of the ‘discharge gates’ due to the needed vertical clearance for the vessels which is shown in Figure A-44. Vertical roller gates are applied in the openings with an extra flap gate on top of the roller gate of the discharge openings in order to be able to ‘fine tune’ the discharge. A discharge till 230 m3/s can be regulated by the flap gates. The vertical roller gates have to be lifted if the discharge rises above 230 m3/s (Het Centrum, 1928). The technical details of a vertical roller gate are already described in the paragraph about weir Lith.

Figure A-44 Weir Borgharen; gates closed and the flap gates are regulating the discharge (photo taken by: den dzjow; www.flickr.com)

“Balgstuw” Bornsebeek and Ramspol Three inflatable barriers are yet constructed in the Netherlands; two in the Bornsebeek and one near the village of Ramspol. The scale of the weir of the Bornse beek (several metres) and Ramspol (a width of 350 metres subdivided in three ‘balgstuwen’ of 80 metres long) differ but the working of the weirs are quite the same. Only the operating buildings are visible when the weir is not in operation; the ‘balg’ is located in its recess in the sill. The weir in the Bornsebeek is inoperative when the discharge in the Bornsebeek is sufficient

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high and a sufficient water level is present. Commercial shipping is not present in the Bornsebeek; the only ‘navigation’ are canoeists in summertime which are able to slide over the fabric in their canoe. The barrier of Ramspol is inoperative during normal conditions. This barrier is inflated with a mixture of water and air during storm conditions. The set-up of water caused bij a north-western storm would otherwise result in floods near the village of Genemuiden and the Kampen. Air is pumped into the ‘balg’ during the first phase of the inflation operation. Secondly water is pumped in to the ‘balg’ and at last air is used to tension the ‘balg.’ The height of the fully inflated weir above the water level is four metres. The fabric of the balg is stored in recesses in the foundation during normal conditions. In this way navigation is able to pass the Ramspol barrier during normal conditions (Volkskrant, 2012).

Barrier Hollandse IJssel The area around the Hollandse IJssel is the lowest lying area and one of the densely populated areas of the Netherlands. During a storm surge the water is not able to flow towards sea due to the raised water level at the Noordzee and in the delta. This increase would cause flooding; levees along the Hollandse IJssel would be overtopped. Two movable gates (placed after each other; one gate is the spare gate in case of failure) of 80 metres wide are lowered by cables to limit the water level of the Hollandsche IJssel when the water level exceeds +2.5m NAP. The navigation with limited height is not hampered by these barriers. Vessels are able to cross the barrier by a lock when the barrier is closed and are able to sail underneath the gates during ‘normal’ situations (Stichting Deltawerken, 2012)

Small weirs A lot of small weirs are constructed beside the larger weirs described above. Only two weirs are presented in this paragraph, because every small weir has nearly the same design. Only weir ‘de Haandrik’ near Gramsbergen and the weir of Hardenberg are described in this section. The weir ‘de Haandrik’ is located in the Vecht near the village of Gramsbergen. The water head at the weir is 1.5 to 2 metres in periods of drought. The weir is regulated by flap gates regulated by cables. The flap gate is connected to the sill of the weir and the cables are connected to the top side of the weir. The top level of the gate decreases by extending the cable, and increases by decreasing the length. In this way the discharge over the weir is being regulated. A small waterpower station is realised next to the weir. A propeller with a diameter of 1.60 metres is applied in the station. The nominal discharge is 7 m3/s and the head is about two metres. The capacity of this small weir is 530.000 KWh/year which is enough for 180 households (Watertononline, 2012). The weir of Hardenberg is build according the same design as weir ‘de Haandrik.’ But now pistons are applied instead of cables. The pistons are connected to the bridge. The discharge over the weir is controlled by adjusting the length of the piston. The weir is shown in Figure A-45.

Figure A-45 Weir 'de Haandrik' (left) and weir Hardenberg (right) (Google maps)

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Design study: open fabric weirs A variation of the Balgstuw is an open fabric weir (also called a parachute dam). The possibilities of fabric barriers are investigated in two master thesis (Hoogenboezem, 1998) and (Ziel, 2009). A parachute barrier blocks the waterway by a fabric which is hoisted or pulled into place. The designs differ in the connection of the cables to the abutments. First the design described by Hoogenboezem is presented and secondly the design of van der Ziel.  The forces resulting from the water level difference are lead towards the abutments by cables. The fabric is held in place by floaters which are connected to the upper cable. The fabric could be stored in a recess in one of the abutments. First the barrier would be pulled out of the recess by winches towards the other side of the waterway in case of a high river runoff or a storm surge from sea. Secondly the cables are connected to the vertical winches. These vertical winches pull down the lower cable towards the bottom. The hydrostatic water pressure would ‘push’ the fabric tightly to the bottom and the abutments.  Another option is the integration of the barrier with a bridge. The barrier is stored underneath the bridge and lowered by cables when necessary. The cables are pulled down and the barrier is tightly connected to the bottom and the abutments. The fabric is held in place by cables which are connected to a support structure like a bridge.

Two impressions of an open fabric barrier/weir are shown in Figure A-46. The left picture is an scale model of an open fabric barrier with floaters and the second picture is a sketch of an open fabric barrier with cables.

Figure A-46 Open fabric barriers; scale model with floaters (left) and impression of a design with cables and bridge (right)

Some opportunities of fabric barriers are:  The fabric is flexible and can be folded. In this way the storage volume of the gate can be reduced compared to a normal (steel or concrete) barrier.  The material is flexible and could divert the loads without bending moments. The behaviour of forces and moments is more efficient compared to a normal gate.  The deterioration of the fabrics is much smaller compared to steel and concrete. So less maintenance has to be performed on this material.

Design study: ESA, FRP gates (ARCADIS & DHV, 2010) A new material which could be used for the weir design are fibre reinforced polymer [FRP] composites. Some bridges and locks are already constructed with this material but the application of this material in hydraulic structures is limited these days (2012). The dimensions of some already designed bridges are shown in Table A-6. The design of a bridge could be compared with the design of a hydraulic gate, the only difference is the direction of the force (a vertical traffic force instead of a horizontal hydraulic

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pressure force) and the magnitude of the force. From the shown projects in Table A-6 could be concluded that a fibre reinforced structure could be applied as a hydraulic gate. Already built FRP hydraulic gates are the new mitre gates of the lock ‘spieringsluis.’ The lock door is 6 metres high and 3 metres wide and replaced an old wooden mitor gate.

Table A-6 Built synthetic bridges

Load Bridge Span [m] [Traffic class] Bridge Oosterwolde 60 12 bridge deck Marebrug 45 9,5 Garstang Mount Pleasant M6 Bridge, Lancashire 45 2 Bridge Den Dungen 45 10 Cyclists and pedestrian bridge Dronten Cyclists and pedestrians 24.5

ARCADIS has examined whether a FRP gate could be applied in the new sluices of the Afsluitdijk. A hydraulic gate of 6 metres high and 35 metres wide with curved convex walls was designed for the sluices. The thickness of the outer plates is 180 mm with a wall thickness of 50 mm. The weight of the gate equals the buoyancy force. The designed gate is presented in Figure A-47.

Figure A-47 Fibre reinforced polymer gate of the new sluice complex

Some other opportunities of a FRP gate are:  A reduction of the dimensions of the hoisting/lifting equipment was possible due to the limitation of weight.  The strength of a FRP gate is higher compared to the strength of a traditional steel gate.  Deterioration of the material is much slower compared to steel and concrete. o Less maintenance has to be performed.  Environmental impact of a FRP gate is lower compared to a steel and concrete gate.  Many options are available for the conservation of the gate; a wide range of additives are developed.

But also some drawbacks are present:  The limited weight of the gate could result in unwanted vibrations.  The stiffness of the material and structure is lower compared to steel and concrete.  FRP is a new material, so the behaviour on the long term (30 to 50 years) is not fully known.

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Design study: high strength concrete weir (Sloten, 2012) A high strength concrete gate has been designed the compartmentalisation for the Amstel. This would reduce the consequential damage caused by a breached levee. A rotatable gate located in a recess in the foundation would rotate over an angle of 900 to close off the water way. The design is illustrated in Figure A-48. The designed gates are 40 metres wide and are able to divert a water level difference of 3.85 metres; the maximum allowable deflection of the gate is 100mm which was the governing design parameter of the gate. This design is nearly equivalent to the design of the Thames barrier in London and the Emssperrwerk near Emden in Germany. The difference is the application of high strength prestressed concrete instead of steel or normal reinforced concrete. The high strength concrete gate has two advantages compared to the steel gates which are applied in London and Emden:  The maintenance costs of the high strength concrete gates are lower than a gate made steel.  The carbon footprint of the high strength concrete gate is also lower compared to a traditional steel gate.

So the application of high strength concrete in hydraulic structures has some benefits compared to the traditional steel gates. The application of high strength concrete gates is not just a case study. A small lock gate made of high strength concrete has already been constructed in IJburg, Amsterdam.

Figure A-48 Rotatable high strength concrete barrier

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B. Information and methods

This appendix elaborates the theories and methods which are applied in this research. The systems engineering methodology is shortly described in B.1. Secondly, the multi criteria analysis method (MCA) is described in B.2. This analysis method is used for the weighting of the variants which are elaborated in each design level. Thirdly, information used for the configuration design is presented in B.3. The detail of the present situation is listed in several tables and presented in several figures. In B.4 the properties of fibre reinforced polymer structures are described. The fabrication process is highlighted and material properties used for the design of the gate are presented.

B.1 SYSTEMS ENGINEERING THEORY

The systems’ engineering (SE) approach is not a ready to use method. It is a toolbox of methodologies which can be applied during a project. Every SE manual describes a (slightly) different approach of SE. A selection of methodologies from different manuals has been sought for in the preparation phase of the graduation research. Methodologies used in this design research originates from the guideline “SE-wijzer” (BAM, 2008), guideline “Systems Engineering – Doen” (ARCADIS, 2008), guideline “Leidraad SE” of Rijkswaterstaat (Rijkswaterstaat & ProRail, 2009), handbook “Systems Engineering Fundamentals” (Department of Defense, 2011), and “Handboek specificeren” (CROW, 2011).

A project is decomposed in several design levels. Every design level represents a certain scale of the project; the first level concerns the design of the large scale system (for example: river branches in the Netherlands), the second level concerns the design of concepts (for example: canalization concepts of the Nederrijn-Lek), and the third level concerns the design of a specific structures (for example: weir designs which can be implemented in the reach Nederrijn-Lek). The design process is not limited to three levels, but can be extended with sequential levels. Every design level consists of three design steps and a design check. A close up of the iterative design process (which consist of the requirements analysis, functional analysis, design synthesis, and verification & validation) is given in Figure B-1. The process output of a level indicated in Figure B-1 is the input for the sequential design level. Each step is clarified in the following enumeration (ARCADIS, 2008):  Requirements are defined by analysing the project area and the stakeholders. o The input of the requirement analysis is a (detailed) description of the project area, the surroundings of the project area, an overview of the stakeholders, and the stakeholder’s needs. The actual requirement analysis identifies and defines the functional requirements, internal constraints, external constraints, boundary conditions, RAMS aspects, etcetera which are applicable for the project. A structured list of requirements is presented at the end of the requirements analysis.  Functions are derived from the project system and the requirements. Requirements elaborated in the first step are allocated to objects which have to perform a certain function.

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o Functions of the system are analysed and decomposed form a higher level function into several lower level functions. The first product of the functional analysis is an overview of functions which the system has to perform. The second step of the functional analysis is the determination of the flow of functions within the system. The functions which the system has to perform are ordered sequentially using a functional flow block diagram or using other methods. The functions are converted/clustered into objects which have to perform a certain function. Finally, the requirements which are derived in the previous step are allocated to the derived objects. The result of this allocation is an overview of requirements per object.  Elaborating the design with use of the allocated requirements and functions. o Several variants are elaborated for the design syntheses. The designs are based on the allocated requirements per object. The variants are compared with each other using decision making tools like a MCA or a trade-off matrix which is described in B.2. The best option is chosen for further elaboration in the next design level.  Verification (“is the job being done right?”) and validation (“is the right job being done?”) of the design. o The chosen design has to be verified and validated with use of the measures of effectiveness (MOE’s).

Figure B-1 Design steps within a level (Department of Defense, 2011)

B.2 MULTI CRITERIA ANALYSIS METHOD

A multi criteria analysis (MCA) is used to select the most promising variant from other variants. This is a method which is used for making rational choices between discrete alternatives on basis of more than one decision criteria. The variants which have to be analysed are defined during a foregoing design synthesis. Variants are made and elaborated in the design synthesis and have been verified. A variant is not an option when it does not meet one or more requirements; only variants which meet every requirement are taken into consideration in the MCA.

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The variants are scored for the predefined design criteria. The criteria are based on lists of environmental impact assessments performed by ARCADIS, a check list established by the British government (Department for Communities and Local Government, 2006), and a check list established by PIANC (PIANC, 2006). A rating per criteria is given per variant in order to compare the variants. The scale of the rating ranges from 1 to 5 for which 1 is ‘very bad’ and 5 is ‘excellent.’ Every aspect is equally important for an unweighted comparison, so in this case no weight reference is specified. However, a weight reference is specified in a weighted comparison in order to take the ‘relative impact’ of each criterion into account. For example: the yearly yield of the aspect commercial shipping could be larger with respect to the yearly yield of recreational boating in the opinion of the client, so the weight factor of the aspect commercial shipping is larger with respect to the weight factor of recreational boating. An unweighted assessment is generally executed by engineering firms. An engineering firm has to advice the client about the impacts on the surroundings of distinct options. The output of the assessment is a factual overview of the impacts on its surroundings. A weighted assessment is executed by the client (this would have been Rijkswaterstaat or the government for a real design project). They have to determine the importance of certain aspects with respect to other aspects (Oosterwijk, 2012). The costs of each variant are not taken into account for the determination of the scores because all criteria except costs are a measure for the performance of the variant. For some occasions it is possible to spend more money in order to realise a better performance of the variant. The costs and (weighted) rating are subsequently plotted in a cost/rating graph as indicated in Figure B-2. With use of the cost/rating graph a distinction can be made between the variants. Figure B-3 indicates the method for which variants A, B, C, and E are assessed with respect to D; four sections can be identified with respect from variant D in this figure (Voortman, 2011):  quadrant bottom-right o The variants (in this case only C) are more expensive and do have a lower score with respect to D. Therefore the variants in this quadrant are always rejected.  quadrant top-left o The variants (in this case only B) are cheaper and do have a better score. Therefore the variants in this quadrant are always an option.  quadrant top-right o The variants (in this case only A) are always better and do have higher costs. The variants do have to be taken into consideration.  quadrant bottom-left. o The variants (in this case only E) do have a lower score and are cheaper. Therefore these variants have to be taken into consideration.

In the example variant C and D are not an option which is determined by repeating this method for different variants as base point (so not only for D). The set of preference variants are A, B, and E for which A has the highest score, for which E is the cheapest, and B has the best cost/score rating.

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Figure B-2 Cost/rating graph (Voortman, 2011)

Figure B-3 Cost/rating graph used for assessment (Voortman, 2011)

B.3 MODEL DESCRIPTION FOR CONFIGURATION VARIANTS

The impact on its surroundings of the variants has to be determined by comparing the impact of the distinct variants. The amount of work per variant is indicated by comparing the present and the future water levels generated by the weirs and the impact of the changed water levels on its surroundings. A model is made in which the new configuration of weirs is modelled. Bridges, harbours, pumping stations, inlets, etcetera are indicated in figures to determine the impact on the surroundings. The model of the present situation is given as a reference for the new configuration variants.

B.3.1 INPUT DATA AND BOUNDARY CONDITIONS OF THE PRESENT SITUATION

Locations of the weirs and the water levels along the Nederrijn and Lek are presented in Table B-1. This table contains the design parameters of the model which are indicated in red. The water level at the

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upstream boundary (IJsselkop) is determined by the rate of flow which is diverted into the IJssel. The water level at downstream boundary is determined by a combination of the tide and the river runoff. A water level of -0.44m NAP is present at weir Hagestein during averaged low water and OLR (Dutch: overeengekomen lage Rivierafvoer) conditions which is the governing situation for commercial shipping.

Table B-1 Design parameters; locations of weirs Distance from the Dam level Location IJsselkop + NAP [km] [m] Upstream boundary 0 8,3 Weir 1 upstream 12,69 8,3 Weir 1 downstream 12,69 6 Weir 2 upstream 43,34 6 Weir 2 downstream 43,34 3 Weir 3 upstream 66,79 3 Weir 3 downstream 66,79 -0,44 Downstream boundary (OLR) 110 -0,44

The first set of boundary conditions are the vertical clearances of the vessels. The design draught used to be 2.80 metres and the designed depth used to be 4 metres (Rijkswaterstaat, n.d.). The present maximum draught of the ships sailing through the Nederrijn is 3 metres (Arnhem, 2012). The actual maximum permitted ship draught in between weir Driel and the connection of the Amsterdam-Rijnkanaal is just 2.70 metres for the last three months.3 This limitation is caused by large shoals located in this reach which are limiting the draught (Arnhem, 2012). The draught of a Va class vessel varies from 2.5 metres to 4.5 metres according to the CEMT classes. The design draught of the Nederrijn and Lek is chosen to be 3.0 metres given the low shipping intensity and the river improvements which are necessary for increasing the draught. A more detailed research has to be performed in order to determine the economical optimal draught, but this is beyond the scope of this research. A keel clearance percentage of 40% has to be present (Rijkswaterstaat, 1999). So, the design depth of the new situation has to be 4.20 metres when the Nederrijn and Lek is equipped for vessels with a draught of 3 metres. The maximum draught of recreational boating is 1.5 metres (class AM). A minimum keel clearance for recreational boating is 10% to 20% of the draught (Rijkswaterstaat, 2011) so a depth of 1,8 metres has to be present. Furthermore a vertical clearance of 9.10 metres has to be present. The set of boundary conditions are shown in Table B-2.

Table B-2 Vertical clearances

vertical clearance Height/depth restrictions [m] Depth commercial shipping 4,2 Depth recreational boating 1,8 Maximum vertical clearance 9,1

The sequential sets of boundary conditions are less stringent with respect to the vertical clearance boundary conditions because the sequential boundary conditions can be changed; changes in of bed levels, heights of bridges etc. are possible by adapting the present situation. The first set of less stringent boundary conditions are minimum water levels for commercial shipping and recreational boating. These water levels are based on the maximum river bed levels presented in the reference book ‘Vaarwegen in Nederland’ on which the online application VIN is based on (Rijkswaterstaat, 2011). The maximum river bed levels are chosen for the design in order to guarantee a safe passage of the vessels sailing through the

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Nederrijn and Lek. The minimum water levels which are necessary for commercial shipping and recreational boating are determine by adding the maximum vertical clearance to the maximum river bed level. The maximum measured river bed levels, the minimum water level for commercial shipping, and the minimum water level for recreational boating are presented Table B-3.

Table B-3 Minimum water levels for commercial shipping and recreational boating Distance Maximum Minimum water level Minimum water level from the riverbed level commercial shipping recreation boating River bed levels IJsselkop +NAP +NAP +NAP [km] [m] [m] [m] IJsselkop 0 3,75 7,95 5,55 Weir Driel 12,69 2,8 7 4,6 Weir Amerongen 43,69 -0,7 3,5 1,1 Weir Hagestein 66,69 -3,7 0,5 -1,9 Schoonhoven 90,69 -5 -0,8 -3,2 Nieuwe Maas 108,84 -5,2 -1 -3,4

These maximum riverbed levels are verified using a data set of SIMONA_wbr08_04. This is necessary because it is not evident what the maximum and minimum bed levels are which are presented in the report ‘Vaarwegen in Nederland.’ Four options for the riverbed levels are possible, which are:  an average minimum and an average maximum bed level  a maximum and a minimum value of the data set of a cross section  the maximum bed levels at the banks.  the maximum end minimum bed level of the fairway.

The bed levels of the fairway are extracted from SIMONA-wbr08_04 and are presented in blue in Figure B-4. On basis of Figure B-4 can be concluded that the maximum river bed levels presented in Table B-3 are representing the maximum fairway bed levels. Therefore the bed levels presented in Table B-3 are used.

Figure B-4 River bed levels (SIMONA, 2012)

The second set of less stringent boundary conditions is the height restrictions generated by the present bridges over the Nederrijn and the Lek. The maximum water level of the Nederrijn and Lek is calculated by subtracting the vertical clearance from the bridge levels. The results are shown in Table B-4. Bridges have to be adapted when the designed new water level is higher with respect to the maximum water levels presented in the last column of Table B-4.

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Table B-4 Maximum water levels

Distance from Bridge levels Maximum water level including clearance Bridges the IJsselkop +NAP +NAP [km] [m] [m] Andrej Sachorovbrug 1,69 25,28 16,18 John D. Frostbrug 4,43 22,85 13,75 Nelson Mandelabrug 5,36 24,58 15,48 Railway bridge 9,27 21,16 12,06 Bridge A50 15,88 22,51 13,41 Bridge Rhenen 30,55 20,74 11,64 Railway Bridge 59,74 16,2 7,1 Hagesteinse Brug 68,67 15,8 6,7 Lekbrug 71,3 15,54 6,44 Jan Blankenbrug 71,36 17,28 8,18

The third set are the dimensions of the present locks located along the Nederrijn and Lek. A minimum water level at the locks is determined by adding the keel clearance and the draught of vessels to the level of the sill. Furthermore, a maximum water level is determined for the Barrier Ravenswaaij, Prinses Irenesluizen, and the Prinses Beatrixsluis. This barrier and the locks are equipped with a vertical lifting gate, which imposes a height restriction. The maximum water level at these locks is calculated by subtracting the vertical clearance from the level of the bottom of the gate. The minimum and maximum water level of the locks is presented in Table B-5 and Table B-6. Locks have to be adapted if the minimum or maximum water level of a lock is undercut or exceeded for the new situation.

Table B-5 Minimum water levels at the locks Distance from the Sill height Minimum water level Ship Lock Sills IJsselkop +NAP +NAP [km] [m] [m] Prinses Marijkesluizen 48,95 -2,35 1,85 Barrier Ravenswaaij 48,95 -2,35 1,85 Prinses Irenesluizen 48,95 -4,6 -0,4 Prinses Beatrixsluis 67,85 -4,5 -0,3 Grote Sluis 70,07 -3,5 0,7 Koninginnesluis 70,07 -3,3 0,9

Table B-6 Maximum water levels at the locks Distance from the Door height maximum water level Ship Locks Door heights IJsselkop +NAP +NAP [km] [m] [m] Barrier Ravenswaaij 48,95 14,65 5,55 Prinses Irenesluizen 48,95 16,75 7,65 Prinses Beatrixsluis 67,85 14,4 5,3

The fourth set are commercial and recreational harbours located along the Nederrijn and Lek. These harbours are subdivided into four groups which are: large commercial harbours, small commercial harbours, large recreational harbours, and small recreational harbours. The water levels of the harbours are set to be equal for the ‘normal dammed’ water levels. The harbour characteristics are presented in Table B-7, Table B-8, Table B-9, and Table B-10. Adaptations are necessary when large deviations occur at specific locations and the available depth is not sufficient anymore

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Table B-7 Characteristics of large commercial harbours Distance from the Present normal water levels Large commercial harbours IJsselkop +NAP [km] [m] Arnhem 3,41 8,3 Rijnhaven Wageningen 24,75 6 Wijk bij Duurstede 47,82 3

Table B-8 Characteristics of small commercial harbours Distance from the Present normal water levels Small commercial harbours IJsselkop +NAP [km] [m] Heteren (paper factory Parenco 18,83 6 Lienden; storage area for fertilizer 30,55 6 Lienden; Skinspark BV 31,25 6 Schoonhoven; stone factory 91,09 -0,44 Groot Ammers; Hartog olie BV 94,26 -0,44 Bergstoep; animal food factory 96,4 -0,44 De Boer beton 104,19 -0,44

Table B-9 Characteristics of large recreational harbours Distance from the Present normal water levels Large recreational harbours IJsselkop +NAP [km] [m] Arnhem 3,41 8,3 Wageningen (water sport centre VADA) 24,75 6 Cut of bend near Weir Amerongen 45,96 3 Vianen 73,08 -0,44 Schoonhoven 91,09 -0,44 Streefkerk 98,69 -0,44 Krimpen aan de Lek 107,59 -0,44

Table B-10 Characteristics of small recreational harbours Distance from the Present normal water levels Small scale recreational IJsselkop +NAP harbours [km] [m] Beusichem 53,19 3 Culemborg 59,34 3 Nieuwegein 70,54 -0,44 Bolsweerd 76,08 -0,44 Nieuwpoort 88,69 -0,44 Lekkerkerk 102,89 -0,44

The fifth set is the characteristics of pumping stations and sluices located along the Nederrijn. The locations of the pumping stations and inlets and the present dammed water levels are presented in Table B-11 and Table B-12. The amount of work for adapting the pumping stations and sluices has to be indicated with use of the changed water levels of the variants.

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Table B-11 Characteristics of pumping stations Distance from Current water levels Pumping stations the IJsselkop pumping stations +NAP [km] [m] Pumping station Drielsedijk 9,07 8,3 Pumping station Mr. G.J.H. Kuijk 23,51 6 Pumping station De Koekoek 84,28 -0,44 Pumping station Elshout 107,68 -0,44

Table B-12 Characteristics of sluices Distance from the Current water levels Sluices IJsselkop sluices +NAP [km] [m] Grebbesluice 29,49 6 Tollewaard 34,86 6 Kromme Rijn sluice 47,61 3 Inlet Haagse Waterleiding 97,47 -0,44

The sixth set are the levels of the river banks and summer levees. The impact on the surroundings of the variant is indicated by the change in water level along the river. River banks and summer levees have to be raised when the normal water level of a variant is higher with respect to the present levels of the river banks and summer levees.

Table B-13 Levels of river banks and summer levees (Actueel Hoogtebestand Nederland, 2012)

Distance from the Floodplains summer levees Levees and Floodplains IJsselkop +NAP +NAP [km] [m] [m] IJsselkop 0 10,5 10,7 Arnhem 5,36 10,7 11 Weir Driel 12,69 9,3 11 Renkum 18,85 7,5 10,2 Wageningen 24,75 7 9,5 Weir Amerongen 43,34 6,3 8,1 Wijk bij Duurstede 48,95 5,2 6,8 Weir Hagestein 66,79 3,5 4,8 Ameide 81,69 1,8 3,4 Schoonhoven 91,09 1,2 3,4 End of the Lek 108,69 1,2 3,4

B.3.2 VISUALIZATION OF THE CONFIGURATION

The data described in the previous chapter is visualized using graphs. The graphs are a schematized cross section of the Nederrijn and Lek. The figure representing the minimum water levels for navigation, the minimum water levels for recreation, the maximum river bed level, and the locations of the harbours indicated for the present dam level is included in the main report. Figure B-5 indicates with the maximum water levels of the spanning bridges with a red dot. Figure B-6 indicates the pumping stations (blue dots) and the inlet stations (green dots). Figure B-7 indicates the minimum and maximum water levels of locks located along the Nederrijn and Lek. The impact on the surroundings of a variant is determined using these graphs as a reference. The amount of work for adapting the existing harbours is estimated using the overview figure included in the main report. The minimal depth of harbours is equal to the depth of the waterway and the keel

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clearance or a depth which equals the draught and a keel clearance of 1 metre (necessary for the high velocities generated by the propeller) (Rijkswaterstaat, 2011). The depth is presented in Equation B-1. The keel clearance of a recreational harbour is 10% to 20% of the draught, so a depth of 1.8 metres must be present at the recreational harbours. Harbours have to be adapted when the depth of the new situation is lower compared to the minimal depth.

Equation B-1 Calculation of the harbour depth

Figure B-5 Bridges along the Nederrijn and Lek

Figure B-6 Pumping stations and sluices along the Nederrijn and Lek

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Figure B-7 Minimum and maximum water levels of locks along the Nederrijn and Lek

The dammed water level presented in Figure B-7 has to remain in between the minimum water level dots (indicated in red) and the maximum water level dots (indicated in purple) for the functioning of the weirs. The water levels for OLR are already lower than the minimum water levels for the Lekkanaal and Merwede Kanaal. Furthermore, an open connection between the Betuwepand and the Nederrijn could be maintained if the new water levels remain in between +3.00m NAP and +5.55m NAP (Rijkswaterstaat, 1998).

B.4 FIBRE REINFORCED POLYMERS (FRP)

Extra information which is not presented in the main report for FRP is included in this section. First, the reinforcements are described in B.4.1; secondly, the resins are described in B.4.2. Thirdly, the cores of sandwich panels are described in B.4.3. The fabrication processes of FRP are described in B.4.5 and at last, the material properties are described in B.4.6. The following literature is used in this appendix: (Kolstein, 2008), (Stichting CUR, 2003), (Chlosta, 2011), (Nijhof, 2006), (Zenkert, 1995), and (Maas, 2011).

B.4.1 REINFORCEMENT

B.4.1.1 GLASS FIBRE REINFORCEMENT

Five distinct types of glass fibre reinforcements are available namely A, C, E, R, and S glass. ‘A’ glass used to be the normal glass fibre, but now the E glass is normally used. E glass has good electrical, mechanical, and chemical resistance properties. C glass has very good chemical resistance properties and is therefore used for surface tissue manufacture. R and S glass do have higher strength properties with respect to E glass and are therefore used for aerospace technology. R and S glass is more expensive with respect to E glass. In civil engineering the codes and design manuals are written for E glass. Therefore the lower strength E glass is chosen for the design of the gates.

B.4.1.2 POLYARAMID FIBRE REINFORCEMENT

Polyaramid fibres have a high tensile modulus, high tensile strength, low weight, high impact resistance, high resistance to abrasion, good creep rupture, and mechanical & chemical resistance for a wide range of temperatures. Carbon and glass fibres fail brittle, but polyaramid fibres fail ductile. The drawback is the low compressive strength of polyaramid fibres. This can be solved by producing a hybrid fabric in which carbon or glass fibre bears the compression and the Polyaramid fibres with tension.

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B.4.1.3 CARBON FIBRE REINFORCEMENT

Carbon fibres have a higher tensile strength with respect to glass fibres and a lower tensile strength than polyaramid fibres. Furthermore, the tensile modulus is higher with respect to polyaramid fibre, so a higher stress is obtained for a lower strain. Carbon fibre is presently used in aerospace technology. No codes or design manuals are yet available for the civil engineering. Therefore, carbon fibre is not used for the design.

B.4.1.4 APPLICATION OF REINFORCEMENT

Reinforcements can be applied in various manners. Several fabrics can be distinguished which are:  Continuous filament rovings. Two types of continuous filament rovings are present, namely a direct roving which consists of parallel filaments and an assembled roving which consists of aspun filaments. Rovings are delivered on a spool and can be used for pultrusion and continuous filament winding, and weaving.  Chopped strand mat (CSM) are composed of chopped rovings which are spread uniformly on a stage. The chopped pieces of roving are bonded together using binders. Chopped strand mats are usually used for hand lay-up, hot pressing, and continuous plate production.  Continuous filament mat (CFM) consists of multiple layers of continuous glass fibres. The glass fibres are bonded together using a binder. After fabrication, the mats are rolled on spools and transported towards the next production plant. CFM’s are used for pultrusion and continuous production methods.  Woven fabrics are made by weaving fibres into a fabric; the fibres are orientated normal to each other. Several kinds of weaves can be produced and each weave has its own properties. The fibres can be orientated in a plain weave. The amount of strands per direction is equal to each other for this weave. The fibres can also be weaved into a unidirectional weave. In this case nearly all the fibres are directed towards one direction and a limited amount of strands is used for holding the gross of the fibres in its place.

B.4.2 RESINS

The fibres have been bonded together to obtain a material for structural applications. The material is able to bear shear and compression forces when they are ‘glued’ together. Furthermore, the fibres are protected from the outside influences (moisture etc.). The fibres can be ‘glued’ together by metal, ceramic, carbon and polymers. Polymers are mainly used as resin for civil engineering applications. A resin is sought for which cures at room temperature because curing ovens are not ready available for large FRP moulds. The applicable resins are:  Unsaturated polyester resins which are syrups consisting of polymer chains dissolved in a reactive organic solvent. A solid three-dimensional structure is obtained by adding a catalyst. Polyesters can be cured at room temperature which makes this material applicable for very large structures. Furthermore the costs of this resin are lower with respect to the vinyl ester resins and epoxy resins.  Vinyl ester resins are an extension of the polyester resin. Vinyl ester resins do have fewer ester linkages with respect to polyester and do have improved chemical resistance. However the costs of the vinyl ester are higher with respect to the polyester resins. Vinyl ester resins can also be cured at room temperature.  Epoxy resins do have the best mechanical properties for FRP. A cured epoxy is not only resistant to moisture and chemicals but does have also good electrical insulation properties. The costs of epoxies are higher with respect to unsaturated polyester resins and vinyl ester resins. The epoxy

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resins can cure at room temperature but the properties are much better when the cure temperature is above 40°C and preferably at 60°C

B.4.3 CORES

Sandwich constructions are often used for a composite structure in order to increase the structural efficiency. Three kinds of cores are presently used namely the foam core, a honeycomb core and a solid core. The properties of each core are described in the following enumeration.  Three kinds of foam cores could be applied namely: o A non-structural foam core used to support the formers for producing a desired shape. Foams can be made from many plastic materials. o Structural foams do have a favourable strength and stiffness to weight ratio. They provide a necessary structural performance for low costs. The skin bears the main loads in tension and compression for a sandwich in flexure. The core is able to bear the transverse shear loads. The core also acts as a spacer for the skin during production. o Reinforced foams are applied in order to increase the stiffness parameters of the foam to obtain a higher performance sandwich. o Honeycomb cores are made of a thin sheet material shaped like a honeycomb. The structural performances are high in direct compression and shear. A honeycomb can be made of glass fabric, metal, paper and other thin sheet materials. o Cores can also be made of solids like wood, bonded micro spheres for lightweight structures. The core of a solid does not have structural properties.

B.4.4 INFRACORE® PANELS PATENTED BY FIBERCORE

Based on: (ARCADIS & DHV, 2010), (FiberCore-Europe, 2012), and (Snijder, 2012).

The disadvantage of a sandwich plate is the loss of bond between the face and the core due to (repetitive) impact due to ship collisions or wave impact. Over a small area the bond is lost between the core and the face and the loss of bond between the core and the face spreads over time. The area of failure is indicated by the red line in Figure B-8. The development of the loss of bond results in a loss of strength properties and to failure.

Figure B-8 Failure of a sandwich plate due to repetitive impact (ARCADIS & DHV, 2010)

The contact area of the face and the core is divided into several square boxes to solve the loss of bond between the face and the core. Furthermore, the vertical walls bears the shear force The alignment of the glass fibre mats is indicated by the green lines in Figure B-9. The tubes are glued together using a resin. However, the resin could fail due to repetitive impact resulting in a loss of bound between the tubes. The failure plains of a glued box beam are indicated by the red lines presented in Figure B-9.

Figure B-9 Failure of a box beam (ARCADIS & DHV, 2010)

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An improved design of the box beam is the InfraCore® plate of the company FiberCore. The same box- beam design is used; however, the glassfibre mats are interwoven between boxes in a Z-profile as indicated by the green lines in Figure B-10. In this way, the failure of the resin in between the boxes is not present anymore. Also for this design, the vertical walls bear the shear force. The present maximum outer wall thickness is 5cm but could be stretched to 7cm. This increase of wall thickness has to be investigated in further detail. The thickness of the vertical walls is in between 4 and 8mm. These walls have to bear the shear force and are used as ‘transport channels’ for the fabrication of the InfraCore® panel. The vertical walls are spaced at every 15 centimetres. A non-structural polyurethane foam (known as PUR) is used as formwork. Currently, FiberCore is investigating the application of low density concrete as core. The advantage of low density concrete is the higher volumetric weight which can be used as ballast in order to prevent floatation of FRP hydraulic structures.

Figure B-10 Glass fibre mat orientation of the InfraCore® beam (ARCADIS & DHV, 2010)

B.4.5 FABRICATION OF FIBER REINFORCED POLYMER PLATES

The fabrication process consists of 5 steps which are presented in the following enumeration:  Mixing the resin and activator.  Positioning reinforcement.  Impregnating the reinforcement with resin.  Dispensing the resin into the mould.  Curing.

These steps can be executed by different techniques with its own advantages and disadvantages. The most used techniques are described in this section.

B.4.5.1 OPEN MOULD PROCESSES

Open mould processes are the simplest FRP production techniques. No pressure and heat is necessary for producing a laminate for this technique. Three common open mould production techniques are described in the following enumeration.  Hand lamination is the oldest production technique. The flexibility of this technique is large, but the process is messy, labour intensive and difficult to control. First a gel coat is applied to the mould using a soft roller or brush. Secondly a layer of resin is placed in the mould. A reinforcement layer is placed on this layer of resin. This process is repeated until the required build-up is achieved.  Saturation is a more advanced technique with respect to hand lamination. The difference with hand lamination is that the resin is sprayed onto the reinforcement layer with a spray gun.  Resin and chopped fibres are spayed onto the mould for the spray up technique. A continuous roving is guided to the gun where it is chopped into pieces. Consolidation of the layer is done by hand. This technique controlled by a machine which results in a better quality laminate. The automated spray up technique is called auto spray-up.  Filament winding is used for the production of pipes, circular sections, and tanks. The mould is fixed on a rotating shaft and the continuous filament roving which is soaked by resin is transported to the mould. The mould rotates and the soaked roving is wounded on the mould till

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the right thickness of the laminate is obtained. The thickness of the laminate is controlled by the rotation speed of the mould.  Spray winding is a combination technique of spray-up and filament winding. A layer of chopped fibres and resin is prayed on the rotating mould. An extra layer of continuous strands is interleaved with a continuous fibre strands which are also winded on the mould to build op thickness.  Centrifugal casting is used for obtaining hollow components. The rotating mould is located at the outside of the new product. A spray gun and fibre chopper is located in the tube and sprays the resin and fibres on the inside of the mould which rotates at high speeds. The air is pressed out of the resin and fibres and the FRP is consolidated well due to the high centrifugal forces.

B.4.5.2 CLOSED MOULD PROCESSES

The following closed mould processes are available for producing a laminate. The closed mould process has a higher production rate but is also more expensive with respect to the open mould process.  Vacuum bag is the simplest closed mould process. Resin and reinforcement are placed on the mould by hand laminating. When the resin and reinforcement are places, an airtight film is placed over the laminate and the air in between the film. The atmospheric pressure is now applied over the surface of the laminate and the air is evacuated. Manual rolling can be applied for better consolidation.  Pressure bag is nearly the same process as vacuum bag processing. But now it does not use the vacuum but an inside pressure bag. Higher pressures can be obtained with respect to a vacuum bag, so better properties can be obtained. The pressure bag is located in between the film which covers the laminate and a stiff cover which is connected by clamps on the mould.  Autoclave uses both techniques. The to be produced laminate is located in a vacuum bag assembly which is located in a pressurized oven, called the autoclave. For producing the laminate, a vacuum, pressure, and heat are applied simultaneously and a very high quality, high fibre content, and low void content FRP is obtained. However the costs are very high and the dimensions of the laminate are limited.  Leaky mould. Resin and reinforcement are placed on the hollow female mould by hand laminating. A male mould is clamped on the female mould and a cavity is created which has the exact shape of the finished component. Excessive resin is able to flow out of the mould at the sides of the mould. Accurate dimensions and good quality finish can be obtained by the leaky mould process.  Cold press is nearly the same production process as the leaky mould process. The difference is a hydraulic press which is capable of exerting pressure. The pressure is used for distributing the resin over the reinforcements and pushing the air out of the laminate. The laminate is finished when the resin is hardened. Accurate components can be produced by this technique with higher performances with respect to leaky mould.  Hot press is nearly the same production process as the cold press, however now heat is used for increasing the rate of production. It is also possible to use prepregs and sheets applied with chopped fibres. The fibres and resin have to be able to flow under the action of heat and pressure. Fine details and close tolerances are obtained but the flow of material can cause variability of strength over the product.  Resin injection uses a male and a female mould. A dry reinforcement pack is placed in between the moulds. Resin is pumped into the mould through an injection point. Resin injection is limited for random reinforcement and low fibre content but can be used for complex shapes.

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 Vacuum-assisted resin injection is and improvement of the above described resin injection technique. Higher fibre content, larger moulds, and a better consolidation are obtained by applying a vacuum in between the male and the female moulds.  Injection moulding uses mixed dough which contains all the ingredients including reinforcements. The dough is pushed into the heated female and cold male mould by a piston. The material properties of the laminate are variable and the sizes are limited.

B.4.5.3 CONTINUOUS PROCESS

The continuous process is used for producing large numbers of identical parts within a limited time span. Several techniques can be used for the continuous process namely:  Continuous laminating; fibre reinforcement and resin are located in between two layers of release films which transport the combined fibre and resin layer to a curing oven. The laminate cures in the oven. The release films are pealed of when the laminate leaves the oven and the laminate is cut in the desired lengths. Flat laminate sheets can be obtained but also corrugated or simple profiles can be obtained by the continuous laminating technique.  Pultrusion; reinforcement is impregnated with resin and pulled through a heated die which forms the fibre and resin mix to its final shape. Mat reinforcement, woven cloth or continuous yarn can be used as reinforcement and high degrees of reinforcement can be obtained. Good longitudinal mechanical properties can be obtained. Mainly beams are produced using this technique.  Continuous filament winding is a method for producing a continuously filament wound pipe. Several winding heads are rotating around the pipe. The profile is cut when the necessary length is obtained.

B.4.6 DESIGN PROPERTIES OF LAMINATES

The stiffness properties and the use of safety factors are briefly discussed in the main report. This section describes the back ground information of FRP and the determination of the design values.

B.4.6.1 STIFFNES PROPERTIES

Nominal stiffness properties of UD lamella, woven lamella, and mat lamella are given in the “CUR aanbeveling 96” (Stichting CUR, 2003). The values for the lamella are given for glass fibre reinforced epoxy FRP without additives. The orthotropic character of the laminate has to be taken into account for design. A coupling between stresses and deformations is present which means that a normal force causes strain and curvature which is presented in Figure B-11. This is caused by the difference in stiffness of the lamellas. The stiffness in the longitudinal direction is larger with respect to the transverse direction. The top layer elongates more with respect to the bottom layer which results in curvature. The coupling behaviour is not present when a symmetric laminate is used.

Figure B-11 Influence of coupling behaviour on asymmetric laminates (Chlosta, 2011)

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The fibre volume fraction (Vf), longitudinal tensile modulus (E1), transverse tensile modulus (E2), shear modulus (G12) and the Poisson’s ratio (v12) influences the strength properties of a lamella. The characteristics of a UD-lamella are given in Table B-14, for a woven lamella in Table B-15, and for a mat lamella in Table B-16.

Table B-14 Nominal values for stiffness characteristics of a UD-lamella (Stichting CUR, 2003)

E E G V 1 2 12 v f [GPa] [GPa] [GPa] 12 40% 30.8 8.9 2.8 0.3 45% 34.3 10 3.1 0.29 50% 37.7 11.3 3.5 0.29 55% 41.1 12.8 3.9 0.28 60% 44.6 14.6 4.5 0.27 65% 48 16.7 5.1 0.27 70% 51.4 19.3 6 0.26

Reduction factor UD-stiffness values (except v12)=0.97

Table B-15 Nominal values for stiffness characteristics of a woven lamella (Stichting CUR, 2003)

E E G V 1 2 12 v f [GPa] [GPa] [GPa] 12 25% 13.4 13.4 2.1 0.21 30% 15.5 15.5 2.3 0.2 35% 17.6 17.6 2.5 0.2 40% 19.8 19.8 2.8 0.19 45% 22.1 22.1 3.1 0.19 50% 24.5 24.5 3.5 1.19 55% 27 27 3.9 0.18

Reduction factor woven-stiffness values (except v12)=0.93

Table B-16 Nominal values for stiffness characteristics of mat lamella (Stichting CUR, 2003)

E E G V 1 2 12 v f [GPa] [GPa] [GPa] 12 10% 6.2 6.2 2.3 0.33 12.5% 6.9 6.9 2.6 0.33 15% 7.6 7.6 2.9 0.33 17.5% 8.3 0.3 3.1 0.33 20% 9.1 9.1 3.4 0.33 22.5% 10.6 10.6 4 0.33 25% 12.2 12.2 4.6 0.33

Reduction factor mat-stiffness values (except v12)=0.91

The maximum tensile and compressive stresses along the main axis and transverse axis are determined using the Hook’s law which is presented in Equation B-2 using the strain and the tensile modulus. The maximum shear stress is determined using the shear modulus [G] and the shear angle γ as presented in Equation B-3.

Equation B-2 Determination of the maximum tensile and compressive stresses

Equation B-3 Determination of the maximum shear stress (Nijhof, 2006)

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The maximum shear angle is not given in the CUR 96 and has to be derived using the strain along the

main axis and transverse axis. The elongations du2/dx2 and du1/dx1 which are presented in Figure B-12 are

equal to the maximum strain along the main axis and transverse axis. The angular rotation ε6 changes into the expression presented in Equation B-4 when the ‘normal axis’ u, v, and w axis are being used.

Figure B-12 The strains ε1 and ε2 and the angle ε6 (Nijhof, 2006)

Equation B-4 The angular rotation by shear (Nijhof, 2006)

The angular rotation is calculated for the maximum strain of 1.2% along the main axis for a dry environment for a 1m x 1m plate is presented in Equation B-5.

Equation B-5 Resulting angular rotation

So, the maximum shear angle is equal to the sum of the maximum strains in the main directions. The maximum strains and the maximum angular rotation are known which are used for calculating the maximum shear stresses of the laminate.

B.4.6.2 SAFETY FACTORS

A structure composed of FRP must comply with Equation B-6 with respect to its strength and load.

Equation B-6 Design check (Stichting CUR, 2003)

In which: S = representative load [N] R = representative strength [N]

= load factor [-]

= material factor [-]

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= conversion factor [-]

Load factors The same load factors as for other hydraulic structures are applicable.

Material and conversion factor The representative stress has to be divided by a material factor and conversion factor for determining the design stress. The following holds for calculating the material factor:

In which:

= 1.35 [-] partial material factor for taking into account the uncertainties of getting the right material properties.

= Dependent of the production methods presented in Table B-17.

Table B-17 partial material factor m,2 depending on the production method (Stichting CUR, 2003)

Partial material factor Production method m,2 Post cured laminate Non-post cured laminate Spray up 1.6 1.9 Hand laminating 1.4 1.7 Vacuum or pressure injection 1.2 1.4 Filament wounding 1.1 1.3 Prepreg 1.1 1.3 Pultrusion 1.1 1.3

The following holds for the conversion factor:

In which:

= 1.1 [-] partial material factor for taking into account the temperature effects.

= 1.0 [-] for dry environment, 1.1 [-] for changing humidity conditions, and 1.3 [-] for prolonged contact with humid environment (surface water, groundwater, etc.).

= In which t = the time span of the load in hours. n = exponent dependent on the fibre reinforcement. 0.01 for UD lamella, 0.04 for woven lamella, and 0.1 for mat lamella.

= 1.1 [-] for fatigue.

Table B-18 gives an overview about the different combinations of conversion factors for six different limit states.

Table B-18 Conversion factors (Stichting CUR, 2003)

Ultimate limit state Serviceability limit state Conversion factor Strength Stability Fatigue Deformations Vibrations First Crack Temperature x x x x x x Moisture x x x x x x Creep x x - x - x Fatigue - x - x x x

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B.4.6.3 MOMENT OF INERTIA OF SANDWICH PANELS

For normal beam and plates models the moment of inertia I is used. For FRP beams the moment of inertia is replaced by the flexural rigidity D taking into account the moment of inertia and the tensile modulus. The flexural rigidity D of the cross section presented in Figure B-13 is depending on the dimensions of the core, the dimensions of the FRP plates, the tensile modulus of the plates, and the tensile modulus of the core. The flexural rigidity of the cross section presented in Figure B-13 is calculated using Equation B-7.

Figure B-13 Dimensions and properties of a sandwich panels (Zenkert, 1995)

∫

Equation B-7 Flexural rigidity D of a sandwich panel (Zenkert, 1995)

In which: D = Flexural rigidity [m4]

Ef = Tensile modulus of the faces [N/m2]

Ec = Tensile modulus of the core {N/m2]

tf = thickness of the faces [m]

tc = thickness of the core [m] d = internal arm [m]

The tensile modulus of a non-constructive core is much lower with respect to the tensile modulus of the faces, and therefore the flexural rigidity of the core can be neglected. Furthermore, for thin faces and a large internal arm the first term of Equation B-7 is much smaller with respect to the mid-term. Therefore, the first term of Equation B-7 can be neglected. Using these statements Equation B-7 reduces to Equation B-8. This expression is used for the design of the sandwich panels.

Equation B-8 Flexural rigidity D of a sandwich panel for a low Ec, small faces and large internal arm (Zenkert, 1995)

B.5 LAMINATE FOR GATE DESIGN

Designing a new laminate for the gates using the classical laminate theory takes too much time within this graduation research. Therefore the characteristics of 0°/90°/+45°/-45° laminates which are already applied in hydraulic structures projects. CUR 96 describes two laminates which are a quasi-isotropic laminate and an anisotropic laminate of which the stiffness characteristics are presented in Table B-19.

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Table B-19 Stiffness properties of a quasi-isotropic and anisotropic laminate (Stichting CUR, 2003)

Stiffness Quasi isotropic Anisotropic laminate Unit properties laminate Vf=50vol% Vf=50vol%

E1 GPa 18.6 25.8 E2 GPa 18.6 15.9 G12 GPa 7 5.6 v12 0.33 0.32

B.5.1 LAMINATE USED FOR THE BENDING MOMENT

The anisotropic laminate is chosen for the plates which are bearing the bending moment. 15% of the fibres are positioned in the four directions which have to bear the multi axial water pressure and the remaining 40% is aligned into the main load bearing direction (0°) as presented in Figure B-14 (SMOZ Project "Kunststof Sluisdeur", sd). The laminate fulfils to the 15% rule (minimal 15% of the fibres must be oriented in the load bearing directions).

Figure B-14 Alignment of the fibres for an anisotropic laminate

Representative stresses of the laminate determined using the 1.2% strain limit and the 0.27% strain limit are presented in Table B-20 (Stichting CUR, 2003). The maximum stresses for SLS and ULS for all the combinations of conversion factors for ULS and SLS are determined using Mathcad of which the sheets are digitally provided.

Table B-20 Representative stress limits of the anisotropic laminate for a strain of 1.2% and 0.27%

Strength Strain limit of Strain limit of properties 1.2% 0,27% [MPa] [MPa]

σ1tR 310 67 σ1cR 310 67 σ2tR 191 43 σ2cR 191 43 τ12R 134 30

B.5.2 LAMINATE USED FOR THE SHEAR FORCE

The isotropic laminate is chosen for the FRP faces which bears the shear force. 25% of the fibres are aligned for the four main directions as presented in Figure B-15. This laminate has a lower tensile modulus for the main load bearing direction with respect to the anisotropic laminate, and a higher tensile modulus for the

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transverse load bearing direction. The stiffness properties of this laminate are already presented in Table B-19. The representative strength for the main load bearing direction, the transverse direction and for shear of the isotropic laminate using the 1.2% strain limit and the 0.27% strain limit are presented in Table B-21. The maximum stresses for SLS and ULS for all the combinations of conversion factors for ULS and SLS are determined using Mathcad which are digitally provided.

Figure B-15 Fibre alignment for an isotropic laminate

Table B-21 Representative stress limit for the isotropic laminate for the 1.2% and 0.27% strain limit

Strength Strain limit of Strain limit of properties 1.2% 0,27% [MPa] [MPa]

σ1tR 223 50 σ1cR 223 50 σ2tR 223 50 σ2cR 223 50 τ12R 168 37.8

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C. Area analyses

C.1 AREA ANALYSIS OF THE RHINE BRANCES (DESIGN LEVEL 1)

The area analysis of the Rhine branches is already covered in section A.3

C.2 AREA ANALYSIS OF THE REACH NEDERRIJN-LEK (DESIGN LEVEL 2)

C.2.1 GENERAL DESCRIPTION OF THE PRESENT SITUATION

The present situation is presented in Figure C-1. The Rhine enters the Netherlands near the village of Lobith and bifurcates at the Pannerdensche Kop into the Waal and Pannerdensch Kanaal. The water levels at Pannerden fluctuate between +7.00m NAP and +15m NAP. The water levels of the Pannerdensch Kanaal are influenced by weir Driel when the discharge at Lobith is lower than 2350 m3/s. The water level at the IJsselkop generated by weir Driel is +8.30m NAP for dammed situations. The aim of the dam regime is diverting 285 m3/s into the IJssel and 25 m3/s into the Nederrijn for preventing a too bad water quality. The Pannerdensch Kanaal is 7 kilometres long and the floodplains, which are mainly used by agriculture, have a surface of 250 hectares. The Pannerdensch Kanaal changes name into Nederrijn at pumping station Kandia (encircled in Figure C-1) which is built in the river bed of the Oude Rijn near the village of Zevenaar. The Nederrijn changes name into the Kromme Rijn at the village of Wijk bij Duurstede, and turns towards the north towards the city of Utrecht. The water from the Nederrijn flows into the Lek and merges to the river Noord after 62 kilometres. The combined Lek and Noord flows as the Nieuwe Maas towards sea. Two other weirs located near the village of Amerongen and Hagestein regulates the water levels of the Nederrijn and a section of the Lek. The normal water level created by weir Amerongen is +6.00m NAP which is maintained by a cylinder valve, waterpower station, and two visor gates. The normal water level regulated by weir Amerongen is +3.00m NAP and is maintained by a cylinder valve and two visor gates. A damped tide is present at the downstream side of weir Amerongen, so the water head over the weir is variable. The water management relation of the Nederrijn with the surrounding areas is poor; the main function of the Nederrijn is the transport of water and recreation. Seven pumping stations and inlets are located along the Nederrijn. The majority of the people living along the Nederrijn live in the Arnhem, Oosterbeek, Renkum, Wageningen, and Rhenen. The Amsterdam-Rijnkanaal, Lekkanaal, and Merwedekanaal are connected to the Lek. The Lek is important for the water management function and the water supply of the Amsterdam-Rijnkanaal. People living along the Lek are concentrated in the villages of Culemborg, Nieuwegein, Vianen, and Schoonhoven. The IJssel branches off from the Nederrijn at the bifurcation IJsselkop and flows into the IJsselmeer. The water management function of the IJssel is large. The IJssel-valley (Dutch: IJsseldal) is connected to the higher grounds by groundwater. The water of the IJssel is stored in the IJsselmeer for periods of drought; the fresh water supply of the northern part of the Netherlands depends on this storage. Also the water

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supply of the ‘Twente kanalen’ from the IJssel is important. So the IJssel influences the water management of a large part of the Netherlands (Rijkswaterstaat, 1994).

C.2.2 DETAILS OF THE PRESENT SITUATION

The details of the Nederrijn and Lek are subdivided in several topics like width, navigation, bridges, etc. Every topic is described in a distinct paragraph.

C.2.2.1 WATERWAY DIMENSIONS

The dimensions of the Pannerdensch Kanaal, Nederrijn, and Lek are presented in Table C-1. The Nederrijn is split up into four sections with different normalised widths. This width of a certain reach presented in Table C-1 representing the average width of the summer bed measured between the groins.

Table C-1 Waterway dimensions (Rijkswaterstaat, 1994)

Normalised width Length of reach Reach [m] [km] Pannerdensch Kanaal 140 7 Pumping station Kandia - IJsselkop 140 4,5 IJsselkop - Arnhem 100 7 Arnhem - Wageningen 115 18 Wageningen - Wijk bij Duurstede 130 23 Schoonhoven 190 43 End of the Lek 100 20

C.2.2.2 RIVER BED LEVEL

The river bed level is not a fixed level because dunes and ripples are propagating in the downstream direction of the river. A minimum and maximum level measured at specific locations is presented in Table C-2.

Table C-2 Minimum and maximum depths measured at different locations (Rijkswaterstaat, 2011)

Minimum and maximum bottom depth Distance from Location compared to the reference level (+NAP) start point [m] [km] Start point of the +3,00 +4,70 0 Pannerdensch Kanaal Weir Driel -0,70 +2,80 24,00 Weir Amerongen -3,70 -0,70 54,65 Weir Hagestein -5,00 -3,70 76,84 Schoonhoven -8,50 -5 102,40 Nieuwe Maas -13.40 -5.20 120,15

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Figure C-1 Overview of the Nederrijn and the upper section of the Lek. The existing weirs and bridges are indicated by a green and red crossing. Furthermore connected waterways are presented (Google maps)

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C.2.2.3 WATER LEVELS OF THE NEDERRIJN AND LEK

Water levels are even more variable with respect to the bed levels. Water levels are determined for distinct discharges represented in ‘reference lines’ (Dutch: ‘betrekkingslijn 1991.0’). A reference line is a graphical representation the water levels along a river. The discharge and water level for a frequency of exceedance, average winter water level, average summer water level, and the lowest allowable water level are presented for several locations along the Pannerdensch Kanaal, Nederrijn, and Lek in Table C-3. The OLR (Dutch: Overeengekomen Lage Rivierafvoer) is based on the discharge which is undershot for 20 ice-free days a year (Rijkswaterstaat, 2011). Two water levels with a frequency of exceedance of 1/1250 are presented. The first row represents the ‘1/1250 new design discharge;’ these water levels are determined for the new governing discharge of 16.000 m3/s at Lobith (Rijkswaterstaat, 2007). The second row is the ‘1/1250 old design discharge’ which represents the old governing discharge at Lobith (15.000 m3/s).The water levels at the weirs represents the expected water level at the upstream side of a weir (Amerongen boven). Water levels of ‘Hagestein beneden’ (downstream side of weir Hagestein) are influenced by the tide and could not be combined in Table C-3. Therefore, values of ‘Hagestein beneden’ are presented in Table C-4. The water levels of ‘Hagestein beneden’ are determined for a given discharge and a given tide. Water levels available at the lower end of the Lek are calculated for Streefkerk and presented in Table C-5.

Table C-3 Water levels, frequency of exceedance and discharge (combined tables from: (Rijkswaterstaat, 2012))

Water level +NAP Discharge Frequency of Pannerdensche Driel Amerongen Hagestein at Lobith IJsselkop exceedance Kop boven boven boven [m3/s] [m] Highest known situation 12600 16,00 14,20 12,25 8,40 5,90 1/1250 new design discharge 16000 16.70 14.80 12.60 8.80 6.5 1/1250 old design discharge 15000 16,65 14,55 12,50 8,65 6,30 1/100 12320 15,90 14,10 12,20 8,35 5,80 1/10 9670 15,15 13,45 11,60 7,75 5,00 1/2 6800 13,90 12,40 10,60 6,65 3,70 1/1 5800 13,30 11,90 10,10 6,10 3,20 average winter 2200 9,50 8,65 average summer 1985 9,10 8,40 OLR 984 7,50 7,30

Table C-4 Water levels at Hagestein beneden (Rijkswaterstaat, 2012)

Water level +NAP Discharge Average tide Spring tide Neap tide at Lobith High water Low water High water Low water High water Low water [m3/s] [cm] 700 89 -48 101 -44 89 -44 OLR = 984 94 -44 107 -41 94 -40 1.400 99 -41 112 -37 99 -35 2.200 162 63 172 65 154 67 3.500 204 148 213 150 197 151 5.000 272 242 278 243 269 244 6.800 360 344 364 345 359 345

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Table C-5 Water levels at Streefkerk (Rijkswaterstaat, 2012)

Water level +NAP Average tide Spring tide Neap tide Discharge High at Lobith High water Low water Low water High water Low water water [m3/s] [cm] 700 77 -39 90 -36 76 -33 OLR = 984 81 -36 93 -33 79 -30 1400 85 -34 97 -30 83 -27 2200 108 -9 119 -4 106 -3 3500 121 5 133 9 113 11 5000 134 28 146 32 126 34 6800 154 62 166 65 146 67 10000 202 135 213 137 195 137

C.2.2.4 DISCHARGES

The discharge of the Nederrijn is related to the water levels at Lobith and the IJsselkop. The discharge at the Nederrijn for a fully dammed situation is 25 m3/s and a aimed discharge of 285 m3/s for the IJssel which is regulated by weir Driel. The corresponding water level of ‘Driel boven’ is in this situation +8.30m NAP. A lower limit of the IJssel discharge is 250 m3/s; the corresponding draught is 2.70m. The relation between the water levels and discharge at the IJsselkop is presented in Figure C-2 and Figure C-3.

Figure C-2 Discharge at Lobith, Nederrijn and IJssel, and the water level at the IJsselkop. Based on: (ARCADIS, 2010)& (Lemans, 2007)

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Figure C-3 An enlargement of the low discharge section of the discharge at Lobith, Nederrijn, and IJsselkop and the water level at the IJsselkop

Weir Driel is gradually opened to maintain a water level of +8.30m NAP at the IJsselkop which results in a higher discharge of the Nederrijn. The gates of weir Driel are fully opened for a discharge at Lobith of 2350 m3/s. For this discharge sufficient depth is obtained for the IJssel and for the upper reach of the Nederrijn. The gates of weir Amerongen and weir Hagestein are fully opened for a discharge of 3000m3/s. Sufficient depth is obtained for this discharge at the lower reach of the Nederrijn and Lek. The discharge of the IJssel is not fully known for low discharges at Lobith. An indication has been made based on the water distribution for dry periods of which the figure is included in the main report. The IJssel discharge is 215 m3/s when the discharge at Lobith is 1200m3/s. A water level of +8.30m NAP at the IJsselkop corresponds to a discharge 285 m3/s according to the master thesis Afvoerverdeling Rijntakken (Lemans, 2007) which is included in Figure C-3. The discharges of the IJssel for discharges higher than 2350m3/s for which the gates of weir Driel are fully opened are based on the water distribution percentages which is also included in the main report.

C.2.2.5 HEIGHT RESTRICTIONS

Six bridges are crossing the Nederrijn and three bridges the Lek. The locations of the bridges are indicated in Figure C-1. Besides these bridges, power lines are crossing the Nederrijn and Lek. The height of the power lines are not taken into account because they are with respect to the height of the bridges. Names, locations, type of bridge, the width, and the height of the bridges are shown in Table C-6.

Table C-6 Bridges over the Nederrijn and Lek (Rijkswaterstaat, 2011)

Height width Bridge Location (rail)way Details +NAP [m] [m] Andrej Sachorovbrug Arnhem N325; 80 km/h 1 fixed opening 115,65 25,28 John D. Frostbrug Arnhem local road; 50 km/h 1 fixed opening 113 22,85 Nelson Mandelabrug Arnhem N225; 80 km/h 1 fixed opening 200 24,58 Railway bridge Oosterbeek Railway 1 fixed opening 122 21,16 Bridge A50 Heteren A50; 120 km/h 1 fixed opening 116 22,51 Bridge Rhenen Rhenen N233; 80 km/h 1 fixed opening 113 20,74 Railway bridge Culemborg Railway 1 fixed opening 140 16,2 Hagesteinse brug Hagestein A27; 100 km/h 1 fixed opening 100 15,8 Lekbrug Vianen To be demolished 1 fixed opening 155 15,54 Jan Blankenbrug Vianen A2; 100 km/h 1 fixed opening 147 17,28

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C.2.2.6 PUMPING STATIONS AND INLETS

Several pumping stations and water inlets are situated along the Pannerdensche Kanaal, Nederrijn, and Lek. The characteristics of each pumping station and inlet are described in Table C-7. The locations of the structures are indicated in the map presented in Figure C-5. The information found for the inlet and pumping stations is based on: (Stowa, 2009), (Waterschap Rivierenland, 2006), (Mazijk, 2005) & (Hoogheemraadschap de Stichtse Rijnlanden, 2012).

Table C-7 Pumping stations and inlets along the Nederrijn and Lek

Location Structure Name Discharge Basin on map Average 4m3/s. Zero discharge when Voedingskanaal Inlet A the water level of the Pannerdensch Betuwe Linge Kanaal has decreased too much Pumping 10 m3/s to 17m3/s depending on the Kandia B Oude Rijn station head. (maximum head is 4 metres) Major part of Arnhem Pumping Zuid and the water of Drielsedijk C 6.2m3/s after renovation station wastewater treatment plant in Zuid Arnhem Pumping Inlet capacity of 4.7m3/s. Discharge station Mr. G.J.H. Kuijk D Linge capacity not found. and inlet Grebbe, Grift, flushing Inlet Grebbesluis E 2.5m3/s of the channels of Amersfoort Drain Tollewaard F To be determined the polder at Tollewaard Historic water defence Inlet Kromme Rijn G 7m3/s line Pumping Koekoek H 11.2m3/s Lopikerwaard station Haagse Inlet I To be determined water treatment plant waterleiding Pumping not on Elshout To be determined waterways in the polder station map

C.2.2.7 CHANNELS

Three major channels are connected to the Nederrijn and the Lek which are the Merwede Kanaal, Amsterdam-Rijnkanaal, and the Lekkanaal. They are indicated with brown (Merwede Kanaal), pale blue (Amsterdam-Rijnkanaal), and purple (Lekkanaal) in Figure C-1 and Figure C-5. The purpose and properties of each channel are described in this paragraph.  The Merwede Kanaal was excavated to improve navigation form Germany to Amsterdam and vice versa in 1880-1892. The channel is divided in two parts, namely a reach at the South of the Lek and a reach on the North side of the Lek. The ‘normal water level’ (Dutch: streefpeil) of the northern part is +0.45m NAP and fluctuates between +0.40m NAP and +0.70m NAP. The northern boundary of this section of the Merwede Kanaal is the ‘Muntsluis’ and the southern boundary is the Koninginnensluis. The ‘Koninginnensluis’ connects the northern reach of the Merwede Kanaal with the Lek. The ‘Grote sluis’ located in Vianen is the northern boundary of the southern reach of the Merwede Kanaal and the ‘Grote Merwedesluis’ located in Gorinchem is the southern boundary. The normal water

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level of the southern reach of the Merwede Kanaal is +0.80m NAP and fluctuates between +0.60m NAP and +1.20m NAP.  The Amsterdam-Rijnkanaal was excavated from the Waal to the city of Amsterdam. Another channel (Noordzeekanaal) was excavated from Amsterdam to IJmuiden for the navigation from Amsterdam towards sea. This channel was an ‘improvement’ for the navigation from Germany to Amsterdam because the Merwede Kanaal became too narrow and shallow for new vessel classes. The reach between the Waal and the Nederrijn is called the ‘Betuwepand.’ Vessels have to cross the prins Bernardsluis which is located at the Waal and the prinses Marijkesluis which is located at the southern side of the Nederrijn to sail from the Waal to the Nederrijn. The reach of Amerongen and Hagestein is in open connection with the ‘Betuwepand’ when the weir gates are closed. The water level of this reach (+3.00m NAP) is the same as the water level of the ‘Betuwepand.’ The barrier of Ravenswaaij which is located next to the prinses Marijkesluis is lifted for this situation. The barrier is closed when the water level of the Lek is equal to or higher than +5.55m NAP; the locks have to be used by the vessels to sail into the ‘Betuwepand.’ The Bernardsluis is used for controlling the water levels of the Betuwepand during normal and high water levels at the Waal. A minimum head difference of 0.3m must be present to prevent chatter of the lock doors. The dam-regime of Amerongen and Hagestein is used for realising this water level difference when the prinses Irenesluis and barrier Ravenswaaij are open. It is possible to maintain this till a water level of +3.15m NAP of the Waal at Tiel. The prins Bernardsluis will be opened if the water level of the Waal at Tiel further decreases. The Waal and the Nederrijn are in open connection for this situation. The prinses Irenesluis is located at the northern side of the Lek and is the entrance of the Amsterdam-Rijnkanaal. Vessels have to cross this lock to sail from the Nederrijn towards Amsterdam and vice versa. The normal water level maintained by this lock is - 0.40m NAP. The Amsterdam-Rijnkanaal has also a water management function. The water boards located along the Amsterdam-Rijnkanaal are extracting water in periods of droughts and are using the Amsterdam-Rijnkanaal as a drain in wet periods. The green and orange area marked in Figure C-4 represents these area which are using the Amsterdam Rijnkanaal as drain.

Figure C-4 Water management function of the northern part of the Amsterdam-Rijnkanaal (Rijkswaterstaat, 2011)

Also the ‘Betuwepand’ is used by the surrounding water boards as a drain during wet periods and as a water source during dry periods (Rijkswaterstaat, 1998). The sluices at Tiel are used as a water inlet for the Betuwepand, the Nederrijn, and the Amsterdam-Rijnkanaal in dry periods. The water of the Waal is used for the regulation of water levels and as a source of fresh water for

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the surrounding polders (Rijkswaterstaat, 2011). Dimensions of the locks of the Amsterdam- Rijnkanaal are presented in Table C-8.  The Lekkanaal connects the northern section of the Amsterdam-Rijn with the Lek. It was excavated in as an improvement of the Merwedekanaal. The prinses Beatrixsluis is the southern boundary of the Lekkanaal. This lock and the prinses Irenesluis form the southern boundary of the Lekkanaal. No locks or sluices are located at the northern boundary of the Lekkanaal; it is in open connection with the Amsterdam-Rijnkanaal so the water levels in the Lekkanaal are equal to the water levels in the Amsterdam-Rijnkanaal. The dimensions of the prinses Irenesluis are presented in Table C-8. The normal water levels of each channel are presented in Table C-9

Table C-8 Lock properties of the Amsterdam-Rijnkanaal, Lekkanaal, and Merwedekanaal (Steenbergen, 2012)

Min Sill height at the Nederrijn Length Height +NAP Structure width and Lek side +NAP [m] [m] [m] [m] Prinses Marijkesluis 260 18 -2,35 - Barrier Ravenswaaij - 80 -2,35 14,65 Prinses Irenesluis 260 24 -4,6 16,75 Prinses Beatrixsluis 230 18 -4,5 14,40 Koninginnensluis 220 11,85 -3,3 - Grote Sluis 240 11,85 -3,5 -

Table C-9 Normal water levels

Normal water level +NAP Channel [m] Merwede Kanaal; southern section +0,80 Merwede Kanaal; northern section +0,45 Amsterdam-Rijnkanaal; Betuwepand +3,00 till +5,55 Amsterdam-Rijnkanaal; northern section -0,40 Lekkanaal -0,40

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Figure C-5 Overview of inlets and pumping stations (Google Maps)

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C.2.2.8 NAVIGATION

Although locks are not designed in this research information has to be available for navigational classes and the intensity of the navigation on the waterways. Every waterway has a certain classification which indicates the size of the allowable vessels. The CEMT classification of the Nederrijn, Lek, Merwede Kanaal, Lekkanaal, and Amsterdam-Rijnkanaal are shown in Table C-10.

Table C-10 CEMT classifications (Ministerie van V&W, 2007) & (Rijkswaterstaat, 2011)

Width Length Draught (loaded) Water way CEMT class [m] [m] [m] Lek Vla 22.8 95-145 2,5-4,5 Nederrijn Va 11.4 95-135 2,5-4,5 IJssel Va 11.4 95-135 2,5-4,5 Merwedekanaal; southern section IV 9,5 80-105 2,5-2,8 Merwedekanaal; northern section Va 11,4 95-135 2,5-4,5 Amsterdam-Rijnkanaal; Betuwepand VIb 22,8 185-195 2,5-4,5 Amsterdam-Rijnkanaal; northern section VIb 22,8 185-195 2,5-4,5 Lekkanaal Vb 11,4 170-190 2,5-4,5

The traffic intensity of the rivers and channels in the Netherlands is determined by counting by the lock operators. These numbers are published in the ‘Kerntallen Scheepvaart.’ A selection of this data has been made for the Nederrijn, Lek, Lekkanaal, and the Amsterdam-Rijnkanaal. The intensities are shown in Figure C-6.

Figure C-6 Vessel passages of the Locks of the Nederrijn (NR), Amsterdam-Rijnkanaal (A-R), and Lekkanaal (LK)

From this figure, it can be concluded that the amount of vessels which change waterways is limited; the amount of vessels which are using the locks of Amerongen, Driel, and Hagestein are about the same. The lock of Amerongen is the most used locks of the Nederrijn. The amount of passages of the prinses Irenesluis and the prins Bernardsluis is also nearly the same. So most of the ships are just crossing the Nederrijn and do not change between waterways. The most intensive used lock is the lock of the Lekkanaal. Vessels from the Lek are able to sail from the port of Rotterdam directly towards the

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Amsterdam-Rijnkanaal and vice versa by using the prinses Beatrixsluis and saving time by avoiding the lock of Hagestein and prinses Irenesluis. A distinction is made between commercial and recreational passages. These passages are estimated and published in ‘Kerntallen Scheepvaart.’ The total amount of passages, cargo passages, and recreational passages are shown in Table C-11. From this table and ‘Scheepvaartinformatie Hoofdvaarwegen 2009’ can be concluded that about the half of the navigation of the Nederrijn is recreational. (7635 commercial vessels and 3747 recreational boats sailed towards the east at weir Driel. And 3787 commercial vessels and 1921 recreational boats sailed towards the west at weir Driel (Rijkswaterstaat, 2009))

Table C-11 Passages of vessels in the Nederrijn, Amsterdam-Rijnkanaal, and Lekkanaal (Noordhoff Atlasproducties, 2011)

Total amount of 'Cargo passages' 'Recreational' passages Waterway passages [x1.000] [x1.000] (2005) [x1.000] Driel-Amerongen ≤ 20 ≤ 20 3,5-10 Amerongen-Hagestein 20-40 ≤ 20 3,5-10 Hagestein-Lek 60-80 40-60 3,5-10 Amsterdam-Rijnkanaal 60-80 60-80 ≤ 3,5 (Northern part) Amsterdam-Rijnkanaal (In between the Nederrijn and 20-40 20-40 ≤ 3,5 the connection of the Lekkanaal) Amsterdam-Rijnkanaal 40-60 40-60 ≤ 3,5 (Southern part) Lekkanaal 40-60 20-40 3,5-10

The trends in navigation are determined using data from 2005 to 2008. A distinction has been made in the total development, the development of commercial shipping, and the development of recreational boating of which the results are shown Table C-12.

Table C-12 Trends in navigation from 2005 to 2008

Total Commercial Recreational Development Development Development navigation shipping boating total commercial recreational Lock Waterway (2008) (2008) (2008) navigation shipping boating [amount [amount [amount [%] [%] [%] /year] /year] /year] Driel Nederrijn 17100 10614 5659 2,4 10,7 -10,7 Hagestein Lek 16346 8297 7504 -11,8 -6,9 -17,5 Prins Amsterdam- Bernardsluis Rijnkanaal 39586 35576 2531 37,9 38,8 8,3 Prinses Lekkanaal Beatrixsluis 54069 46528 4650 -5,3 -2,2 -31,2 Prinses Amsterdam- Irenesluis Rijnkanaal 39718 35894 2077 44,4 44,6 21,3

Also a distinction is made with respect to the upstream and downstream navigation. Twice as much navigation sails to the west with respect to the navigation sailing towards the east for the Nederrijn. This difference is explained by the flow velocities of the Nederrijn-Lek and Waal. The velocity of the Nederrijn is nearly zero when the weirs of Amerongen, Hagestein, and Driel are closed. A counter flow has to overcome when sailing upstream on the Waal, which costs more fuel than sailing upstream on the Nederrijn-Lek with no counter flow. The drawback of the Nederrijn-Lek route is the time delay caused by the locks which have to be used. Vessels without a very tight time constraint could use the Nederrijn-Lek

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route to safe fuel. The amount of vessels in the upstream and downstream direction of the Nederrijn is shown in Table C-13. Also the amount of vessels sailing at the Waal and the IJssel are presented as a reference. The amount of vessels sailing on the Nederrijn is in the same order as the vessels sailing on the Gelderse IJssel. Also a difference in upstream sailing vessels and downstream sailing vessels is present at the IJssel. More ships are sailing with the flow direction towards the north. About 40% of the ships would not sail back on the IJssel and are taking another route. Perhaps they are avoiding the counter flow by sailing back on the IJsselmeer and Markermeer. The difference between upstream sailing vessels and downstream sailing vessels on the Waal is less compared to the Nederrijn and IJssel because ships are not able to take another route for bypassing the counter flow.

Table C-13 Navigation of waterways in different directions

2005 2006 2007 2008 Development Waterway Direction [vessels [vessels [vessels [vessels [%] /year] /year] /year] /year] East 25965 36886 36076 29608 14 Nederrijn West 13961 18516 18288 15158 11,2 East 65613 65840 62955 60723 -7,5 Waal (Lent) West 78121 70801 63638 62024 -20,6 East 71611 67475 63976 62260 -13,1 Waal (Lobith) West 97816 79888 70618 68558 -29,9 North 24562 28092 27346 25549 4 Gelderse IJssel South 16361 18531 17565 16707 2,1

C.2.2.9 LEVEES

Levee heights are measured using a digital height elevation map using the Actueel Hoogtebestand Nederland (Actueel Hoogtebestand Nederland, 2012).

C.2.2.10 HARBOURS

The harbours located along the Nederrijn and Lek are divided in three groups which are: recreational, commercial, and military harbours. The harbours are sorted from east to west and summed up below. Recreational harbours are located near or in:  Arnhem  Wageningen (water sport centre VADA)  a cut off bend near Wijk bij Duurstede (water sport centre De Lunenburg)  Beusichem  Culemborg (jachthaven De Helling)  Nieuwegein (water sport centre ZV de Lek)  Vianen (wsv de Peiler)  Bolsweerd  Nieuwpoort  Schoonhoven  Liesveld. Commercial harbours of cities and factories are located near:  Arnhem (a harbour basin and a quay wall along the river)  Heteren (paper factory Parenco)  Rijnhaven Wageningen (a harbour basin)  Lienden; storage area for fertilizer (a harbour basin)  Lienden; Skinspark BV (harbour located between groins)

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 Wijk bij Duurstede (small harbour basin)  Schoonhoven; stone factory (quay wall along the river)  Groot Ammers; Hartog olie BV (quay wall along the river)  Bergstoep; animal food factory (small harbour basin)  Nieuw-Lekkerland; Sand transhipment JW de Lange BV (quay wall along the river)  Nieuw-Lekkerland; Den Boer Beton (small harbour basin) Military harbours are located near:  Arnhem  fort ‘t Spoel  fort t Goilberdingen  fort Honswijk.

The size of the recreational harbours and their location are presented in Figure C-7. Most of the recreational harbours are located in Utrecht and two large harbours are located in Gelderland.

Figure C-7 Recreational harbours in the middle section of the Netherlands (Noordhoff Atlasproducties, 2011)

The commercial harbours are presented in Figure C-8. More transhipment locations are shown in this figure with respect to the list of the commercial harbours. The harbours of the enterprises shown in the enumeration are the large harbours. The other dots in Figure C-8 represents all transhipment points; also some very small scale transhipment points with limited cargo transport (Figure C-9 left) and harbours which are not used any more (Figure C-9 right). The harbours presented in the enumeration are mapped by Rijkswaterstaat in December 2011. So this is the most actual and accurate database of transhipment points along the Nederrijn and are used for this reason in this design study.

Figure C-8 Transhipment locations along the Nederrijn (Noordhoff Atlasproducties, 2011)

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Figure C-9 Small scale transhipment point at Rheden (left) and an harbour near Wageningen which is out of use (right) (bing.nl/maps)

C.2.2.11 RECREATION

Other kinds of recreation are present along the Nederrijn and Lek besides the water sport centres described in the previous paragraph. The other kinds of recreations are:  Several (water) scouting clubs. They are using the Nederrijn for sailing and other activities.  People who are using the sand deposits between the groins as a beach in summertime. Furthermore some small side lakes are created along the Nederrijn which serve as a recreational lake. As example: a bend which is cut off for the construction of weir Amerongen is changed into a nature area with a recreational beach.  Campsites are located in the floodplains along the Nederrijn. People are able to stay here due to the low and regulated water levels in summer.  Natural areas like the cut off bend near Amerongen have been created. People are enjoying the nature in these natural areas by visiting these areas.

C.2.2.12 NATURE

Nature restoration and preservation projects are executed along the Nederrijn and Lek. The floodplains of the Nederrijn and Lek are part of the ‘ecologische hoofdstructuur.’ The floodplains of the 40 km long reach Arnhem-Amerongen are changed into an area with swamps, grasslands, small lakes etc. Furthermore an ecological connection of the floodplains of the Nederrijn with the higher grounds (Veluwe and Utrechtse Heuvelrug) is created. The floodplains of the Nederrijn are part of the ‘ecologische hoofdstructuur.’ Also the floodplains of the Lek are part of the ‘ecologische hoofdstructuur.’ A natural area is created between Everdingen and Hagestein by excavating gullies, swamps, and creating grasslands. The clay which has been excavated is used for levee reinforcement. Furthermore, two sections of the floodplains are part of Natura 2000, namely the floodplains of Ameide and the floodplains of the reach Wijk bij Duurstede-Renkum. Natura 2000 is an instrument for protecting the biodiversity in EU for protecting valuable species and habitats. The areas which are part of the Natura 2000 are shown in Figure C-10 and Figure C-11 (Ministerie van Economische Zaken, Landbouw en Innovatie, 2012). It is not permitted to build a structure in these areas unless there are compelling arguments and sufficient nature compensation works are introduced. Natura2000 areas have to comply with the bird and habitat directives which are briefly explained in the following enumeration:  Solutions implemented in the floodplains must comply with the Birds directive: o The conservation and regulation of all naturally occurring wild birds in the European territory of the member states, including their eggs, nests and habitats. o Regulating the exploitation of these species.

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 Solutions must comply with the habitat directive: o Applying a decent management of the landscape elements which are essential for the migration, dispersal and genetic exchange of wild species. o Setting up a system of strict protection of certain endangered species and determine whether reintroduction of certain species in their territory is desirable. o Forbidding the use of non-selective approaches for collecting, capturing or killing certain animals and plant species.

Figure C-10 Natura 2000 area from Renkum to Wijk bij Duurstede (Ministerie van Economische Zaken, Landbouw en Innovatie, 2012)

Figure C-11 Natura 2000 area near Ameide (Ministerie van Economische Zaken, Landbouw en Innovatie, 2012)

C.2.2.13 AGRICULTURE

Agricultural areas are present along the Nederrijn and Lek. A sufficient groundwater level is important for the production of the farmlands. The groundwater level is established by the waterways between the farmlands and the rivers. Fresh water is let into the polder for maintaining the (ground) water levels in periods of drought. This water is extracted by the described inlets from the Nederrijn.

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C.3 AREA ANALYSIS OF THE REACH WIJK BIJ DUURSTEDE TILL WEIR HAGESTEIN (DESIGN HPASE 3)

The new weir complex has to be built near the village of Culemborg. In this paragraph the area spanning from the crossing of the Amsterdam-Rijnkanaal till weir Hagestein is analysed and uses as input of the requirements analysis.

C.3.1 WATERWAY DIMENSIONS

The waterway dimensions are presented in Table C-14. The normalised width (the width in between the groins) is measured for every 2 kilometres and starting at the crossing of Amsterdam-Rijnkanaal. The normalised width varies between the 130 and 140 metres.

Table C-14 Waterway dimensions in between the Amsterdam-Rijnkanaal and weir Hagestein

Normalised width Length of reach Location [m] [km] Start of the Lek 130 0 2 km from the starting point 130 2 4 km from the starting point 145 4 6 km from the starting point 135 6 8 km from the starting point 150 8 10 km from the starting point; railway 140 10 bridge Culemborg 12 km from the starting point 145 12 14 km from the starting point 145 14 17 km from starting point; 140 16 Weir Hagestein

C.3.2 WATER LEVELS

The upstream dammed water level of weir Culemborg is +5.00m NAP for a fully dammed operation which is presented in Figure C-12. The average water levels according to the reference lines (Dutch: ‘betrekkingslijn 1991.0’) are presented in Table C-15. The downstream water level is generated by a combination of tide and runoff. The water levels measured and predicted by Rijkswaterstaat for the project area are presented in Table C-4.

Table C-15 Predicted water levels at Culemborg (Rijkswaterstaat, 2012)

Discharge Water level +NAP Frequency of exceedance at Lobith Culemborg (Lek railway bridge) [m3/s] [m] Highest known 12600 6,60 1/1250 15000 7,00 1/100 12320 6,50 1/10 9670 5,70 1/2 6800 4,55 1/1 5800 4,00 average winter 2200 average summer 1985 OLR 984

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C.3.3 BED LEVEL

The maximum bed levels which are presented in Figure C-12 has to be maintained for the implementation of the weir. No river bed adaptations have to be implemented when ‘weir complex Culemborg’ is places at 55 kilometres from the IJsselkop as indicated in Figure C-12 to generate sufficient draught for recreational boating.

Figure C-12 Longitudinal profile of the chosen configuration variant of design level 2

C.3.4 DISCHARGES

The same dam regime described for design level 2 holds for the weir design. A minimum flushing discharge of 25 m3/s must be present and the weir is fully opened when a discharge of 3000 m3/s is present at Lobith. The discharge of the Nederrijn and Lek which has to generate sufficient draught is circa 550 m3/s (Rijkswaterstaat, 1994). This discharge is higher with respect to the minimum discharge of 412 m3/s for which the gates of weir Driel are fully lifted which is caused by the divergence in width; the river is wider at the downstream side so a higher discharge needs to be present to maintain sufficient depth for navigation.

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C.3.5 HEIGHT RESTRICTIONS

One height restriction is located in the area of interest which is the railway bridge of Culemborg. The span of the bridge is 140 metres and the bottom level of the bridge is located at +16,20m NAP (Rijkswaterstaat, 2011). The northern ramp of the bridge consist is composed of many smaller spans as indicated in Figure C-13.

Figure C-13 Railway bridge Culemborg (Google Maps)

C.3.6 PUMPING STATIONS AND INLETS

No pumping stations and inlets are located at in between Wijk bij Duurstede and weir Hagestein. (Rijkswaterstaat, 2011).

C.3.7 CHANNELS

One channel is of importance for the design of weir complex Culemborg which is the Amsterdam- Rijnkanaal. The weir has to control the water levels at the Amsterdam-Rijnkanaal connection. The water levels at the crossing must remain lower than +5.55m NAP in order to maintain an open connection of the Nederrijn with the Betuwepand.

C.3.8 HARBOURS

Two professional harbours are located in between weir Hagestein and the Amsterdam-Rijnkanaal, namely:  A small ‘harbour’ located in the village of Culemborg. o This quay wall offers three mooring places along the bank. People can acces their vessels by walkways as presented in Figure C-14.  A quay wall in between the groins near the village of Hagestein owned by a wholesale company called: Lekzicht b.v. o The quay wall is used for unloading and loading vessels with sand and gravel. The harbour is presented in Figure C-14.

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Figure C-14 'Harbour' Culemborg (left) and harbour 'Lekzicht' (right) (Bing Maps)

C.3.9 RECREATION

Two recreational harbours are located in between weir Hagestein and the Amsterdam-Rijnkanaal namely:  Jachthaven Beusichem o Jachthaven Beusichem has capacity of 100 boats. It is located near a campsite for 100 places as presented in Figure C-15.  Jachthaven De Helling located in Culemborg. o Jachthaven the Helling has a capacity of 150 boats and is located near a small campsite. The harbour is owned by the municipality of Culemborg. The harbour is presented in Figure C-15.

Figure C-15 Recreational harbour Beusichem (left) and recreational harbour de Helling (right) (Google Maps)

C.3.10 NATURE

Natura2000 areas are not located in between weir Hagestein and the Amsterdam-Rijnkanaal, so solutions does not have to comply with the bird and habitat directives. However, natural areas have been created in the floodplains. By-passes of the river have been excavated and the agricultural areas are changed into meadows. Oxes and many bird species are living in these natural areas.

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C.3.11 AGRICULTURE

The part of the floodplains which are not changed into natural areas is cultivated by farmers. Many grasslands for cattle are present at the floodplains in between weir Hagestein and the Amsterdam- Rijnkanaal.

C.4 AREA ANALYSIS OF THE WEIR LOCATION (DESIGN LEVEL 4)

The weir complex is going to be built at the floodplains upstream from the village of Culemborg. The project area of the weir is described in this section.

C.4.1 WEIR LOCATION

The weir is located in between the village of Culemborg and Wijk bij Duurstede. An impression of the floodplains is given in Figure C-16. The photos are made from the winter levee directed south towards the Lek.

Figure C-16 Present floodplains at the new weir location

C.4.2 GLOBAL WEIR DIMENSIONS

The total width of the openings should be larger than 100 metres and smaller than 140 metres which is determined in appendix O. The sill level of the weir is -2.2m NAP.

C.4.3 ROAD CONNECTIONS

The new weir complex is located at a kind of island as indicated in Figure C-17. 6 bridges are available for reaching the ‘island’ of which 2 are spanning over the Lekkanaal and 4 over the Amsterdam Rijnkanaal. The characteristics of each bridge are presented in Figure C-17 and Table C-16. The bridges presented in Figure C-17 are numbered and the description of each numbered bridge is given in Table C-16.

Table C-16 Description of the available bridges width Bridgenumber Channel Road classification [lanes] 1 Lekkanaal 2 lanes for both directions Local road (50km/h) 2 Lekkanaal 1 lane for both directions Local road (50km/h) 2 lanes for one direction. Divided 3 Amsterdam Rijnkanaal Highway (120km/h) by a mid-berm 4 Amsterdam Rijnkanaal 2 lanes for both directions Local road (80km/h) 5 Amsterdam Rijnkanaal 2 lanes for both directions Local road (80km/h) 6 Amsterdam Rijnkanaal 2 lanes for both directions Local road (80km/h)

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Figure C-17 Road connections (Source: Google maps)

The two main roads in the project area are the Provincialeweg, the Beusichemseweg and the Lekdijk. These roads are located on top of a levee. An impression of the roads is given in Figure C-18

Figure C-18 Roads heading to the project area (left: Provincialeweg; middle: Lekdijk; right: Beusichemseweg

C.4.4 WATERWAY CONNECTIONS

The weir is located in the Lek, connected to the Nederrijn, Amsterdam Rijnkanaal, and the Lekkanaal. Vessels have to cross lock complex Hagestein, the Prinses Irenesluis, the prinses Marijkesluis, or the lock of weir complex Driel. These complexes limit the dimension of the components which have to be transported towards the new weir complex. Lock dimensions are given in Table C-17.

Table C-17 Lock dimensions Height of lifting gate Depth Width Length Lock +NAP [m] [m] [m] [m] Locks of Hagestein 4 18 220 Locks of Driel 4 18 260 Prinses Irenesluis 4 24 260 16.75 Prinses Marijkesluis 4 18 260

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C.4.5 SOIL PARAMETERS

Soil parameters are obtained by the internet application of TNO (TNO-DinoLoket, 2012). Three standard cone penetration tests nearby the location of the weir have been executed which are presented in Figure C-19, Figure C-20, and Figure C-21. The upper layer presented in Figure C-19 and Figure C-20 is weak. This is caused by the levee in which the sounding is made. The sill is located at -2.2m NAP, so the sill of the weir is located at:  5.84 metres for Figure C-19  11.07 metres for Figure C-20  6.31 metres for Figure C-21

A cross section of the bottom is given in Figure C-22. A Holocene layer is present from +4m NAP till a depth of -7.5m NAP, from there on a sand layer is present. The sand formation is called ‘the formation of Kreftenheye.’ This sand deposit is composed of course sand and gravel layers (Nederlands instituut voor Toegepaste Geowetenschappen TNO, 2003).

Figure C-19 CPT at the northern winter levee

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Figure C-20 CPT at the southern winter levee (sill level at 11.07 metre)

Figure C-21 CPT at the floodplain (sill level at 6.31 metres)

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Figure C-22 Cross section of the floodplains

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D. Stakeholder analysis

The purpose of a stakeholder analysis is to obtain an overview of all legitimate stakeholders, including their goals, interests, and their values. A list of stakeholders which are involved in any project related to the Nederrijn and Lek. are presented in this paragraph. First a description of the stakeholder is given. Secondly the interest, problem perception, goal and resources of the actors are listed the table of actors.

D.1 GOVERNMENT (RIJKSOVERHEID, 2012)

Two ministries are important for the this project, namely the ministry of Infrastructuur and Milieu and the ministry of Economische Zaken, Landbouw, en Innovatie. The ministry of Infrastructuur and Milieu focuses on the aspects of liveability and accessibility. The ministry is working on robust infrastructural connections by rail, road, water and air, protection against flooding, and the improvement of the quality of air and water. It has three primary tasks which are: making and developing of policy, implementation of policy, and inspection. The royal Dutch meteorological institute (KNMI), Rijkswaterstaat, and the inspection of environment and transport are agencies of this Ministry. The ministry of Economische Zaken, Landbouw, and Innovatie stands for the entrepreneurial Netherlands, with a strong aim at international competitiveness and attention to sustainability. They strive for an excellent business climate by creating the right conditions, creating a possibility for entrepreneurs for innovation and growth, focussing to our nature and environment, and by encouraging collaboration between researchers and entrepreneurs. They are building towards leading positions for agriculture, industry, services, and energy. A robust infrastructure and a well-functioning water system for agriculture and horticulture is present when the structures are reliable and sufficient available. Therefore the government has an interest in a sufficient reliable and available solution for the Nederrijn.

D.2 RIJKSWATERSTAAT (RIJKSWATERSTAAT, 2009)

Rijkswaterstaat manages and develops the national network of roads and waterways on behalf of the minister and secretaries. They are responsible for the management and maintenance of the main waterways, the flow of traffic on the waterways and roads, and the safety level. Rijkswaterstaat is the implementing organisation of the ministry of Infrastructuur and Milieu. The organisation is split up in sub-sections. Each section has the responsibility of a specific area of the Netherlands. The sub sections of Rijkswaterstaat are: IJsselmeergebied, Limburg, Noord-Brabant, Noord-Holland, Noord-Nederland, Oost- Nederland, Utrecht, Zeeland, and Zuid-Holland. The Nederrijn is located in the sub-section Oost- Nederland, the Lek in the sub-section Zuid-Holland, and the crossing channels (Merwedekanaal, Lekkanaal, and Amsterdam-Rijnkanaal) in the sub-section Utrecht. The tasks of Rijkswaterstaat are split up in three function groups, namely:  basic functions o safety o sufficient clean water

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o ecological healthy water  navigational functions o reliable and safe traffic on the waterways (for every traffic participant)  usage functions o drinking water o recreation o fishery o swimming water.

Rijkswaterstaat maintains the three basic functions of the waterways, so they are responsible for the condition of the weirs located in the Nederrijn and Lek. They have an interest in a reliable and available solution for the Nederrijn.

D.3 PROVINCES

The provinces are the supervisors of the water boards regarding the flood defences. Plans about primary flood defences have to be approved by the province on basis of the Wet op de Waterkeringen. The province also determines the distribution and the use of the provincial area in a ‘structuurvisie.’ Municipalities located in the province are using the ‘structuurvisies’ as a directive for their policy. Furthermore the province is responsible for the ground water management. Three provinces are located in the project area, namely the province of Gelderland, Utrecht, and Zuid- Holland. They want to establish a safe environment for the inhabitants and a well-functioning ground water system. Therefore, they are interested in a reliable and available solution for the reach Nederrijn-Lek which meets the safety levels and realises a well-functioning ground water system.

D.4 WATER BOARDS

The water boards are managing and controlling the water management function of a certain area. They are responsible for the flood defences, water quality, water quantity, and local waterway management. They are nearly always the initiator of flood defence projects. The water boards applicable in this project are: Stichtse Rijnlanden, Vallei & Eem, and Rivierenland. Their goal is the realisation of a safe environment for the inhabitants and a well-functioning water system. Therefore they are interested in a reliable and available solution for the reach Nederrijn-Lek which meets the safety levels and a well-functioning water system.

D.5 MUNICIPALITIES

Municipalities which are located along the Nederrijn and Lek are important for the design of a new canalization of the Nederrijn-Lek. ‘Bestemmingsplannen’ are made by the municipalities based on the ‘structuurvisies.’ The ‘bestemmingsplan’ determines the kind of use of a specific area. It describes the locations of stores, restaurants, businesses, etc. They are also the initiator of ‘small’ local civil engineering projects. The municipalities which are important in this project are shown in Figure C-5. (Recreational) harbours, quay walls, sluices, and other connections with the Nederrijn are located in these municipalities. People living in these municipalities are using the Nederrijn for recreation (swimming, sunbathing at the beaches at the floodplains, fishing, sailing, etc.) and profession. Municipalities are interested in the realisation of a living area which meets the demands of the inhabitants. They have to provide job opportunities and creating an area with good living circumstances, healthy conditions, and educational opportunities.

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D.6 INHABITANTS

People living along the Nederrijn have certain demands and wishes. They want to live in a nice, healthy, and safe environment with sufficient possibilities for employment. The inhabitants of a village or city are not represented by an organisation which defends their interests; they can organise themselves in (temporary) organisations. The problem perception varies per citizen. For example: fishers benefits from a free flowing river without weirs, because more fish is able to migrate through the river. But people with a yacht are not able to sail on the Nederrijn for a certain period of time and benefits from a river with weirs. Generally stated: inhabitants are interested in a safe environment and good living circumstances in which they can make money, enjoy themselves, undertake recreational activities etc.

D.7 KONINKLIJKE SCHUTTEVAER (KONINKLIJKE SCHUTTEVAER, 2012)

Koninklijke Schuttevaer is the representative of the interests of the inland navigation for more than 160 years. The interests represented by the Koninklijke Schuttevaer are nautical-technical issues of their members and the waterways which are important for the inland shipping. Some of these topics are:  More influence on the policies about the waterway and budgets.  Taking care of timely maintenance of the waterways and budgets for maintenance.  Promoting the widening and deepening of the waterways and harbours.  Is in favour for the optimal operation of the locks and bridges.  Is in favour for sufficient overnight accommodation, resting and waiting areas and their related facilities.

D.8 HISWA (HISWA, 2012)

The HISWA is an organisation for the recreational boating. This organisation represents two thirds of the turnover and employment of this sector. The activities of the HISWA are split up into two sections:  Supporting their members (1200 members) in their businesses. Members could contact the HISWA for judicial advice, questions about human resource management, health and safety (Dutch: arbo), permits, and environment.  Looking after the collective interests of the recreational boating. HISWA is the only organisation which defends the interests of the water sport industry. HISWA is lobbying on an international, European, national, and regional policy level. Topics of the lobby activities are: wet infrastructure, promotion of water sports, ‘bestemmingsplannen,’ dredging works, and taxes.

HISWA defends the interests of water recreation. Water recreation could take place when a safe situation is present and a sufficient draught for the recreational boats is available.

D.9 UTRECHTS LANDSCHAP (UTRECHTS LANDSCHAP, 2012)

Utrechts Landschap (founded in 1927) protects nature and culture located in the province of Utrecht. The main tool of this organisation is the purchase of land and historic buildings followed by a careful management and maintenance. The natural areas provide a habitat for plants and animals and contribute to the welfare of man. Utrechts Landschap defends the interests of nature and landscape of the province. The organisation operates with governments, businesses, and individuals to achieve these goals. They benefits from a situation with sufficient water of good quality for the natural areas.

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D.10 MOOI GELDERLAND (MOOI GELDERLAND, 2012)

The organisation ‘Mooi Gelderland’ is an umbrella organisation of the foundations ‘Geldersch Landschap’ and ‘Geldersche Kasteelen.’ ‘Mooi Gelderland’ manages, preserves, and protects the nature reserves, castles, and estates located in this province. 38.000 financiers, 500 volunteers, governments, funds, sponsors, and companies are supporting this organisation. ‘Mooi Gelderland’ defends the interests of nature and landscape of the province. The main tool of this organisation is the purchase of land and historic buildings followed by a careful management and maintenance. They benefits from a situation with sufficient water of good quality for the natural areas.

D.11 HART VOOR NATUUR (HART VOOR NATUUR, 2012)

‘Hart voor natuur’ is an umbrella organisation of more than 30 organisations committed to the future of the nature and the landscape in the Netherlands. Nature conservation organisations, water distribution companies, hiking platform, fishing associations, etcetera are combined in this coalition. ‘Hart voor natuur’ is preserving the natural areas in the Netherlands and is in favour for the ‘ecologische hoofdstructuur.’ This organisation is interested in the preservation and protection of nature in the future and the completion of the half implemented projects which have been stopped due to budget cuts.

D.12 ENTERPRISES

Several enterprises are the owner of a harbour. They are using this harbour for the transportation of cargo from and towards their factories. These factories are in favour of a well-functioning waterway for the transportation of their products and raw materials. A well-functioning waterway has sufficient draught for the vessels for a certain number of days per year and a well-functioning locks system. A list of enterprises located along the Nederrijn is given in C.2.2.10.

D.13 ENERGY SUPPLIERS

Two waterpower plants are located in the Nederrijn, one plant is realised in the middle pier of weir Hagestein and the other plant is realised next to weir Amerongen. Power plant Hagestein is out of service and property of Rijkswaterstaat. Power plant Hagestein is still exploited by the NUON. Both plants will be out of service or have to be adjusted if a new configuration of the Nederrijn is realised. The NUON and other energy suppliers are interested in ‘green’ and profitable solutions for the generation of electricity which could be implanted in the new solution.

D.14 LTO NEDERLAND (LTO NEDERLAND, 2012)

LTO Nederland represents nearly 50.000 farmers and horticultures in the Netherlands. This organisation is committed to the economic and social position of the farmers and horticultures. The agricultural sector in the Netherlands contributes to 10% of the national economy and more than 660.000 people are working in this sector. Managers and policy specialists of LTO are trying to influence ‘bestemmingsplannen,’ ‘streekplannen,’ and the policy of the government in favour of the agriculture and horticulture. The LTO is represented in the ‘Sociaal Economische Raad’ as an employer’ and employee organization. LTO Nederland is interested in a well-functioning (ground) water system from which farmers and horticultures can benefit.

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D.15 TABLE OF ACTORS

Table D-1 Overview of actors

Actor Interests Problem perception Goals Resources Importance A well-functioning water system in the Netherlands. The goals of the Dutch government are providing:  Water safety  Water quality Wellbeing of the inhabitants Losing an internal 'valve' (weir Driel) of the Financial resources, Government of the  Enough freshwater supply 1 and a well-functioning Netherlands. This leads to a not well- legalisation power ++ Netherlands  Cultural inheritance economy of the Netherlands functioning water system. decision power  Application of sustainable energy  Recreation  Shipping and navigation  Agricultural economy Maintaining the basic Losing an internal 'valve' (weir Driel). This functions and the Financial resources, 2 Rijkswaterstaat leads into a not well functioning water Providing the three function groups ++ infrastructure of the decision power system. waterways A not well functioning water system and on  Elaborating the three basic functions on Financial resources, Wellbeing of the inhabitants the long term, deviating ground water levels provincial scale 3 Provinces decision power, + of a province along the Nederrijn when a weir is out of  Responsible organisation for the ground 'structuurplannen' operation. water The tasks of the water boards are: Change in water levels in channels and rivers  Maintaining levees and other hydraulic (especially in Overijssel, Utrecht and structures at a sufficient safety level. Providing a safe and robust Financial resources, 4 Water boards Gelderland) which hampers the water  Supply, discharge, and (temporally) storage 0 water system decision power management function when a weir is out of of water. operation.  Purification of surface water and wastewater. The goals and tasks of the municipalities are:  Creating enough living space Wellbeing of the inhabitants  A well-functioning traffic system Financial resources, Losing certain (water management) functions 5 Municipalities of a city or village and a well-  Providing recreational areas decision power, 0 when a weir is out of operation. functioning local economy  Owner of municipal ports and exploiting bestemmingsplannen them  Responsible organisation for sewerage Good and safe living A (negative) change in their living conditions Living and working in a good and safe living area. With Demonstrating, 6 Inhabitants - conditions when a weir is out of operation sufficient jobs, recreation possibilities, etc. diffusive power

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Actor Interests Problem perception Goals Resources Importance

The goals of Koninklijke Schuttevaer are: Not meeting the interests of the inland  widening and deepening of the waterways Meetings with policy- Koninklijke Representing the interests of navigation when a weir is out of operation. and harbours makers, 7 + Schuttevaer the inland navigation Too low water levels will be present in the  optimal operation of locks and bridges media, IJssel and Nederrijn in this case.  enough overnight accommodations, resting lobby. and waiting areas, and their related facilities

The goals of HISWA are: Not meeting the interests of the water Meetings with policy-  Supporting the members in their business. Representing the interests of recreation when a weir is out of operation. makers, 8 HISWA  Representing the collective interests of the + the water recreation industry Too low water levels will be present in the media, recreational boating (well-functioning IJssel, Nederrijn and other local channels. lobby. waterways for recreation).

 Maintaining and expanding natural and Meeting with policy  Changed water levels which affect cultural areas in the province of Utrecht. makers, Protecting the environment the nature when a weir is out of Utrechts  Creating a broad support for the protection enterprises, 9 and cultural heritage of the operation. + Landschap of nature and landscape through public individuals, province of Utrecht  The reduction of natural areas in education and recruitment of 'nature media, the future. protectors' lobby.

Meeting with policy  Changed water levels which affect  Preserving and protecting the nature makers, Protecting the environment the nature when a weir is out of reserves, castles and estates. enterprises, 10 Mooi Gelderland and culture of the province of + operation.  Creating a broad support for the protection individuals. Gelderland  The reduction of natural areas of nature and landscape. Media, lobby.

Meeting with policy  Changed water levels which affect makers, Commitment to the future of the nature when a weir fails. Preserving and protecting nature in the future and enterprises, 11 Hart voor natuur the nature and landscape in  Budget cuts in the nature finishing the half implemented projects which have 0 individuals, the Netherlands conservation which leads to half been stopped due to the budget cuts. media implemented projects Lobby.

Diffusive power, Changed water levels which makes blocking power, Enterprises along Making (more) profit by using waterway transport which 12 Making profit transhipment impossible during a certain through + the Nederrijn is cheaper compared to road transport. amount of days demonstrations, lobby

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Actor Interests Problem perception Goals Resources Importance

Diffusive power, blocking power Changed water levels which influences the Generating as much electricity as possible using water 13 Energy suppliers Making profit through 0 yield of the waterpower plant heads generated by structures demonstrations, Lobby. Meeting with policy Representing the interests of Changing water levels which influences the Creating an as good as possible environment for the makers and 14 LTO Nederland + farmers and horticultures yield of the farms and horticultures agriculture along the Nederrijn-Lek individuals. Media and lobby.

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E. Requirements analyses

E.1 REQUIREMENT ANALYSIS OF THE REACH SELECTION (DESIGN LEVEL 1)

E.1.1 CUSTOMER EXPECTATIONS (STEP 1)

Defining and quantifying the customer expectations is the first step of the requirements analysis. The purpose of this step is to determine what the customer wants the system to accomplish and how well each function has to be fulfilled. The main customer in this project is Rijkswaterstaat. This actor wants to establish a well-functioning water system in the Netherlands. The expectations of the customer (Rijkswaterstaat) have to be implemented in the new design. The expectations of Rijkswaterstaat are covered within the three main function groups (Rijkswaterstaat, 2009):  Basic functions: o Defending the hinterland for flooding. . Retaining water. . Transportation of the maximum discharge at Lobith (16.000 m3/s in 2015 and 18.000 m3/s in 2050) through the delta towards sea and IJsselmeer. . Realising certain water levels by steering the runoff. . A safe manageable transportation of water. o Supplying clean and ecological healthy water. . Supplying water with a base quality (Kaderrichtlijn Water). . Providing passages for fauna. . Maintaining (and restoring) the natural habitats.  Navigational functions: o Enough draught has to be available for navigation.  Usage functions: o These functions are subordinate to the basic functions and the navigational functions; favourable conditions for the usage functions are created when the basic functions are fulfilled (Rijkswaterstaat, 2011).  Furthermore the alternative of renovation has to be implemented within one decennia.

E.1.2 CONSTRAINTS (STEP 2 & 3)

Constraints of the project are divided in two parts, namely internal and external constraints. Internal constraints (project and enterprise constrains) and external constraints are impacting the design solutions. The constraints considered on the reach level are internal constraints:  Water discharge capacity of the distinct branches must be equal to the applied discharge from the upper boundary.  A minimum flushing discharge of 25m3/s must be present in the Nederrijn.  A minimum discharge of 70m3/s must flow into the IJssel to guarantee the water buffer function.

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 Navigation on the main transport route (the Waal) may not be hampered. And external constraints:  The water which enters the Netherlands at the upper boundary must leave the Netherlands at the lower boundary.

E.1.3 OPERATIONAL SCENARIOS (STEP 4)

One operational scenario is taken into account, which is the increase of the maximum discharge from 16.000m3/s to 18.000m3/s. Taking into account the navigational prospects is beyond the scope in this graduation research.

E.1.4 MEASURES OF EFFECTIVENESS (STEP 5)

The measures of effectiveness and suitability define quantitatively how well the system performs with respect to the expectations of the customer. The expectations which have to be quantified are enumerated in the following enumeration:  The amount of water which has to flow into the Waal during dry periods.  The amount of water which has to flow into the Pannerdensch Kanaal during dry periods.  The amount of water which has to flow into the Nederrijn during dry periods.  The amount of water which has to flow into the Lek during dry periods.  The amount of water which has to flow into the Waal during wet periods.  The amount of water which has to flow into the Pannerdensch Kanaal during wet periods.  The amount of water which has to flow into the Nederrijn during wet periods.  The amount of water which has to flow into the Lek during wet periods.

E.1.5 SYSTEM BOUNDARIES (STEP 6)

The system boundary goes beyond the geographical boundary of a system. The system boundaries define which part is controllable by the engineer, and which part is not. Furthermore the interactions between system elements are described. The system boundaries are included in the main report and therefore not included in the appendices.

E.1.6 INTERFACES (STEP 7)

Interfaces are considered from an internal and external point of view. Internal interfaces address the elements inside the boundaries of the system. External interfaces address the relationship outside the system boundaries.

Internal interfaces are:  Water flowing from the upstream boundary into the Boven-Rijn.  Water flowing from the Boven-Rijn into the Waal.  Water flowing from the Boven-Rijn into the Pannerdensch Kanaal.  Water flowing into the Merwedes from the Waal.  Water flowing into the Nederrijn from the Pannerdensch Kanaal.  Water flowing into the Lek from the Nederrijn.  Water flowing into the IJssel from the Nederrijn.

External interfaces are:  Water entering the system.

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 Water leaving the system.

E.1.7 UTILIZATION ENVIRONMENTS (STEP 8)

Two scenarios are taken into account, namely a dry (summer scenario) and a wet (winter) scenario. The Rhine branches have to be able to transport the water from the upper boundary to the lower boundary for both scenarios.

E.1.8 LIFE CYLCLE PROCESS CONCEPTS (STEP 9)

The life cycle process requirements could be defined for the development, construction, deployment, operation, support, disposal, training, and verification of an object. These aspects are not defined in this design level, because the life cycle process concepts have to be defined for the structures. They are not strictly necessary for the selection of river branches.

E.1.9 FUNCTIONAL REQUIREMENTS (STEP 10)

Functions describe what the system has to perform. The functions are described for the subsystems of the ‘water supplier.’ In the next step the functional requirements are quantified. The subsystems and the functional requirements are:  the Boven-Rijn o waterway navigation & discharge  the Waal o waterway navigation & discharge  the Merwedes o waterway navigation & discharge  the Noord o waterway navigation & discharge  the Pannerdensch Kanaal o waterway navigation & discharge  the Nederrijn o waterway navigation & discharge  the Lek o waterway navigation & discharge  the IJssel o waterway navigation & discharge  bifurcation the IJsselkop o water distribution water management  bifurcation the Pannerdensche Kop o water distribution water management  confluence Noord and Lek o water distribution water management  confluence Beneden Merwede and Noord o water distribution water management

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E.1.10 PERFORMANCE REQUIREMENTS (STEP 11)

E.1.11 MODES OF OPERATION (STEP 12)

The Rhine branches are a partly dammed system and have therefore different modes of operation. Three modes of operation can be defined namely:  a fully dammed operation o A minimum flushing discharge flows through the dammed reach.  an undammed operation o Normal discharges flow through the rivers; the river is undammed.  a partly dammed operation. o Discharges are being regulated by one or more structures.

E.1.12 TECHNICAL PERFORMANCE MEASURES (STEP 13)

E.1.13 PHYSICAL CHARACTERISTICS (STEP 14)

Physical characteristics (like: colour, size weight, etc.) have to be defined in this step. For now, one physical characteristic is defined which is the width of the waterway. The width of the winter levees must remain the same to prevent major river improvement works which would take longer than 10 years.

E.1.14 HUMAN FACTORS (STEP 15)

The last step of the requirements analysis is de definition of human factors like physical space limits, climatic limits, ergonomics, etc. which affects the design. Human factors are not defined for now, because they are not relevant for the configuration of the Nederrijn. Human factors could be defined for the requirements of the weir, so human factors are elaborated in a different level.

E.2 REQUIREMENT ANALYISIS OF THE CONFIGURATION DESIGN (DESIGN LEVEL 2)

The requirements analysis described in E.1 is extended in this appendix. The requirements analysis does not have to be repeated again completely, because the steps are merely to analyse the project area and the opportunities. The total area of the Nederrijn and Lek has already been analysed, so overlap would be present when the total system is analysed once again. Steps which were not fully elaborated in E.2 are (further) elaborated in this chapter because they are applicable for the design level of the configuration itself.

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The requirement analysis starts with an input of objectives, description of the surroundings, and the stakeholder needs. This data is described in C.2 and D. Secondly functional requirements, performance requirements, internal constraints, and external constraints etc. are identified using the 15 steps for requirements analysis (Department of Defense, 2011).

E.2.1 CUSTOMER EXPECTATIONS (STEP 1)

Defining and quantifying the customer expectations is the first step of the requirements analysis. The purpose of this step is to determine what the customer wants the system to accomplish, and how well each function has to be fulfilled. The main customer in this project is Rijkswaterstaat. This actor wants to establish a well-functioning water system in the Netherlands. The expectations of this ‘customer’ is derived from the three main functions and two additional needs. The expectations are presented in the following enumeration (Rijkswaterstaat, 2009) and applied to the Nederrijn and Lek.  Fulfilling the basic functions: o safety . Prevention of too high water levels in the IJssel and Lek which could result in flooding. o sufficient clean water . Diversion of sufficient water into the IJssel. . Providing sufficient water for counteracting the salt intrusion from the North sea o (ecological) healthy water. . Guaranteeing a sufficient water quality by refreshing the Nederrijn-Lek.  Fulfilling the navigational functions: o a ‘smooth’ transport . As less hindrance and delays as possible caused by congested waterways, locks, and bridges. . Sufficient draught for navigation. o reliable and safe traffic on the waterways (for every traffic participant). . A robust waterway system.  Fulfilling the usage functions: o drinking water o recreation o fishery o swimming water.  Maintenance should be easily applied.  The structure must function over a lifetime of 100 years.

E.2.2 CONSTRAINTS (STEP 2 & 3)

Constraints of the project are divided in two parts, namely internal and external constraints. Internal constraints (project and enterprise constrains) and external constraints are impacting the design solutions. The constraints applicable for the canalization of the Nederrijn are presented in the following enumerations. Internal constraints:  Solutions should comply with the Euro codes.  A minimum flushing discharge must be present in the Nederrijn-Lek for maintaining a sufficient water quality.  A minimum of pumping stations and inlets should be redesigned for the implementation of the new solution.  Fish migration may not be negatively affected with respect to the present situation.

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 The floodplains must remain part of the ‘ecologische hoofdstructuur.’  Increased water levels in the Nederrijn-Lek caused by construction works (caused by temporally restrictions of the river profile) should be limited.  The regulation of discharge by the present weirs may not be hampered during construction.  The maximum capacity [m3/s] of the present system may not be negatively affected during construction.  Navigation may not be fully blocked during construction.  Existing harbours must be accessible during construction.  Pumping stations and inlets located along the Pannerdensch Kanaal, Nederrijn, and Lek should function normally during construction.  A new solution must be implemented within 10 years.

External constraints:  Sufficient fresh water has to be diverted into the IJsselmeer during dry periods.  Navigation over the Amsterdam-Rijnkanaal may not be hampered.  Navigation over the Lek Kanaal may not be hampered.  The maximum discharge capacity of the reach Nederrijn-Lek may not be decreased.  Solutions must comply with the European legislation: o kaderrichtlijn Water o natura 2000 (Europe direct, 2012) o high water directive (Dutch: hoogwaterrichtlijn) o swimming waters directive for protected areas (Dutch: Zwemwaterrichtlijn) o fishing waters directive for protected areas (Dutch: Viswaterrichtlijn) o shellfish waters directive for protected areas (Dutch: Schelpdierwaterrichtlijn).  Solutions must comply with the national legislation o water law (Dutch: Waterwet) o environmental maintaining law (Dutch: Wet milieubeheer) o natural protection law 1998 (Dutch: Natuurbeschermingswet 1998) o flora and Fauna law (Dutch: Flora en Fauna wet) o spatial planning law (Dutch: Wet ruimtelijke ordening).  Solutions must comply with the other norms and directives. o management plan of the waters (Dutch: Beheersplan Rijkswateren 2009-2015) o water level agreements (Dutch: Peilbesluiten) o water agreements (Waterakkoorden).

E.2.3 OPERATIONAL SCENARIOS (STEP 4)

Operational scenarios are used to describe plausible ‘futures’ and to extend the view of opportunities and threats. Scenarios can involve the evolvement of the political environment, social attitudes, changing climate, and economy. Operational scenarios are not predictions of the future, it are ‘just’ hypotheses of a possible future and the consequences of this future. They are used for the extension of our thinking of the opportunities and threats that the future might hold. Four aspects are defined in this research namely:  economic growth  economic decline  wetter climate conditions  drier climate conditions.

The economic effects are restricted to the trends in recreational boating, commercial shipping, and the available budgets. A project team of economists has to be set up to determine the effects of economic

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growth and economic decline in a thorough way for a real project. Possible consequences of economic growth and decline presented in this paragraph are based on the intuition of the author. A more detailed elaboration of operational scenarios is beyond the scope of this graduation research. The environmental effects are split up into two extremes, namely a wetter environment and a drier environment. Climate scenarios are defined by the KNMI reports published in the’ Nationaal Waterplan’ (Rijkswaterstaat, 2009). The reports of the KNMI states that the winters are becoming wetter and the summers drier, furthermore the behaviour of the weather is more fluctuating. These scenarios are partly taken over in this paragraph. Furthermore a ‘zero scenario’ is elaborated for the future developments of this system. The economic developments and the climate conditions are not changed with respect to the present situation for this scenario.

E.2.3.1 ZERO SCENARIO

Nothing is changed for the ‘zero scenario’ with respect to climate conditions and economic growth or decline. A minimum flushing discharge is present in the Nederrijn-Lek in a fully dammed situation which occurs during dry periods (discharges at Lobith lower than 1500m3/s). Furthermore, the discharge at the IJssel should be higher than 250m3/s and aimed to be 285m3/s for providing sufficient draught in the IJssel and realising a sufficient fresh water supply for the IJsselmeer in dry periods. The discharge of the Nederrijn-Lek is increased gradually by opening the gates for maintaining the aimed IJssel discharge of 285m3/s. An open Nederrijn is present for discharges at Lobith larger than 2350m3/s. On average, the system is fully closed for 3 months a year and partly closed during 6 months a year (Lemans, 2007). The water levels of the Nederrijn are regulated at the aimed water levels of +3m NAP for weir Hagestein and +6m NAP for weir Amerongen. Commercial and recreational traffic patterns remain the same as shown in C.2.2.8. The main cargo transport route is north-south orientated. The Nederrijn is more intensively used by recreation compared with the Amsterdam-Rijnkanaal and Lekkanaal.

E.2.3.2 ECONOMIC GROWTH

Due to the economic growth, more transport will take place in the system,. The crossing channels and the reach Nederrijn-Lek will become busier. Next to the increase of commercial vessels, more recreational boats will be purchased. This increase could hamper the cargo transport on its turn and could result in unsafe situations. A specialisation of river branches could be a solution for solving this interference. The Nederrijn could be specialised in a ‘local’ waterway for local transport and recreation and the Waal in a ‘main cargo waterway’ for large cargo vessels to separate both functions. Large vessels are not strictly necessary in the Nederrijn corridor given the size of the commercial harbours located along the Nederrijn; only small scale harbours of enterprises and harbours owned by cities like Wageningen and Arnhem are present as presented in C.2.2.10. Vessels which are using the Nederrijn corridor as transport route have to use the Amsterdam-Rijnkanaal, Lekkanaal, and the Waal in order to sail from the harbours located in Rotterdam and Amsterdam towards Germany and vice versa. This results in an increased use of locks located in the Amsterdam-Rijnkanaal and the Lekkanaal. Therefore the capacity of these locks has to be increased. Designs are already being made for the enlargement of the capacity of the Beatrixsluis and the widening of the Lekkanaal (Rijkswaterstaat, 2012). Adaptation works for other locks could be implemented in this scenario because sufficient funds are probably available in a growing economy. River improvements works after the implementation the project Room for the River could be implemented in the Nederrijn-Lek with use of the available budgets. The present maximum discharge of the Lek is set at 3.376m3/s according to the Nationaal Waterplan. A larger discharge is not preferred due to the narrow profile along the Lek; river improvements are too costly in the opinion of the government. However, the

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maximum discharge of the Lek could be increased when sufficient budgets are available in order to reduce the impact of river improvements along the Waal and the IJssel. At last, the recommendation of the second delta committee for the closable but open Rijnmond (Dutch: ‘afsluitbaar open Rijnmond’) could be implemented; sufficient budgets are probably available for the implementation of this recommendation into this scenario.

E.2.3.3 ECONOMIC DECLINE

Less recreation and less navigation are present in this system with repsect to the zero scenario, so the present river system has sufficient capacity for the vessels and boats. The majority of vessels on the Nederrijn which are sailing towards the east are sailing back along with the flow on the Waal to save fuel. No budget is available for the specialisation of waterways due to the economic decline, so recreational and commercial shipping are still sharing the waterways. The hindrance between commercial and recreational boating is less severe compared to the previous scenario which is caused by the decreased shipping intensity. River improvements at the downstream side of the system (the Lek) are too costly due to the narrow river profile. Budgets could better be used for the increase of the discharge capacity of the Waal and IJssel because more available space is present along these rivers. Space is already been reserved for the implementation of the second phase of the project Room for the River (increasing the maximum discharge of the Rhine branches to 18.000 m3/s). Furthermore fewer budgets are available for the conservation of nature along the Nederrijn and Lek. The implementation and the improvement of the ecologische hoofdstructuur could be stopped due to budget cuts.

E.2.3.4 WETTER CLIMATE CONDITIONS

The summer and winters are wetter and the average discharge and maximum discharge of the Rhine is increased with respect to the zero scenario. More space in the delta has to be reserved for the second phase of the project Room for the River. The water levels and discharge of the rivers in the Netherlands are higher compared to the zero scenario, so less water has to be diverted towards the IJssel by weir Driel. This results into a decreased usage of the weirs located in the Nederrijn with respect to the zero scenario. The project ‘closable but open Rijnmond’ could be implemented in this situation to prevent the Rijnmond from flooding during a storm surge and a high river runoff. Te same minimum of discharge has to be diverted into the IJssel for realising a sufficient water level during dry periods. This occurs less often with respect to the present situation. Furthermore, more water is available in a wetter climate which results in an increased flushing discharge of the Nederrijn. So, the minimum flushing discharge is less often present in this situation with respect to the zero scenario which is beneficial for nature. Finally, the pumping stations located along the Nederrijn and Lek have to be renovated, replaced, or built to increase their capacity. This is needed because more participation will be present in the Netherlands, so more water has to be pumped out of the polder. Furthermore, the capacity of the Nederrijn has to be high enough for the discharge of the polder water.

E.2.3.5 DRIER CLIMATE CONDITIONS

Less discharge and longer droughts takes place compared to the zero scenario. The Nederrijn-Lek is longer fully or partly dammed with respect to the zero scenario. A dammed Nederrijn-Lek could even occur for a whole year (Lemans, 2007). Water levels of the IJssel could be maintained at a sufficient level to the detriment of the flushing discharge. Navigation on the IJssel and Nederrijn could be hampered by the low

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draughts caused by the longer droughts. The water which is let into the polder by inlets is of a lower quality with respect to the zero scenario due to the lower flushing discharge of the Nederrijn. This will affect the quality of the area located along the local waterways. Furthermore, natural areas located in the system are losing their quality due to the low water quality and droughts. The wet cross section of the Nederrijn could be decreased to keep the water in the system at a sufficient quality by using a lower flushing discharge. In this way, the quality of the water which is let in by inlets located at the levees along the Nederrijn is of the same quality as the water quality in the zero scenario because the relative flushing rate remains the same.

E.2.4 MEASURES OF EFFECTIVENESS (STEP 5)

The measures of effectiveness and suitability define quantitatively how well the system performs with respect to the expectations of the customer (Rijkswaterstaat). The expectations which have to be quantified are enumerated in the following enumeration:  The total amount of water which has to flow into the IJssel during dry periods.  The total amount of water which has to flow into the Nederrijn during dry periods.  The total amount of water which has to flow into the Lek during dry periods.  The maximum amount of water which has to flow into the Nederrijn during design conditions.  The maximum amount of water which has to flow into the Lek during design conditions.  The maximum water levels of the Nederrijn during design conditions.  The maximum water levels of the Lek during design conditions.  Fish migration in the Nederrijn and Lek.  The design water levels for navigation during a dammed situation.  The reliability of the structures which are built for the new situations.  The availability of the structures which are built for the new situation.  The minimal dammed water levels of the Nederrijn.  The minimal dammed water levels of the Lek.

E.2.5 SYSTEM BOUNDARIES (STEP 6)

E.2.6 INTERFACES (STEP 7)

Internal interfaces are:  flow of water from the Nederrijn into the Lek  transportation of sediments from the Nederrijn into the Lek  interaction of species living in the system  draughts of harbours located in the system caused by the water level in the Nederrijn and Lek  interaction between the (new) water levels and the surroundings  (ground) water levels of natural areas in the floodplains  navigation moving through the system.

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External interfaces are:  water and sediments entering the system  water and sediments leaving the system  properties of pumping stations o pump curves of the pumping stations  dimensions of inlets o sill heights o rates of flow  normal water levels (Dutch: streefpeilen) of connected channels o Amsterdam-Rijnkanaal o Lekkanaal o Merwedekanaal  dimensions of present sluices o prinses Irenesluizen (northern part of the Amsterdam-Rijnkanaal) o prinses Marijkesluis (southern part of the Amsterdam-Rijnkanaal) o prinses Beatrixsluis (Lekkanaal) o Koninginnensluis (northern part of the Merwedekanaal) o Grote Sluis (southern part of the Merwedekanaal)  dimensions of a present barrier o barrier Ravenswaaij  legislation  the recommendations of the second delta committee o closable but open Rijnmond  municipalities located along the Nederrijn  migration of species living in the Rhine branches  navigation crossing, entering, and leaving the system.

E.2.7 UTILIZATION ENVIRONMENTS (STEP 8)

The utilization environments (weather conditions, temperature ranges, topologies) are not defined for configuration design. They are not relevant for the second phase.

E.2.8 LIFE CYCLE PROCESS CONCEPTS (STEP 9)

The life cycle process requirements could be defined for the development, construction, deployment, operation, support, disposal, training, and verification. These aspects are not defined in this level because the life cycle process concepts have to be defined for the structures.

E.2.9 FUNCTIONAL REQUIREMENTS (STEP 10)

Functions describe what the system has to perform. The functions are described for the subsystems of the ‘water supplier.’ In the next step the functional requirements are quantified.  bifurcation the IJsselkop o water distribution water management  Nederrijn section (IJsselkop-Wijk bij Duurstede) o (fish migration) link between IJsselkop and Lek ecology o waterway navigation & discharge o sediment transport morphology

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 Lek section (Wijk bij Duurstede-confluence Lek & Noord) o link between Nederrijn and the river Noord ecology o waterway navigation & discharge o sediment transport morphology o implementation of ‘closable but open Rijnmond’ water safety  floodplains o nature ecology o flood protection water safety o farm land agricultural o drains water management  levees and hydraulic structures along the Nederrijn and Lek o design heights water safety o roads transport o harbours transport o municipalities living  waterway connections of channels along the Nederrijn and Lek o waterways navigation o drains water management o water levels water management  pumping stations situated from the IJsselkop to confluence Lek & Noord o flood protection of the polders water safety o water quantity water management  inlets situated from the IJsselkop to confluence Lek & Noord o water quality environment & ecology o water quantity water management

E.2.10 PERFORMANCE REQUIREMENTS (STEP 11)

A performance requirement describes how well a system functions or how well it has to be accomplished. A performance requirement is expressed in terms like a degree, rate, quantity, quality, etc. The performance requirements are defined for the subsystems of paragraph E.2.9  bifurcation the IJsselkop o Supplying a minimum discharge to the IJssel of 250 m3/s. o Supplying an aimed discharge to the IJssel of 285 m3/s. o Supplying a minimum discharge to system ‘The water supplier’ of 25 m3/s. o Processing a maximum discharge of 5800 m3/s 2015. o Processing a maximum discharge of 6150 m3/s in 2100.  Nederrijn section (IJsselkop-Wijk bij Duurstede) o Processing a maximum discharge of 3.376 m3/s in 2015. o Processing a maximum discharge of 3.376 m3/s in 2100. o Navigational class Va must sail in the Nederrijn. o Transport of sediments must be possible in an fully open situation.  Lek section (Wijk bij Duurstede- confluence Lek&Noord) o Processing a maximum discharge of 3.376 m3/s in 2015. o Processing a maximum discharge of 3.376 m3/s in 2100. o Navigational class Vla must sail at the Lek o Transport of sediments must be possible for a fully open situation.  floodplains o The roughness of the floodplains may not increase with respect to the present situation.

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 levees and hydraulic structures along the Nederrijn and Lek o Functioning at a design flood of 1/1250 per year along the Nederrijn. o Functioning at a design flood of 1/2000 per year along the Lek.  waterway connections of channels along the Nederrijn and Lek o Minimal lock dimensions have to be equal or larger with respect to the present situation which is presented in Table C-8. o Traffic intensity has to be equal or higher with respect to the present situation.  pumping stations situated from Arnhem to confluence Lek&Noord o As little as possible adaptations of pumping stations should be applied.  inlets situated from Arnhem to confluence Lek&Noord o As little as possible adaptations of inlets should be applied.  Reliability of the structures in the systemi o The reliability of the waterway is equal to or higher compared to the fully renovated situation. Corresponding requirements are defined for the requirements for the structures in the next level.  Availability of the structures in the system o The availability of the waterway is equal to or higher with respect to the present renovated situation.  Maintainability o Maintenance should be easily applied  Safety o Sudden release of water is not allowed. Releasing of water must take longer than 3 hours. o The system has to be designed with the design codes for implementing a sufficient safety level. o The health and safety legislation (Dutch: ARBO) must be taken into account.

E.2.11 MODES OF OPERATION (STEP 12)

Three different modes of operations are distinguished namely a closed operation with a minimum flushing discharge, an open operation, and an intermediate operation. The state of operation is determined by the amount of discharge flowing through the Pannerdensch Kanaal. System ‘The water supplier’ is fully closed when the aimed discharge of the IJssel is not met. First sufficient water (285 m3/s) has to flow into the IJssel before more water than the flushing discharge is let in the system. The corresponding maximum water level at the IJsselkop is +8.30m NAP for the realisation of this discharge (Lemans, 2007). The system is in an intermediate operation when the discharge of the Pannerdensch Kanaal is higher compared to the sum of the aimed discharge of the IJssel (285 m3/s) and the flushing discharge (25 m3/s). The division of discharge over the ‘The water supplier’ and the IJssel is regulated by the system. The system is in a fully open operation when the discharge of the Nederrijn is higher compared to the maximum discharge which can be controlled by the system. Nowadays this maximum controllable discharge by weirs is 2350 m3/s at Lobith (Rijkswaterstaat, 2011).

E.2.12 TECHNICAL PERFORMANCE MEASURES (STEP 13)

Technical performance measures are measurements or controls which are executed during the design and manufacturing process. The outcomes of the measurements are used for the verification of the design. Technical performance measures in the configuration design level are:  minimum draughts [m] measured for the three modes of operation presented in paragraph E.2.11  minimum discharge [m3/s] measured in the IJssel

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 minimum discharge [m3/s] measured in the Nederrijn and Lek  maximum discharge capacity [m3/s] of the Nederrijn and Lek.

These technical performance measures have to be verified to the design which has been made in the end of the first design level.

E.2.13 PHYSICAL CHARACTERISTICS (STEP 14)

Physical characteristics (like: colour, size, weight, etc.) have to be defined in this step following the handbook. This is not be done for this level because defining the physical characteristics constricts the solution space of the project. Properties about the weight or size could differ per variant which has to be elaborated in the next level.

E.2.14 HUMAN FACTORS (STEP 15)

The last step of the requirements analysis is the definition of human factors like physical space limits, climatic limits, ergonomics etc. which affects the design. The human factors are not defined because they are not relevant for the configuration of the Nederrijn. Human factors could be defined for the requirements of the weir, so human factors will be elaborated during the next design level.

E.3 REQUIREMENTS ANALYSIS OF THE WEIR LOCATION AND WEIR CONFIGURATION (DESIGN LEVEL 3)

The requirements analysis described in E.2 is extended in this section. The requirements analysis does not have to be repeated again completely, because these steps are merely to analyse the project area and the opportunities. The total area of the Rhine branches and the Nederrijn and the Lek is already analysed so overlap would be present when the total system is analysed once again. Therefore, the requirements analysis presented in this section is focused on the reach: Wijk bij Duurstede-Culemborg. Steps which were not fully elaborated in E.1 and E.2 are (further) elaborated in this chapter because they are applicable for the design level of the structure itself. The requirement analysis starts with an input of objectives, description of the surroundings, and the stakeholder needs. This data is described in appendix C.3 and D. Secondly functional requirements, performance requirements, internal constraints, external constraints, etc. are identified using the 15 steps for requirements analysis (Department of Defense, 2011).

E.3.1 CUSTOMER EXPECTATIONS (STEP 1)

The expectations of the customer Rijkswaterstaat are not changed with respect to E.2.1 because the customer has not been changed.

E.3.2 CONSTRAINTS (STEP 2 & 3)

Constraints of the project are divided in two parts, namely internal and external constraints. Internal constraints (project and enterprise constraints) and external constraints are impacting the design solutions. The constraints applicable for the weir location are described in the following enumerations: Internal constraints:  No extra internal constraints are identified. The internal constraints are already covered by the constraints presented in E.1.2 and E.2.2. External constraints:

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 Sufficient space needs to be present for the implementation of the weir.  The building location must be easily accessible.  Ships must be able to sail easily into the approach channel of the weir.  The weir may not be a too large constriction for the river.  The weir may not be an enlargement of the cross section of the river.

E.3.3 OPERATIONAL SCENARIOS (STEP 4)

The operational scenarios are used to describe plausible ‘futures’ and to extend the view of opportunities and threats in this graduation research. The scenarios described in E.2.3 are used for the list of requirements presented in F.2. There is no need for further elaboration of this list because the extracted requirements for the future are already taken into account in the configuration variants.

E.3.4 MEASURES OF EFFECTIVENESS (STEP 5)

The measures of effectiveness and suitability of the total system are also applicable for the design of the weir. The measures of effectiveness of the total system are fulfilled during well-functioning of the water retaining object (weir complex Culemborg). However, some extra measures of effectiveness could be identified for the weir, namely:  the length of the straight approach channel  the minimum width of the weir openings  the maximum width of the weir openings  the width of each opening.

E.3.5 SYSTEM BOUNDARIES (STEP 6)

E.3.6 INTERFACES (STEP 7)

Internal interfaces are:  the flow of water from the upstream boundary to the downstream boundary through the new object  the transportation of sediments from the upstream boundary to the downstream boundary through the new object  the interaction of species in the system  water levels and depths of the system  navigation sailing from the upstream boundary to the downstream boundary through the object.

External interfaces are:  water and sediments entering the system  water and sediments leaving the system

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 migration of species living in the Rhine branches  minimum water levels at the constriction at Wageningen  minimum water levels at weir Driel  maximum water levels at Ravenswaaij.

E.3.7 UTILIZATION ENVIRONMENTS (STEP 8)

The utilization environments (weather conditions, temperature ranges, topologies) are now yet defined for the project area of weir complex Culemborg. The utilization environments are defined for the requirements analysis of the weir design.

E.3.8 LIFE CYCLE PROCESS CONCEPTS (STEP 9)

The life cycle process requirements could be defined for the development, construction, deployment, operation, support, disposal, training and verification. These aspects are not defined in this level of the design project because the life cycle process concepts have to be defined for the structures.

E.3.9 FUNCTIONAL REQUIREMENTS (STEP 10)

Functions describe what the system has to perform. The functions are described for the subsystem.  reach Wijk bij Duurstede-weir Hagestein. o link between the Nederrijn and the downstream side of weir Hagestein water management o waterway navigation & discharge o sediment transport morphology  levees and hydraulic structures o design heights water safety o roads transport o harbours transport o municipalities living  floodplains o nature ecology o flood protection water safety o farm land agriculture  weir Culemborg o water level control water management o water discharge control water management o fish passage ecology o navigational passage navigation

E.3.10 PERFORMANCE REQUIREMENTS (STEP 11)

A performance requirement describes how well a system functions or how well it has to be accomplished. A performance requirement is expressed in terms like a degree, rate, quantity, quality, etc. Performance requirements are already defined for the Nederrijn and Lek. Now the performance requirements for the system for weir location are defined.  Reach Wijk bij Duurstede – weir Hagestein o Processing a maximum discharge of 3.376 m3/s in 2015. o Processing a maximum discharge of 3.376 m3/s in 2100.

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o Transportation of sediments must be possible for an open situation. o Navigational class Va must be able to sail through the reach for open river conditions.  Levees and hydraulic structures o Functioning at a design flood of 1/2000 per year.  Floodplains o The roughness of the floodplains may not increase with respect to the present situation. o Sufficient space must be available for the construction of the new weir.  weir complex Culemborg o Generating a minimum draught of 4.2 metres at the reach weir complex Driel - weir complex Culemborg in order to generate sufficient draught for Va vessels. o Let through a minimum discharge of 25m3/s for fully dammed conditions. o Let recreational AM class boats pass the complexes during dammed periods. o Let Va class vessels pass the complexes during open periods.

E.3.11 MODES OF OPERATION (STEP 12)

The modes of operation are already covered in the description presented in E.2.11.

E.3.12 TECHNICAL PERFORMANCE MEASURES (STEP 13)

Technical performance measures are measurements or controls which are executed during the design and manufacturing process. The outcomes of the measurements are used for the verification of the design. As example: the strength of concrete in N/mm2 or the minimum discharge through the system in m3/s is a technical performance measurement which can be verified in the project. Technical performance measures for the reach Wijk bij Duurstede – weir Hagestein are:  minimal upstream water level at weir complex Culemborg  maximal upstream water level at Ravenswaaij  minimal upstream water level at Wageningen  minimal upstream water level at weir complex Driel  minimal depth at the sill of the weir complexes  minimal length of the approach channels  minimal and maximal width of the approach channels.

E.3.13 PHYSICAL CHARACTERISTICS (STEP 14)

Physical characteristics have to be defined in this step following the handbook. Physical characteristics for this level are the dimensions of the approach channels (width, curvature of the channel, length, straight line of sight etc.)

E.3.14 HUMAN FACTORS (STEP 15)

A straight line of sight for navigation must be present for the shippers. A straight line of sight with the navigational openings is to facilitate the incoming vessel to enter and exit the structure without the need to make sharp turns.

E.4 REQUIREMENTS FOR WEIR DESIGN (DESIGN LEVEL 4)

The requirements analysis described in E.3 is focused on the weir location. The requirement analysis does not have to be repeated completely again, because the 15 steps are merely to analyse the project area and

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the opportunities. The requirements analysis presented in this section is focussed on the weir structure located at the floodplains upstream from the village of Culemborg. The requirement analysis starts with an input of objectives, description of the surroundings, and the stakeholder needs. Secondly; functional requirements, performance requirements, internal constraints, external constraints, etc. are identified using the 15 steps for requirements analysis (Department of Defense, 2011).

E.4.1 CUSTOMER EXPECTATIONS (STEP 1)

The expectations of the customer Rijkswaterstaat are not changed with respect to E.2.1 because the customer has not been changed.

E.4.2 CONSTRAINTS (STEP 2 & 3)

Internal constraints:  Solutions should comply with the Euro codes.  A maintenance friendly solution is preferred.  Navigation may not be fully blocked during construction.  Existing harbours must be accessible during construction.  The gates have to be adjustable in height for the discharge regulation.  The gates have to be fully removed for creating enough clearance.

External constraints:  Vessels have to be able to pass the complex.  Minimum water levels have to be regulated for the upstream river.  Solutions must comply with the national legislation given in E.2.2.  Solutions must comply with the norms and directives given in E.2.2.  Solutions must comply with the European legislation given in E.2.2.

E.4.3 OPERATIONAL SCENARIOS (STEP 4)

E.4.4 MEASURES OF EFFECTIVENESS (STEP 5)

The measures of effectiveness and suitability of the total system are also applicable for the design of the weir. The measures of effectiveness of the total system and of the system Wijk bij Duurstede – weir Amerongen are fulfilled during well-functioning of the water retaining object (weir complex Culemborg). However, some extra measures of effectiveness could be identified for the weir, namely:  The depths generated by the weir for the reach Driel-Culemborg  The average number of days for which the river is open.

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E.4.5 SYSTEM BOUNDARIES (STEP 6)

E.4.6 INTERFACES (STEP 7)

Interfaces are considered from an internal and external point of view. Internal interfaces address the element inside the boundaries of the system. External interfaces address the relationship outside the boundaries.

Internal interfaces are:  The flow of water from the upstream side of the weir to the downstream side of the weir.  The transportation of sediments from the upstream side of the weir to the downstream side of the weir.  Navigation sailing from the upstream side of the weir to the downstream side of the weir and vice versa.  The transition from the river bed to the upstream bottom protection.  The transition from the upstream bottom protection to the foundation.  The transition from the foundation to the downstream bottom protection.  The transition from the downstream bottom protection to the downstream river bed.  The water tight closure of the gates with the foundation for a fully dammed operation.  The connection of the gates with the pylons.  The connection of the pylons with the foundation.  The mechanical equipment necessary for moving the gate.  Upstream (dammed) water levels.  Downstream (dammed) water levels.

External interfaces are:  Water and sediment entering the system.  Water and sediment leaving the system.  The connection of the weir with the embankment.  The connection of the weir foundation to the soil.

E.4.7 UTILIZATION ENVIRONMENTS (STEP 8)

The weir complex has to be able to operate during and within the described environments.  The weir complex must cope with the (normal and extreme) weather conditions in the Netherlands.  The weir complex must cope with the temperature ranges present in the Netherlands.  The weir must be operational 24 hours a day.  The weir must remain stable for all conditions.  The weir must be ready for the predicted future conditions.

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E.4.8 LIFE CYCLE PROGRESS CONCEPTS (STEP 9)

The system life is split up in 8 life cycle functions which are: development, construction, deployment operation, support, disposal, training, and verification. The total life span of the system is analysed by addressing these life cycle functions:  development o The environment of the sub-system has to be analysed in order to create an optimal solution. The characteristics of the surroundings (dimensions, rate of flow, etc.) have to be taken into account and the stakeholders must be involved as much as possible. Furthermore a location for the weir has to be found with ‘good’ soil qualities. This minimizes the foundation works and the costs of the weir. The determined location should also be easily accessible in order to build the object.  construction o The impact on the surroundings during construction must be as limited as possible. Nature may not be negatively affected on the long term and the ‘ecologische hoofdstructuur’ and fish migration must remain during construction. Furthermore navigation may not be hindered during construction.  deployment o The separate systems which have been constructed have to be integrated into one well- functioning system. The separate systems (like the gates) have to be tuned to the discharge control and water level control system of the weir complex.  operation o Manuals have to be made for the operation of the weir complexes for the deployed people. The manuals must cover all situations which could occur.  support o A maintenance plan and maintenance schedule has to be made in order to maintain the weir complexes. The maintenance schedule should include the visual checks, when certain materials should be ordered, and when maintenance has to be performed. The maintenance plan describes the maintenance activities; it describes how certain systems work, and how to maintain them. Easy access to the site is important in order to be able to perform maintenance during operation (PIANC, 2006).  disposal o A plan has to be made about how the structures have to be disposed. The disposal of the structures has to be taken into account during the design.  training o People have to be trained in order to operate the new weir complexes. They have to learn how to control the rate of flow, water levels, fish migration, lock passages, etc.  verification. o A verification plan has to be made which has to be used for quality checks. These quality checks have to be performed during design, construction, operation, and disposal of the complexes in order to guarantee a well-functioning system.

E.4.9 FUNCTIONAL REQUIREMENTS (STEP 10)

Functions describe what the system has to perform. The functions are described for the ‘weir system.’ In the next step, the functional requirements are quantified.  Upstream approach channel o Providing a straight line of sight. o Providing a straight channel.

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o Providing enough discharge capacity.  Downstream approach channel o Providing a straight line of sight. o Providing a straight channel. o Providing enough discharge capacity.  Upstream bottom protection o Defending the river bed near the weir for scour. o Defending the river bed near the weir for erosion.  Downstream bottom protection o Defending the river bed near the weir for scour. o Defending the river bed near the weir for erosion. o Absorbing the energy of the flow through the complex.  Gates o Diverting water. o Regulating discharge. o Transferring the hydrostatic and hydrodynamic forces to the pylons.  Pylons o Transferring the forces from the gates to the foundation. o Providing space for the gate operation machinery.  Foundation o Transferring the loads to the subsoil. o Providing stability of the weir complexes. o Providing a stiff foundation for the weir complex. o Preventing excessive seepage. o Preventing piping.

E.4.10 PERFORMANCE REQUIREMENTS (STEP 11)

A performance requirement describes how well a system functions or how well it has to be accomplished. A performance requirement is expressed in terms like a degree, rate, quantity, quality, etc. The performance requirements are defined for the subsystems of E.4.9.  Upstream approach channel o Providing a line of sight of 1 kilometre (based on reference projects). o Processing a maximum discharge of 3.376 m3/s in 2015. o Processing a maximum discharge of 3.376 m3/s in 2100.  Downstream approach channel o Providing a line of sight of 1 kilometre (based on reference projects). o Processing a maximum discharge of 3.376 m3/s in 2015. o Processing a maximum discharge of 3.376 m3/s in 2100  Upstream bottom protection o Remaining stable for the maximum flow speed. . Dimensions and configuration needs to be determined in further investigation. o Defending the subsoil from erosion. . Dimensions and configuration needs to be determined in further investigation. o Preventing the development of a large scour hole near the structure. . Dimensions and configuration needs to be determined in further investigation.  Downstream bottom protection o Remaining stable for the maximum flow speed. . Dimensions and configuration needs to be determined in further investigation.

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o Defending the subsoil from erosion. . Dimensions and configuration needs to be determined in further investigation. o Preventing the development of a large scour hole near the structure . Dimensions and configuration needs to be determined in further investigation. o Absorbing the energy. . Dimensions and configuration needs to be determined in further investigation.  Gates o Maintaining a maximum upstream water level of +5.00m NAP o Remaining water tight for a fully dammed operation . Maximum deflection of 100mm for a segment gate (Sloten, 2012) o Being adjustable in height from +5.00m NAP (fully dammed) to -2.2m NAP (sill level)  Pylons o Transferring the forces from the gates to the foundation . Dimensions and configuration needs to be determined in further investigation. o Providing space for the gate operation machinery . The width of the pylon is assumed to be 12.5 metres which is based on the Thames barrier and Ems barrier. o The top of the pylons must be higher with respect to the maximum design water level . The top of the pylon must be located at the maximum design water level +1 metre.  The top of the pylons are located at +7.5m NAP or higher (Table C-3).  Foundation o Being stable for all load configurations . Dimensions and configuration needs to be determined in further investigation. o Being resistant for piping . Dimensions and configuration needs to be determined in further investigation. o Preventing excessive seepage . Dimensions and configuration needs to be determined in further investigation.

E.4.11 MODES OF OPERATON (STEP 12)

The modes of operation presented in E.2.11 also hold for the weir structure.

E.4.12 TECHNICAL PERFORMANCE MEASURES (STEP 13)

Technical performance measures are measurements or controls which are executed during the design and manufacturing process. The outcomes of the measurements are used for the verification of the design. As example: the strength of concrete in N/mm2 or the minimal discharge through the Nederrijn in m3/s. Technical performance measures for the weir are:  sliding stability of the weir complex  overturning stability of the weir complex  piping length of the weir complex  maximum soil pressure of the subsoil  deformations of the soil and weir complex  deflection of the gates  maximum stress in the concrete  maximum stress in the gates  maximum strain in the concrete  maximum strain in the fibre reinforced polymers

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 stability of the bottom protection  depth of the upstream and downstream scour hole  discharge through the complex  upstream dammed water levels.

E.4.13 PHYSICAL CHARACTERISTICS (STEP 14)

Physical characteristics (like: colour, size, weight, etc.) have to be defined in this step following the handbook. The physical characteristics are defined for the weir structure. The structure must be heavy enough in order to remain stable for all conditions, so concrete is used for the substructure. The gate could be made out of steel, concrete, fibre reinforced polymers, or other materials. A well balanced trade-off should be made in order to determine the right material for the design of the gates. However, this trade-off is not executed in this graduation research; a design choice for fibre reinforced polymers is made in the context of this graduation research. The total width of the openings of the weir must larger than 100 metres and smaller than 140 metres. The minimal width per opening is 41 metres which is sufficient for 2 lane Va vessels.

E.4.14 HUMAN FACTORS (STEP 15)

The last step of the requirements analysis is the definition of human factors like physical space limits, climatic limits, ergonomics etc. which affects the design of the structure. The human factors are taken into account by implementing the ARBO legislation into the design.

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F. Lists of requirements

The next step of systems engineering methodology is the subdivision of the requirements. The requirements are subdivided into five categories namely (ARCADIS, 2008) & (CROW, 2011):  functional requirements (FR) o Requirements about what the system must be able to perform.  aspect requirements (AR) o Requirements about the non-functional performances.  internal interface requirements (IR) o Requirements about the internal interfaces.  external interface requirements (ER) o Requirements about the external interfaces.  realisation requirements (RR) o Requirements about the implementation of the new object and systems.

F.1 REACH REQUIREMENTS (DESIGN LEVEL 1)

First, the requirements for the reach design are subtracted from the requirements analysis described in E.1. The requirements are subdivided into the five categories as stated earlier in this appendix.

F.1.1 FUNCTIONAL REQUIREMENTS

FR 1) Sufficient discharge capacity must be available. FR 1.1) The river branches must have a discharge capacity of 16.000m3/s in 2015. FR 1.2) The river branches must have a discharge capacity of 18.000m3/s in 2100. FR 2) An ecological healthy situation must be present. FR 2.1) A minimum flushing discharge of 25m3/s must be present in the Nederrijn and Lek. FR 2.2) A minimum discharge of 70m3/s must be present in the IJssel for dry periods (Rijkswaterstaat; Directie bovenrivieren, 1979). FR 2.3) A minimum discharge of 110m3/s must be present in the IJssel for extreme dry periods (Rijkswaterstaat; Directie bovenrivieren, 1979). FR 3) Sufficient draught must be present in the waterways. FR 3.1) Sufficient draught must be present in the Waal. FR 3.2) Sufficient draught must be present in the Nederrijn and Lek. FR 3.3) Sufficient draught must be present in the IJssel.

F.1.2 ASPECT REQUIREMENTS

No aspect requirements are identified for this level. The aspect requirements are elaborated in sequential levels.

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F.1.3 INTERNAL INTERFACE REQUIREMENTS

IR 1) The system has to comply with the legal safety levels presented in Figure A-12. IR 2) Navigation on the main transport route (the Waal) may not be hampered. IR 3) The width of the waterways must remain in the same order to prevent major river improvements.

F.1.4 EXTERNAL INTERFACE REQUIREMENTS

ER 1) The combined discharge capacity must be equal to the maximum discharge entering the Netherlands. ER 1.1) The discharge capacity must be 16.000m3/s in 2015. ER 1.2) The discharge capacity must be 18.000m3/s in 2100. ER 2) The discharge capacity of the downstream water system must be equal to the maximum discharge capacity of the system. ER 2.1) The discharge capacity must be 16.000m3/s in 2015. ER 2.2) The discharge capacity must be 18.000m3/s in 2100.

F.1.5 REALISATION REQUIREMENTS

RR 1) Implementation of the system may not influence the present situation. RR 1.1) Navigation may not be fully blocked during execution. RR 1.2) Damming operation of the present situation may not be affected during execution. RR 2) The new canalization must be implemented within one decade.

F.2 CONFIGURATION REQUIREMENTS (DESIGN LEVEL 2)

The next step is the expansion of the requirements which are defined in F.1. The requirement analysis presented in E.2 is used as input for this second list of the requirements. The requirements are subdivided into five categories and enumerat4ed in the sequential sections (ARCADIS, 2008) & (CROW, 2011).

F.2.1 FUNCTIONAL REQUIREMENTS

FR 4) The system has to realise sufficient draught in the IJssel. FR 4.1) A minimum of 250 m3/s has to be diverted into the IJssel. FR 4.2) An aimed discharge of 285 m3/s has to be diverted into the IJssel. FR 4.2.1) A minimum water level at the IJsselkop of +8.30m NAP has to be realised. FR 5) An ecological healthy situation must be present. FR 5.1) A minimal flushing discharge of 25 m3/s must always be present in the system. FR 6) A sufficient high water discharge capacity must be present. FR 6.1) A discharge capacity of 3.376 m3/s must be available in 2015. FR 6.2) A discharge capacity of 3.376 m3/s must be available in 2100. FR 7) The same kind of navigation has to be able to use the system. FR 7.1) CEMT class Vla has to be able to navigate on the Lek. FR 7.2) CEMT class Va has to be able to navigate on the Nederrijn. FR 7.3) Recreational class MA has to be able to use the Nederrijn and Lek FR 8) Sufficient vertical clearance for commercial shipping must be present during high water. FR 8.1) A minimal clearance needed for ships with 4 layers of containers must be present. FR 8.1.1) The minimum vertical clearance is 9.10m (including a 0.30m marge). FR 9) Nature should not be negatively affected.

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FR 9.1) The ‘ecologische hoofdstructuur’ must be implemented. FR 9.2) Natura2000 conditions must be taken into account.

F.2.2 ASPECT REQUIREMENTS

AR 1) The system must have a sufficient reliability. AR 1.1) The reliability of the system must be equal to or higher compared to a fully renovated present system. AR 2) The system must have a sufficient availability. AR 2.1) The availability of the system must be equal to or higher compared to a fully renovated present system. AR 2.2) The system must function for 100 years. AR 3) The solutions must comply with the European legislation presented in E.2.2. AR 4) The solutions must comply with the national legislation presented in E.2.2. AR 5) The solutions must comply with the norms directives presented in E.2.2.

F.2.3 INTERNAL INTERFACE REQUIREMENTS

IR 4) Nature may not be negatively affected. IR 4.1) (Fish) migration must be provided in the system. IR 5) Harbours along the Nederrijn and Lek must be accessible for the new situation. IR 5.1) A draught of 3.0 metres must be available in the commercial harbour (basin). IR 5.2) A draught of 1.50 metres must be available in the recreational harbour (basin). IR 6) Minimal lock dimensions must remain equal or become larger with respect to the present situation. IR 7) The transport capacity of the Lek must equal the capacity of the Nederrijn IR 7.1) The water discharge capacity of the Lek must equal the discharge capacity of the Nederrijn. IR 7.2) The sediment transportation capacity of the Lek must equal the sediment transport capacity of the Nederrijn.

F.2.4 EXTERNAL INTERFACE REQUIREMENTS

ER 3) The system has to facilitate a well-functioning water system of the connected waterways. ER 3.1) Inlets have to be able to extract the same amount of discharge with respect to the present situation. ER 3.2) Pumping stations have to be able to drain the same amount of discharge with respect to the present situation. ER 4) Cross system functions may not be hindered. ER 4.1) Traffic on the crossing bridges may not be hindered. ER 4.2) Bridges crossing the Nederrijn and Lek may not be closed unless planned. ER 4.3) Navigational transport of the crossing waterways may not be hindered. ER 4.3.1) Navigation may not face an increase in delay or hindrance compared to the present situation. ER 4.4) Fish migration between river branches must be provided. ER 5) The water system of other branches may not be negatively affected by water set up caused by the system. ER 5.1) Water levels at the IJssel may not increase during maximum design discharge. ER 5.2) Water levels at the Waal may not increase during maximum design discharge.

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F.2.5 REALISATION REQUIREMENTS

RR 3) A minimum of pumping stations and inlets should be redesigned for the implementation of the new solution.

F.3 REQUIREMENTS FOR WEIR LOCATION AND CONFIGURATION (DESIGN LEVEL 3)

The next step of systems engineering is the expansion of the requirements which are defined in F.2. The area analysis presented in C.3 and the requirement analysis presented in E.3 is used as input for this second list of the requirements. The requirements are subdivided into five categories (ARCADIS, 2008) & (CROW, 2011).

F.3.1 FUNCTIONAL REQUIREMENTS

FR 10) Vessel class Va must be able to sail through the system. FR 10.1) A width equal or larger with respect to the Lek must be present. FR 11) Recreational class AM must be able to sail through the system. FR 11.1) A width equal or larger with respect to the Lek must be present FR 12) Flow velocities must be limited for navigation in a new situation. FR 12.1) Flow velocities must be lower than 2 m/s averaged over the cross (Meijer, 2012). FR 13) The head over a fully opened weir must be limited. FR 13.1) Depending on the flow velocity, so defining a value for the head is twofold. Therefore, this requirement is expired.

F.3.2 ASPECT REQUIREMENTS

AR 6) The weir complex alignment has to be designed according the PIANC design manual for weirs and storm surge barriers (PIANC, 2006).

F.3.3 INTERNAL INTERFACE REQUIREMENTS

IR 8) The structure should be situated in order to provide a straight line of sight with the structure navigational openings to facilitate the incoming vessel to enter and exit the gate without the need to make sharp turns (PIANC, 2006). IR 9) The structure should be situated in a position that minimizes cross current in the area where ships navigate (PIANC, 2006). IR 10) The capacity of the upstream river section must be equal to the capacity of the downstream river section. IR 10.1) The upstream discharge capacity must be equal to the downstream capacity IR 10.2) The sediment transportation capacity must be equal to the downstream capacity.

F.3.4 EXTERNAL INTERFACE REQUIREMENTS

ER 6) The capacity of the system must be equal to the capacity of the Nederrijn. ER 6.1) The discharge capacity must be equal to the discharge capacity of the Nederrijn. ER 6.2) The sediment transportation capacity must be equal to the capacity of the Nederrijn. ER 7) The described minimum or maximum dam regime of appendix O.1 must be maintained. ER 8) Winter levees and floodplains may not be submerged during a dammed operation.

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F.3.5 REALISATION REQUIREMENTS

RR 4) Easy access to the site during the construction period and operation must be available (PIANC, 2006). RR 5) Modifications to the existing river channel should be selected in a way to minimize the environmental impact of the structure and its size. RR 5.1) The structure should be built at a site where there is a natural restriction of the river (PIANC, 2006).

F.4 REQUIREMENTS FOR WEIR DESIGN (DESIGN LEVEL 4)

The next step is the expansion of the requirements which are defined in F.3. The requirement analysis presented in E.4 is used as input for this second list of the requirements. The requirements are subdivided into five categories (ARCADIS, 2008) & (CROW, 2011).

F.4.1 FUNCTIONAL REQUIREMENTS

FR 14) Vessel class Va must be able to sail through the weir. FR 14.1) A width of 29 metres (22,8 metres + 0,05*length of vessel) must be present for one lane traffic per opening. (Rijkswaterstaat, 2011). FR 14.2) A width of 41 metres must be present for two lane traffic per opening (Rijkswaterstaat, 2011). FR 14.3) A vertical clearance of 9,1 metres must be present (Rijkswaterstaat, 2011). FR 14.4) A depth of 4,2 metres must be present (Rijkswaterstaat, 2011). FR 15) Recreational class AM must be able to sail through the weir. FR 15.1) A width of 25 metres must be present (Rijkswaterstaat, 2011). FR 15.2) A height of 3.75 metres should be available (Rijkswaterstaat, 2011). FR 15.3) A depth of 1,80 metres should be available (Rijkswaterstaat, 2011). FR 16) A water level difference of 5.5m must be retained FR 16.1) A maximum upstream water level of +5.00m NAP must be realised for zero discharge. FR 16.2) The minimum downstream water level is -0.5m NAP. FR 17) The structure must remain stable for all conditions. FR 17.1) Piping must be avoided. FR 17.2) The structure may not slide away. FR 17.3) The structure may not overturn. FR 17.4) The maximum soil stress may not be exceeded. FR 18) The structure must be protected for bed degradation. FR 18.1) The object should be protected for scour holes. FR 18.2) The object should be protected against bottom erosion. FR 19) The pylons must be located above the high water reference line. FR 19.1) The top of the pylons are located at +7.5m NAP or higher. FR 20) Sufficient space needs to be present at the pylons for the operating systems. FR 20.1) The width of the pylons is 12.5 metres. FR 21) A minimum discharge of 25m3/s must be regulated during fully dammed conditions. FR 21.1) A watertight closure of the gates with the foundation has to be realised for a fully dammed operation.

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F.4.2 ASPECT REQUIREMENTS

AR 7) Maintenance and inspection has to be easily applied. AR 8) A safe solution must be obtained during a (party) closed operation. AR 8.1) Sudden release of water is not allowed. AR 8.1.1) Releasing water must take longer than 3 hours. AR 9) The weir complexes have to be designed according to the working conditions legislation (Dutch: ARBO wetgeving). AR 10) The structures have to be designed according to the Eurocode. AR 11) The gates have to be designed using FRP. AR 11.1) The fibre reinforced parts of the structures have to be designed according to CUR96.

F.4.3 INTERNAL INTERFACE REQUIREMENTS

IR 11) The depth at the weir and approach channels must be equal or larger than 4.2 metres. IR 12) The gates have to be adjustable from +5.00m NAP to -2.2m NAP.

F.4.4 EXTERNAL INTERFACE REQUIREMENTS

ER 9) The capacity of the system must be equal to the capacity of the Lek. ER 9.1) The discharge capacity must be equal to the discharge capacity of the Lek. ER 9.2) The sediment transportation capacity must be equal to the capacity of the Lek.

F.4.5 REALISATION REQUIREMENTS

The realisation requirements are already covered in the previous sections.

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G. Functional analyses

G.1 FUNCTIONAL ANALYSIS OF REACH SELECTION (DESIGN LEVEL 1)

The functional analysis is the step in between the overview of requirements and the elaboration and selection of variants. The outputs of the requirements analysis are converted into a description of system functions which are helpful for the design synthesis.

G.1.1 SYSTEM FUNCTIONS

The system functions are covered in the main report

G.1.2 SYSTEM OBJECTS AND ALLOCATION OF REQUIREMENTS

The requirements subtracted from the requirement analysis are related to the sub-systems which are presented in the object tree (Rijkswaterstaat & ProRail, 2009). The result of the allocation is presented in Figure G-1.

Figure G-1 Requirement specification tree for level 1. FR=functional requirement, IR=Internal constraint requirement, ER=External constraint requirement, and RR=realisation requirement.

G.2 FUNCTIONAL ANALYSIS OF CONFIGURATION DESIGN (DESIGN LEVEL 2)

The functional analysis of design level 1 is elaborated in further detail in this section. The functions elaborated in paragraph G.2.1 are transformed into objects which are shown in paragraph G.2.2. In the last step the developed requirements are allocated to derived objects.

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G.2.1 SYSTEM FUNCTIONS

First the system ‘The water supplier’ is defined in functional terms. These top-level functions are decomposed into sub-functions in the second breakdown (Department of Defense, 2011) & (CROW, 2011). The functional tree is included in the main report. The start of the sequence for this design level is the upper boundary of the system which represents the water inflow of the system. The end of the sequence is the lower boundary of the system which represents the water outflow of the system. Several steps are located in between the upper and lower boundary represented by five blocks in Figure G-2. Every block represents a sub-function of the system. First, water enters the system at the IJsselkop. Secondly, the water is transported from the IJsselkop towards the subsystem which distributes the water over the IJsselkop and the Nederrijn-Lek. The quantity of the inflow is measured and a certain amount of discharge is allowed to pass the system depending on the inflow rate. Thirdly, the outflow of water of the sub system is transported towards a system which retains water. This system has to regulate a certain water level for complying with the navigational functions, ecological functions and user functions. A specific amount of discharge is passed depending on the measured water level. The outflow of this system is transported towards the lower boundary located at the confluence of the Lek-Noord. The water leaves the main system at this lower boundary.

Figure G-2 Functional flow block diagram of the system for design level 2

G.2.2 SYSTEM OBJECTS AND ALLOCATION OF REQUIREMENTS

System ‘The water supplier’ is split up in three sub-systems based on the functional flow block diagram presented in Figure G-2. The system consists of water transportation objects, objects which retain water, and objects which determine the discharge regulation over the IJssel. The object tree is included in the main report. The requirements of the requirements analysis are related to the sub-systems. The allocation of requirements is presented Figure G-3. The allocated requirements specifications per object are the starting point of the design synthesis.

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Figure G-3 Requirements specification tree for design level 2 (FR=Functional requirement, AR=Aspect requirement, IR= internal interface requirement, ER=external interface requirement, and RR=Realisation requirement)

G.3 FUNCTIONAL ANALYISIS OF WEIR LOCATION AND CONFIGURATION (DESIGN LEVEL 3)

The functional analysis of design level 2 is elaborated in further detail in this section. The functions elaborated in paragraph G.3.1 are transformed into objects which are shown in paragraph G.3.2. In the last step, the requirements are allocated to the derived objects.

G.3.1 SYSTEM FUNCTIONS

The functions which the system has to fulfil are included in the main report. A functional flow block diagram is used for the transformation of the functions to objects. The start of the sequence for this design level is the upper boundary of the system which represents the inflow of the system. The end of the sequence is the lower boundary of the system which represents the outflow from the system. Several steps are located in between the upper and lower boundary, represented in three blocks in Figure G-4. Every block represents a sub-function of the system. First, the water enters the system from the upstream Lek section. Secondly, the water and sediments are transported towards the object. The amount of water which flows through the object is regulated by the object. A percentage of the upstream water flow is retained at the upstream side of the object in order to generate a sufficient water level. Thirdly, the water and sediments are transported towards the lower boundary. In the end, the water and sediments leaves the system and flows back into the downstream section of the weir.

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Figure G-4 Functional flow block diagram for design level 3

G.3.2 SYSTEM OBJECTS AND ALLOCATION OF REQUIREMENTS

The system ’the water retainer’ is split up in two sub-systems which is based on the functional flow block diagram of Figure G-4. The sub-systems which are derived from the functional analysis are included in the main report. The allocated requirements are presented in Figure G-5.

Figure G-5 Requirements specification tree for design level 3 (FR=Functional requirement, AR=Aspect requirement, IR= internal interface requirement, ER=external interface requirement, and RR=Realisation requirement)

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G.4 FUNCTIONAL ANALYSIS OF WEIR DESIGN (DESIGN LEVEL 4)

The functional analysis described in G.3 could now be elaborated in further detail for the weir complex. The ‘water retention object’ presented in the second row of Figure G-5 is subdivided in four sub-objects as presented in Figure G-6. A subset of requirements is derived for every sub-object. However, the focus is aimed at the weir structure within this graduation research. Therefore, no requirements are defined for the lock complex, the fish passage, the banks, and the floodplains.

Figure G-6 The sub-objects of a water retention object

G.4.1 SYSTEM OBJECTS AND ALLOCATION OF REQUIREMENTS

The object to be designed in the context of this graduation research is the ‘weir construction’ which is described as ‘weir Culemborg.’ This weir has to regulate the upstream water levels, has to let through the navigation for an open river situation, and should contribute to the maximum discharge capacity during high water events. The weir itself is composed of five objects as presented in the object tree in the main report. The inlet structure is the part which protects the upstream bottom for erosion. The substructure forms the foundation of the superstructure and the gate. The substructure transfers the forces to the subsoil and has to limit the ground water flow underneath the structure. The superstructure is fixed on the substructure and transfers the forces of the gate towards the substructure. The superstructure is located in the embankment and the river. Therefore, the superstructure has to withstand a part of the hydraulic loads. The superstructure supports also the gates and the operating machinery. The gates are the water retaining element of the weir complex and have to withstand the water flow and the head. The gate(s) transfer the (hydraulic) forces to the superstructure. The outlet structure has to protect the bottom from erosion and eventually to prevent even more seepage underneath the structure. Furthermore an energy dissipating structure like a stilling basin could be designed at the outlet structure. The requirements presented in F.4 are allocated to the objects defined in the last row of the object tree. The allocation of the requirements is presented in Figure G-7. The allocated requirements specifications per object are the starting point of the design synthesis of the weir itself.

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Figure G-7 Requirements specification tree for design level 4 (FR=Functional requirement, AR=Aspect requirement, IR= internal interface requirement, ER=external interface requirement, and RR=Realisation requirement)

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H. Design criteria

H.1 DESIGN CRITERIA FOR REACH SELECTION (DESIGN LEVEL 1)

Design criteria are necessary for the comparison of the variants. Design criteria used for the reach selection are:  duration of the implementation  minimum flow rate through the river branches  maximum flow rate through the river branches. A qualitative selection is performed for the selection of the most suitable redesign of the river branches in the Netherlands.

H.2 DESIGN CRITERIA FOR CONFIGURATION DESIGN (DESIGN LEVEL 2)

Design criteria are necessary for the comparison of variants. Criteria important for the design of the configuration of the Nederrijn deals with: direct-, indirect-, secondary-, short-, medium-, long term-, permanent-, temporary-, positive-, and negative impacts of the project on its surroundings. An environmental impact assessment has to be executed in order to determine the effects of a new situation. The criteria, which are normally used in an environmental impact assessment, are listed in Table H-1. This list of criteria originates from the environmental impact assessment applied by ARCADIS, the manual: ‘Environmental Impact Assessment: A guide to procedures’ (Department for Communities and Local Government, 2006) and a checklist for weirs presented in ‘Design of movable weirs and storm surge barriers’ (PIANC, 2006). The application of design criteria differs from the application of requirements. Variants do have to comply with the requirements. Only the variants which fulfil the requirements can be compared in a multi criteria analysis using the design criteria. The design criteria are used in a multi criteria analysis for determining the score. A score indicates how ‘good’ a variant performes with respect to other variants. A full scale environmental impact assessment is not performed in this research due to the set time constraint of this thesis. A selection between the main criteria and the irrelevant criteria for this design level has to be made for the multi criteria analysis. This selection iss presented in the last column of Table H-1.

Table H-1 General Environmental impact assessment criteria.

Topic Aspect Criteria Application within the project Supplying a minimum Not important in this level. This will be regulated by the objects Water quality flushing discharge of the and not by the configuration. Nederrijn and Lek. Diverting enough water Important in this level. Location of the most upstream weir Water quantity into the IJssel determines the discharge steering capacity. Water Modification of the bed Important in this level. Implementation of weirs affects the Morphology profile and the sediment bottom profile and the sedimentation patterns which on its characteristics. turn affects the need for dredging Drainage Capacity of pumping Important in this level. Pumping stations and sluices have to be patterns stations and sluices able to fulfil their functions (with or without adaptations).

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Topic Aspect Criteria Application within the project Natural Maintaining natura2000 Important in this level; the operation of the objects affects protection law goals and areas the (ground)water level of the natural areas 1998 Maintaining the Nature Ecologische Not important in this level. Objects will be built in the Ecologische hoofdstructuur summer bed, so linkage on the floodplains is maintained. Hoofdstructuur. Enabling fish to pass the Important in this level. The more objects are built in the river Fish migration weirs. bed; the more difficult it is to use the Nederrijn-Lek. Integration of the new Spatial Not important in this level. The exact location of the weirs situation in the planning will be chosen in the next level. surroundings Daily impact on local Important in this level. The impact of changed water levels on communities (jobs, Social impact the municipalities along the Nederrijn and Lek has to be taken (Living) economy, transport, into account. environment recreation etc.) Visual disturbance of new Not important in this level. Layout of the objects will not be Visual aspects objects designed in this level. Nuisance Hindrance (noise, emission Not important in this level. Layout of the objects will not be caused by the etc.) designed in this level. new situation Important in this level. Sufficient vertical clearance, keel Commercial Safety and capacity. clearance, and navigational width should be present in the shipping (distances, currents, etc.) new situation. Recreational activities on the river banks, Important. A changed configuration results in a shift/change Recreation Usage recreational boating, of recreational activities. functions fishery. Capacity of roads, Important. A changed configuration results in a changed Infrastructure railways, and waterways. transport network. Agriculture and Ground water quality and Important. Changed groundwater quality or quantity results horticulture quantity. in a changed yield of agricultures and horticultures. Safety level with respect to Not important in this level. Governing high water levels will Water safety high waters. not significantly change in the new configuration. Regulating the water levels Not important. This is determined by the objects which will Water level in the Nederrijn and Lek in be built in the Nederrijn and Lek, so this is important in the control a full and partial dammed next level. Technical operation. functioning Not important. This is determined by the objects which will Maintainability of the new Maintenance be built in the Nederrijn and Lek, so this is important in the situation. next level. Not important. This is determined by the objects which will Design A robust or adaptable be built in the Nederrijn and Lek, so this is important in the characteristics design next level. Hindrance for Temporally closure of Not important in this level; this will be elaborated in the next road transport bridges design level. Hindrance for Temporally reduction of Not important in this level; this will be elaborated in the next commercial navigation design level. shipping Hindrance for Temporally reduction of Not important in this level; this will be elaborated in the next recreational Construction navigation design level. navigation Reducing Enabling the minimum and Not important in this level; this will be elaborated in the next discharge maximum discharge design level. capacity capacity Important in this level; the adaptations of the present Construction Implementation within 10 configuration must be implemented within 10 years from time years now.

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H.3 DESIGN CRITERIA FOR WEIR LOCATION AND CONFIGURATION (DESIGN LEVEL 3)

Design criteria are needed in order to compare the different variants. The design criteria are based on the list of design criteria presented in H.2 and the handbook ‘design of hydraulic gates’ (Erbisti, 2004). The design criteria used for comparing the different options is presented in Table H-2. A score will be assigned per variant for every aspect in order to compare the variants.

Table H-2 Design criteria for the weirs

Aspect Criteria Sensitivity to malfunctions, human errors, ship collisions Reliability Vulnerability to foundation distortions, vibrations etc. Vulnerability to sediments, ice, debris, etc. Capacity and accuracy of river control Convenience of operation Operation Unavailability for operation due to maintenance Construction time Aspect Criteria Construction impact on navigation conditions Navigation Maintenance impact on navigation conditions Navigation safety and convenience Maintainability of all areas and details Access to maintenance sensible components Maintenance Maintainability under operation conditions Health and safety of maintenance crews Aesthetics Social impacts Noise

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I. Redesign of the waterways

This chapter presents the full reasoning per reach for the redesign of the waterways. The opportunities of water managing structures per reach (IJssel, Waal, Pannerdensch Kanaal, and the Nederrijn-Lek) are defined. The design choice is included in the main report.

I.1 WAAL

I.1.1 LOWER DISCHARGES

The Waal could be dammed off using weirs or the discharge could be decreased by river retaining works to increase the depth of the reach Nederrijn-Lek and IJssel and diverting sufficient water into the IJssel in periods of droughts. This is not a preferred solution because the navigation would be hindered by the structures built in the waterway or by the decreased depth resulting from the decreased discharge. The traffic intensity of the Waal is larger compared to the IJssel, Nederrijn-Lek, and Pannerdensch Kanaal; so damming this reach has a bigger disadvantage compared to river works located in the other river branches. A reduced discharge in periods with high discharges is not preferable because the Waal is the main water discharge route and the capacity of the Nederrijn-Lek could not be increased further due to the narrow river profile of the Lek to take over the discharge of the Waal. A further increase of capacity of the IJssel during high discharge introduces levee reinforcements over a length of 125 km which is undesired given the time constraint of 10 years and the implementation costs.

I.1.2 EQUAL DISCHARGE

Equal discharges for periods with low discharge are necessary for counteracting the salt intrusion and realising a sufficient draught for navigation. Equal discharges in periods with high discharge are possible; the system is able to facilitate the design discharge at Lobith of 16.000 m3/s after implementation of Room for the River.

I.1.3 HIGHER DISCHARGES

Higher discharges during dry periods are not preferable because less water is diverted into the IJssel and the Nederrijn-Lek which is detrimental for the fresh water supply of the IJsselmeer and the draught of the IJssel and the Nederrijn-Lek. Structures have to be built in the Nederrijn and IJssel in order to realise sufficient draught. So, not only one reach has to be redesigned, but two reaches which takes more time and is more costly than redesigning one reach. During wet periods, higher discharges after the implementation of Room for the River are possible because space is already reserved for the implementation of a second phase of this project. The Waal is able to transport more than the present 64% of the total runoff. This has to be regulated by a by a water

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regulating structure at Pannerdensche Kop. Furthermore this option could reduce the risk for high waters at the IJsselmeer when insufficient water can be discharged by the sluices into the Waddenzee. The drawback of implementing a discharge regulating device at the Pannerdensche Kop (which controls the water division by river width reduction or river depth reduction) is the set-up of water in the upstream direction. Diverting 419 m3/s more into the Waal causes a set-up of 0.39 metres when the MHW-discharge is present. River widening projects or heightening of levees is necessary to guarantee a sufficient safety level which is expensive (Hermeling, 2004). Furthermore, the implementation of a discharge regulation device and the river adjustments would probably not be implemented within 10 years.

I.2 IJSSEL

I.2.1 LOWER DISCHARGE

The discharge in the IJssel could be decreased from a navigational point of view when weirs are built which regulates the water levels of the IJssel. A drawback of this solution is the water supply of the IJsselmeer and the Twente Kanaal. A minimum of 70 m3/s has to flow through the IJssel to guarantee the water buffer function of the IJsselmeer during dry periods. During an extreme dry summer, a minimum discharge of 110 m3/s has to flow through the IJssel (calculated for the summer of 1976 with a minimum discharge of 850 m3/s at Lobith and a fully supplied IJsselmeer). (Rijkswaterstaat; Directie bovenrivieren, 1979). Less discharge is available for the Nederrijn-Lek, so draughts are smaller in this reach and less water is available for refreshing the Nederrijn and Lek with respect to the present situation. Therefore, more water has to flow towards the Nederrijn-Lek which negatively affects the draught of Waal. River improvements have to be executed in the Waal for increasing the draught which would take longer than the stated time constraint or weirs have to be built or renovated in the Nederrijn-Lek. Both solutions are not preferred because river improvements have to be executed in the IJssel, Nederrijn or Waal in order to guarantee a well-functioning system which is more costly than redesigning one reach. A lower discharge in the IJssel in wet periods is advantageous for relieving the IJsselmeer during high waters when the needed discharge capacity at the Afsluitdijk is not met. The drawback of this option is an increase of capacity which has to be realised in other river branches. The capacity of the reach Nederrijn- Lek could not be further increased due to the narrow river profile at the Lek. So river improvements have to be executed along the Waal. These river improvements works would probably last for longer than 10 years, so this is not an option in this research project.

I.2.2 EQUAL DISCHARGE

An equal discharge is possible; the system is able to facilitate the design discharge at Lobith of 16.000 m3/s after the implementation of Room for the River. Furthermore a sufficient draught is realised in dry periods.

I.2.3 HIGHER DISCHARGE

Higher discharges for dry periods are beneficial for the draught of ships. A minimal draught of 2.5 metres is realised for a discharge of 250 m3/s. The probability per year of an undercut of this discharge is roughly 21% (based on: (ARCADIS, 2010)). The discharge needed for the increase in draught has to be extracted from the Waal or Nederrijn-Lek. Water extraction from the Waal hampers the more intensive navigation. The consequences of a too low draught at the Waal compared to the consequences of a too low draught of the IJssel are larger, so the available water could better be used for an increase of draught at the Waal

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(Lemans, 2007). Water extraction from the Nederrijn-Lek is not desirable because only a flushing discharge of 25 m3/s is present in dry periods. During wet periods, higher discharges are possible because space is already reserved for the implementation of a second phase of ‘Room for the River’. More capacity could easier be obtained with respect to the Waal because less critical trajectories are present along the IJssel (Rijkswaterstaat, 2009). The layout of bifurcation IJsselkop and Pannerdensche Kop has to be adjusted in order to divert a higher fraction of water into the IJssel compared to present situation.

I.3 NEDERRIJN AND LEK

I.3.1 LOWER DISCHARGES

Lower discharges could be possible if the Nederrijn-Lek is fully dammed or the discharge in this reach is further reduced for dry periods. In the first case, zero discharge is present which results in major changes in the surrounding (ground) water levels and waterway connections. This possibility is not easily implementable because major adaptations have to be made in order to divert the river runoff in case of a flood wave and major adaptations of the water management systems around the Nederrijn-Lek have to be executed. Another option is a lower discharge compared to the present situation (25 m3/s) but this is not beneficial for the base quality of water. So reducing the discharge of the Nederrijn in dry and wet periods is not a solution on a timescale of 10 years.

I.3.2 EQUAL DISCHARGE

An equal minimum discharge is possible and an equal maximum discharge through the Nederrijn-Lek is possible when the weirs are being replaced or the same dam-regime is implemented. Less disadvantages occur when this option is implemented. This option could be implemented within the set time constraint.

I.3.3 HIGHER DISCHARGES

A higher discharge is possible in case of a removal of the weirs or increasing the minimum flushing discharge in periods of low discharge. The drawback of this option is a decrease of draught in the IJssel which hampers the navigation and a decrease of quantity of water which is diverted into the IJssel. River retaining structures like weirs or constrictions have to be built in the IJssel in order to generate sufficient draught in the IJssel. Another option is diverting more water into the Pannerdensch Kanaal by a ‘control device.’ This device extracts water from the Waal which leads into a decrease of draught which is unfavourable. So, a higher discharge for dry periods results in major river improvements of the IJssel and the Waal. This will take more than 10 years and is not an option in the scope of this project. Higher maximum discharges in the Nederrijn are not preferred due to the narrow river profile at the Lek. Measures which could be implemented to increase the discharge capacity are not easily implementable and lead to high costs. (Rijkswaterstaat, 2009). Increasing the capacity of the IJssel and Waal is a better solution than additional river works in the Nederrijn. Space is already been reserved for the increase of capacity for the second phase of Room for the River.

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J. New configuration of the Nederrijn-Lek

This chapter describes the design process of the configuration variants. The variants are developed during several brainstorm sessions. The brainstorm sessions are individually performed during four days. The results of the brainstorm sessions were discussed with H.G. Voortman for review. A selection of variants is made after each brainstorm session. Only the most promising variants are elaborated in further detail during the sequential brainstorm sessions and elimination rounds.

J.1 FIRST BRAINSTORM SESSION

The amount of weirs and the location of the weirs of the reach Nederrijn-Lek are the design variables of the first brainstorm session. Five options were taken into account which are: one weir configurations, two weirs configurations, three weirs configurations, four weirs configurations, and ‘many’ weirs configuration. The options were grouped by the amount of weirs. So, a variant group ‘one weir solutions,’ a variant group ‘two weir solutions,’ etc. was made. Infinite configurations (any number of weirs located at any locations) are possible using this approach, so a more sophisticated approach had to be used. However a decision could be made for one weir configurations and the ‘many’ weirs configurations using this approach, namely:  Variants with one weir are not feasible. River works and additional works necessary for adapting the present locks, pumping stations, sluices, etc. are too extensive to be implemented within the set time constraint of 10 years.  A solution with many weirs; many weirs with limited water heads could be present along the Nederrijn and Lek and an upstream weir at the IJsselkop which regulates the discharge distribution over the IJssel and the Nederrijn. The complexity and the scale of the small weirs are lower compared to the present weirs located near Driel, Amerongen, and Hagestein. Theoretically spoken, a solution which consists of many small inexpensive weirs could even be cheaper compared to a solution which is composed of two to three large weirs. However this option is not a suitable solution in the context of this graduation research because this solution is not applicable within the set time constraint of 10 years. Furthermore, disadvantages for commercial shipping and recreational boating are large for this solution than a solution composed of 2 or 3 weirs.

J.2 SECOND BRAINSTORM SESSION

Four system aspects were made to order the next brainstorm session. The weirs are positioned using the experience of the previous brainstorm session. Variants which make no sense like 3 weirs located in between the connection of the Lekkanaal and the Amsterdam-Rijnkanaal are not an option in this brainstorm session. The system aspects used for the elaboration of variants are:

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 closable but open Rijnmond o A weir is located downstream of the village of Lexmond. This weir is part of the new canalization of the Nederrijn and Lek and acts like a flood barrier in case of high river runoff and a storm surge from sea. The flow of the Lek has to be diverted towards the south by the barrier as indicated in Figure J-1. The water depth at this reach has to be equal or larger than 4.2 metres for navigation to sail from Amsterdam towards the Lekkanaal. So, in short: . A weir should be located in between the confluence of the Lek&Noord and the village of Lexmond. . A minimal depth of 4.2 metres must be present at this reach for commercial shipping.

Figure J-1 Lexmond barrier (Wijdenes, 2010)

 commercial shipping o Optimization of the navigational functions of the Nederrijn and Lek by placing the minimal amount of weirs at the optimal locations is the aim of this system aspect. The amount of lock passages of a vessel is minimized by placing the complexes on the right locations, so a solution equipped with one weir on the reach Amsterdam-Rijnkanaal- IJsselkop is a better variant compared to a solution equipped with two weirs on the reach Amsterdam-Rijnkanaal-IJsselkop because vessels have to use one lock less with respect to the present situation. Furthermore, a variant with no weirs on the reach confluence Noord&Lek is a better variant with respect to a variant with weirs on this reach. In short: . Sufficient depth must be present at the Nederrijn and Lek . Ships have to pass as little locks as possible on the main shipping routes.  separation of recreational boating and commercial shipping o The reach Nederrijn-Lek will be downgraded into a local waterway. Commercial shipping has to use the Waal instead of the Nederrijn to travel from east to west and vice versa. Vessels are still able to reach certain harbours but they have to decrease their draught. Currently, the navigation in between the connection of Lekkanaal and Amsterdam-Rijnkanaal is very limited; less than 5 million tons of cargo is transported over this reach with respect to 25-50 million tons of cargo over the Lekkanaal and more than 50 million tons over the Waal. Furthermore, the upstream cargo transport over the Nederrijn is 3 to 5 times larger compared to the downstream cargo transport (based on: (Rijkswaterstaat, 2009) & (Noordhoff Atlasproducties, 2011)). This difference is explained by the flow direction of the Nederrijn. The flow velocity is nearly zero when the weirs of Amerongen, Hagestein, and Driel are closed. The velocity of the Waal is not

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equal to zero, so the vessels have to overcome a counter flow when sailing upstream on the Waal which costs more fuel than sailing in the upstream direction on the Nederrijn. Vessels sailing on the Nederrijn could be directed towards the Waal instead of the Nederrijn for the new situation to minimuse the lock dimensions and to minimise the number of weir complexes. Several options for the separation of recreational boating and commercial shipping are presented in this chapter. However, many ships do use the Lekkanaal and the Amsterdam Rijnkanaal. Therefore, sufficient draught must be present at the crossing of the Amsterdam Rijnkanaal with the Nederrijn and sufficient draught must be present downstream from the Lekkanaal. Separation variants should be eliminated according to requirement FR 7.2). This requirement states that the same kind of navigation has to be able to use the system for the new situation. However, a reduction of building and maintenance costs could be obtained by relocation the commercial shipping from the Nederrijn towards the Waal. It could be possible that the opportunities generated by this option overrule the disadvantage of the relocation of commercial shipping from the Nederrijn to the Waal. This is already been investigated and proposed by Rijkswaterstaat (Havinga, 2012). Therefore these variants are elaborated in further detail, in despite they do not fulfil requirement FR 7.2). So in short:  Commercial shipping is removed from the Nederrijn and Lek  Enough draught must be present at the crossing of the Amsterdam Rijnkanaal and the reach Lekkanaal –confluence Noord&Lek  ecology. o The system is optimized with respect to ecology by placing the right amount of weirs on the correct locations; the amount of fish passages and deviations in ground water levels is minimized by placing the complexes on the right locations. Zero weirs located in the Nederrijn and Lek would be the best solution for the fish migration but is not an option, because a weir is needed at the upstream boundary in order to control the discharge over the Nederrijn and IJssel. The second best option for fish migration is the application of one weir because fish has to pass just one obstacle, but is also not an option as described earlier in this appendix. In short: . As little as possible weir complexes have to be placed in the Nederrijn and Lek for fish migration . Large deviations in ground water level must be avoided.

About 12 options per system aspect are drawn up during the second brain storm session. The complexity of the project area is reduced to the Nederrijn, Lek, and the connected waterways as presented in Figure J-2. Options are generated by drawing weirs and dammed reaches in this figure. Options are organized from few to many weirs as shown in a copy of a brainstorm session in Figure J-3. The main advantages and disadvantages of a system aspect are described in the left upper corner; the advantages and disadvantages per option are presented next to the drawing of specific variant; the planned weirs are indicated as blue V’s which are rotated counter clockwise over an angle of 90O.

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Figure J-2 Basis drawing used in the brainstorm session

J.2.1 PRELIMINARY SELECTION OF VARIANTS

A selection is made for the 47 variants which were developed during the second brainstorm session. Some variants could immediately be eliminated, because they do not meet the elaborated list of requirements described in F.2. Secondly variants which perform worse on certain aspects with respect to other variants were eliminated. For example; variant 5 of Figure J-3 is a better variant compared to variant 6 of Figure J-2 because the upstream weir of variant 5 is situated closer to the bifurcation which results in a better discharge control over the IJssel and Nederrijn. But variant 4 of Figure J-3 is a better variant with respect to variant 5 because navigation is less hampered by the downstream weir. Navigation heading from Rotterdam towards Amsterdam has to pass one weir of the Lekkanaal instead of two weirs as presented in variant 5 of Figure J-3. In this way every system aspect is filtered until a selection of about 3 to 4 best options per system aspect was obtained. Variants of different system aspects which showed much equality are combined during. After reconsideration, the closable but open Rijnmond variants were not as good as previously assumed. The location of the downstream weir near the village of Lexmond has some negative side effects. A minimum average depth of 4 metres is already present from the downstream boundary (confluence Lek and Noord) till the village of Hagestein (65 kilometres from the IJsselkop during low tide and OLR. So the effect of this weir for improving the minimum draught is limited. Furthermore navigation heading from Rotterdam towards Amsterdam and vice versa has to pass an extra lock which results in extra time delay. Based on these arguments, the closable but open Rijnmond variants are eliminated and are not further elaborated. Furthermore, only the options with an upstream weir at the most upstream location are elaborated in further detail. The most upstream weir may not be located too far away from the IJsselkop in order to control the discharge accurate enough over the Nederrijn and IJssel and not too far away from the bifurcation in order to reduce the amount of weir complexes (Rijkswaterstaat, n.d.). So the weir has to be placed as close as possible near the IJsselkop just like weir Driel (13 kilometres away from the IJsselkop). The elimination of variants results in 9 best variants presented in Figure J-4, which are elaborated in further detail in J.3.

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Figure J-3 Impression of the brainstorm session of the system aspect commercial shipping

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Figure J-4 Overview of the remaining variants

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J.3 ELABORATION OF CONFIGURATION VARIANTS

The remaining variants which are shown in Figure J-4 are now elaborated in further detail. First a global scan is performed on the impacts of the variants on its surroundings. Variants are eliminated if the impact on the surroundings is much larger with respect to comparable variants. The remaining variants are elaborated in further detail in J.4.

J.3.1 ELIMINATION OF VARIANTS

The implementation works of variant “2w;Driel&A-R Kanaal;com.ship”4 and variant “2w;Driel&A-R Kanaal;recr” which are shown in Figure J-5 are larger than variant “2w;Driel&Lekkanaal;recr” and variant “2w;Driel&Culemborg;recr.” The tide in combination with the minimum flushing discharge is not able to generate sufficient depth at the downstream side of the Amsterdam-Rijnkanaal; only a depth of 1 metre is present at this location during low tide and OLR conditions. The water level at the downstream side of the second weir of variant “2w;Driel&Lekkanaal;recr” and variant “2w;Driel&Culemborg;recr” are higher so more depth is available and less adaptation works have to be implemented. A solution for increasing the depth could be found by the implementation of river improvement works (dredging etc.). The amount of river works for implementing sufficient draught of variant “2w;Driel&A-R Kanaal;com.ship” and “2w;Driel&Lekkanaal;recr” is larger than variant “2w;Driel&Lekkanaal;recr” and “2w;Driel&Culemborg;recr”, so variant “2w;Driel&Lekkanaal;recr” and “2w;Driel&Culemborg;recr” performs better than variant “2w;Driel&A-R Kanaal;com.ship” and “2w;Driel&Lekkanaal;recr”.

Figure J-5 Elimination of variant “2w;Driel&A-R Kanaal;com.ship” and “2w;Driel&A-R Kanaal;recr”

Variant “4w;Driel&Wageningen&A-R Kanaal&Lekkanaal;com.ship” is also eliminated. A solution consisting of 4 weirs does not have major advantage compared to a solution consisting of 3 weirs because one extra weir has to be constructed. The costs per weir of variant “4w;Driel&Wageningen&A-R Kanaal&Lekkanaal;com.ship” are not significantly lower compared to a solution consisting of 3 weirs which is presented in Figure J-6 (back ground information about the estimation of costs is presented in appendix L). The construction costs of a solution composed of four weirs is 50 million more and an extra weir results in an extra dammed reach with a changed water level, so even more adaptations have to be implemented with respect to a solution consisting of three weirs. Therefore, an option composed of 3 weirs is cheaper with respect to an option composed of 4 weirs and has the preference.

4 The number of weirs is described before the first semicolon, the locations in between the semicolons, and the system aspect after the second semicolon.

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Figure J-6 Estimated costs as a function of the amount of weirs situated in the Nederrijn

J.4 ELABORATION OF CONFIGURATION VARIANTS

Finally, the seven remaining variants are elaborated in further detail. The impact of each variant on its surroundings is determined by using the model described in B.3. Separation variants should be eliminated according to requirement FR 7.2). This requirement states that the same kind of navigation has to be able to use the system in the new situation. However a reduction of building and maintenance costs could be possible by relocation the commercial shipping from the Nederrijn to the Waal. The separation variants are elaborated in further detail to explore the advantages of these variants, in despite they do not fulfil requirement FR 7.2). Every upstream weir of each variant is placed at the upstream side of the Lekkanaal connection, so nothing changes at the reach Lekkanaal-confluence Noord&Lek for each variant. Because of this, river adjustment works located at 67 kilometres or further from the IJsselkop are not taken into account in this elaboration of configuration variants, because no adaptations of the profile takes place with respect to the present situation except the costs of the overdue maintenance. The overdue maintenance costs are taken into account for the quantitative estimation of costs. Furthermore, the new ‘dammed’ reaches do not influence the bridges which cross the Nederrijn and Lek. The variant with the smallest clearance is variant “2w;Driel&Lekkanaal;com.ship,” which is presented in Figure J-7. A clearance of 11 metres is still present for this variant which is still larger compared to the minimal clearance of 9.10 metres. The clearances of other variants are larger with respect to variant “2w;Driel&Lekkanaal;com.ship,” so there is no need for the adjustment of bridges for the new dammed situation for each variant. Thus requirement FR 8.1.1) (The minimum vertical clearance is 9.10m) is fulfilled for every variant. The amount of adaptation works of every variant is presented in Table J-4. This table is used for the ranking of the variants with respect to the implementation costs.

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Figure J-7Variant “2w;Driel&Lekkanaal;com.ship” and the crossing bridges

J.4.1 “2W;DRIEL&LEKKANAAL;COM.SHIP”

Two weirs have to generate sufficient draught for this variant. Weir 2 presented in Figure J-8 is positioned near weir Hagestein and has a maximum head of 5.5 metres. Sufficient depth is available at the downstream side of this weir, so no river improvement has to be implemented near weir two. The dammed reach is 50 kilometres long; which is too long for generating sufficient water levels in between weir 1 and 2. Therefore, adaptation works have to be implemented to generate sufficient draught. The maximum water level in between weir 1 and 2 is equal to +5.00m NAP which is set by the limitation of barrier Ravenswaaij. This limitation is caused by the dimensions of barrier Ravenswaaij and the levees along the Betuwepand of the Amsterdam Rijnkaal. The barrier and the levees are designed for a maximum water level of +5.55 M NAP (Rijkswaterstaat, 2011) and have to be redesigned and adapted for a new situation with higher water levels which is not an option given the time constraint of 10 years. The impact on the surroundings of this variant is presented in Figure J-9, Figure J-10, and Figure J-11 and summarized in Table J-4.

Figure J-8 Two weir variant; commercial shipping on the Nederrijn and Lek and weir 2 placed at the upstream side of the Lekkanaal connection (units: kilometres)

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Figure J-9 “2w;Driel&Lekkanaal;com.ship.” Harbour and river characteristics

According to Figure J-9 the following has to be changed/adapted:  The Nederrijn has to be deepened in between km15 and km30 with 1,5 metres to create sufficient draught.  The Lek has to be deepened in between km65 and km85 with 1,5 metre to create sufficient draught.  Levees have to be raised from km50 and km65 with 1,2 metre to keep the river in the summer bed.  1 large commercial harbour (Wageningen km 25) has to be adapted for a water level decrease of 1 metre.  1 large recreational harbour (Wageningen km25) has to be adapted for a water level decrease of 1 metre.  3 small commercial harbours (km 18, km 31, km 32) have to be adapted for a water level decrease of 1 metre.

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Figure J-10 “2w;Driel&Lekkanaal;com.ship.” Pumping stations and sluices

According to Figure J-10 the following has to be adapted:  1 pumping station has to be adapted for a water level decrease of 1 metre.  2 inlets have to be adapted for a water level decrease of 1 metre.  1 inlet has to be adapted for a water level increase of 2 metres.

Figure J-11 “2w;Driel&Lekkanaal;com.ship.” Lock characteristics

The water level at km48 remains in between the upper and lower boundary according to Figure J-11, so no adaptations have to be implemented.

J.4.2 “2W;DRIEL&CULEMBORG;COM.SHIP”

Two weirs have to generate sufficient depth in this variant for implementing sufficient draught for commercial shipping. Weir 2 has been relocated in the upstream direction with respect to weir 2 of variant “2w;Driel&Lekkanaal;com.ship”, which is presented in Figure J-8. In this way, the length of the reach weir 1-weir 2 is shortened. However, the maximum water level of this reach is still +5.00 metres which is set by

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the limitation of barrier Ravenswaaij and the Betuwepand. The difference with respect to variant “2w;Driel&Lekkanaal;com.ship” is the river improvements at weir 2. The river bed has to be lowered at the downstream side of weir 2 in order to generate sufficient water depth instead of raising the river banks and summer levees. The impact on the surroundings is presented in Figure J-13, Figure J-14, and Figure J-15. The amount of work is given in Table J-4 and is used for the estimation of the costs.

Figure J-12 Two weir variant; commercial shipping on the Nederrijn-Lek and weir 2 placed in between the Lekkanaal and Amsterdam-Rijnkanaal connection (units: kilometres)

Figure J-13 Variant “2w;Driel&Culemborg;com.ship.” Harbour and river characteristics

According to Figure J-13 the following has to be changed/adapted:  The Nederrijn has to be deepened in between km15 and km30 with 1,5 metres to create sufficient draught.

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 The Lek has to be deepened in between km55 and km85 with 2,5 metre to create sufficient draught.  Levees have to be raised from km50 and km55 with 0,5 metre in order to keep the river in the summer bed.  1 Large commercial harbour (Wageningen km25) has to be adapted for a water level decrease of 1 metre.  1 large recreational harbour (Wageningen km25) has to be adapted for a water level decrease of 1 metre.  3 small commercial harbours (km 18, km 31, km 32) have to be adapted for a water level decrease of 1 metre.  1 small recreational harbour (km59) has to be adapted for a water level decrease of 3,6 metres

Figure J-14 Variant “2w;Driel&Culemborg;com.ship.” Pumping stations and inlets

According to Figure J-14 the following has to be adapted:  1 pumping station has to be adapted for a water level decrease of 1 metre.  2 inlets have to be adapted for a water level decrease of 1 metre.  1 inlet has to be adapted for a water level increase of 2 metres.

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Figure J-15 Variant “2w;Driel&Culemborg;com.ship.” Lock characteristics

The water level at km 48 remains in between the upper and lower boundary according to Figure J-15 so no adaptations have to be implemented.

J.4.3 “2W;DRIEL&LEKKANAAL;RECR”

The recreational function and the commercial shipping function of the waterways are split up for this variant. Commercial shipping is not able to sail from weir 1 till 25 kilometres downstream from the IJsselkop. Commercial shipping sailing from the Amsterdam-Rijnkanaal towards Germany and vice versa has to use the Waal instead of the Nederrijn. In this way a solution consisting of two weirs is possible with limited river adaptation works. Furthermore, locks dimensioned for commercial shipping are no longer required. The Lek and the downstream side of the Nederrijn do have a sufficient depth for commercial shipping. The water level at the connection of the Amsterdam-Rijnkanaal is +4.00m NAP, so an open connection between the Nederrijn and the Betuwepand of the Amsterdam-Rijnkanaal is still present during dammed conditions. The disadvantages are the adaptation works of harbours and inlet stations along the Nederrijn and the hindered commercial shipping. Waterway connections with factories and harbours are lost which results in economic damage. Furthermore vessels have to use more fuel for sailing upstream on the Waal than sailing upstream on the Nederrijn. The impact on the surroundings per variant is presented in Figure J-16, Figure J-17, and Figure J-18. The elements which have to be adapted are presented in Table J-4.

Figure J-16 Variant “2w;Driel&Lekkanaal;recr.” Separation of functions and weir two at the upstream side of the Lekkanaal (units: kilometres)

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Figure J-17 Variant “2w;Driel&Lekkanaal;recr.” Harbour and river characteristics

According to Figure J-17 the following has to be changed/adapted:  The Nederrijn has not to be deepened because sufficient draught for recreational boating is present.  The Lek has to be deepened in between km65 and km85 with 1 metre to create sufficient draught for navigation heading from the Lek towards the Lekkanaal.  Levees have to be raised from km60 and km65 with 0,5 metre to keep the river in the summer bed.  1 Large commercial harbour (Wageningen km 25) cannot be used anymore in dammed configuration.  1 large recreational harbour (Wageningen km25) has to be adapted for a water level decrease of 2 metre.  3 small commercial harbours (km 18, km 31, km 32) cannot be used anymore in dammed configuration.

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Figure J-18 Variant “2w;Driel&Lekkanaal;recr.” Pumping stations and sluices

According to Figure J-18 the following has to be adapted:  1 pumping station has to be adapted for a water level decrease of 2 metre.  2 inlets have to be adapted for a water level decrease of 2 metre.  1 inlet has to be adapted for a water level increase of 1 metre.

Figure J-19 Variant “2w;Driel&Lekkanaal;recr.” Lock characteristics

The water level at km 48 remains in between the upper and lower boundary according to Figure J-19 so no adaptations have to be implemented.

J.4.4 “2W;DRIEL&CULEMBORG;RECR”

The recreational and commercial use of the waterway is split up for this variant. The length in between weir 1 and 2 is 10 kilometres shorter compared to the previous variant. Commercial shipping are able to use the Lek til the Lekkanaal connection and are not able to sail at the downstream side of weir 2 due to restricted water levels, so this part of the river has a recreational function. An open connection with the

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Betuwepand is still present in this variant because the water level in between weir 1 and 2 is +4.50m NAP in a dammed situation. The disadvantage is the adaptation works of harbours and inlet stations along the Nederrijn. The water level has dropped 1.5 metres with respect to the present situation and the passages of commercial shipping is hindered. The impact on the surroundings of the new water levels is presented Figure J-20, Figure J-21, and Figure J-22. The elements which have to be adapted are presented in Table J-4.

Figure J-20 Separation of functions and a weir located in between the connection of the Lekkanaal and Amsterdam- Rijnkanaal (units: kilometres)

Figure J-21 Variant “2w;Driel&Culemborg;recr.” Harbour and river characteristics

According to Figure J-21 the following has to be changed/adapted:

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 The Nederrijn has not to be deepened because sufficient draught for recreational boating is present.  The Lek has to be deepened in between km 65 and km 85 with 1 metre to create sufficient draught for navigation sailing from the Lek towards the Lekkanaal.  1 large commercial harbour (Wageningen km 25) cannot be used anymore in dammed configuration.  1 large recreational harbour (Wageningen km25) has to be adapted for a water level decrease of 1,5 metre.  3 small commercial harbours (km 18, km 31, km 32) cannot be used anymore in dammed configuration.  1 small recreational harbour (km58) has to be adapted for a water level decrease of 3,5 metres.

Figure J-22 Variant “2w;Driel&Culemborg;recr.” Pumping stations and sluices

According to Figure J-22 the following has to be adapted:  1 pumping station has to be adapted for a water level decrease of 1,5 metre.  2 inlets have to be adapted for a water level decrease of 1,5 metre.  1 inlet has to be adapted for a water level increase of 1,5 metres.

Figure J-23 Variant “2w;Driel&Culemborg;recr.” Lock characteristics

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The water levels at km 48 remain in between the upper and lower boundary according to Figure J-23 so no adaptations have to be implemented.

J.4.5 “2W;DRIEL&CULEMBORG;COM.SHIP&RECR”

The description of this variant is included in the main report.

J.4.6 “3W;DRIEL&AMERONGEN&HAGESTEIN;COM.SHIP”

Variant six is equal to the present situation. New weirs would be implemented within 200 metres from the present locations as shown in Figure J-24. The impact on the surroundings is limited with respect to the other variants. Still river improvement works have to be executed to generate sufficient depth along the Nederrijn. The impact on the surroundings is presented in Figure J-25, Figure J-26, and Figure J-27. The amount of work of this variant is also listed in Table J-4.

Figure J-24 Variant “3w;Driel&Amerongen&Hagestein&com.ship.” Present situation (units: kilometres)

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Figure J-25 Variant “3w;Driel&Amerongen&Hagestein;com.ship.” Harbour and river characteristics

According to Figure J-25 the following has to be changed/adapted:  The Nederrijn has to be deepened in between km 15 and km 30 with 1,0 metre to create sufficient draught.  The Lek has to be deepened in between km 65 and km 85 with 1,5 metre to create sufficient draught.

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Figure J-26 Variant “3w;Driel&Amerongen&Hagestein;com.ship.” Pumping stations and sluices

Nothing has to be adapted according to Figure J-26.

Figure J-27 Variant “3w;Driel&Amerongen&Hagestein;com.ship.” Lock characteristics

The water level at km 48 remains in between the upper and lower boundary according to Figure J-27 so no adaptations have to be implemented.

J.4.7 “3W;DRIEL&A-R KANAAL&HAGESTEIN;COM.SHIP”

The location of the weir 2 is changed with respect to variant “3w&Driel&Amerongen&Hagestein;com.ship.” The present location of the weir 2 is the upstream side of the Amsterdam-Rijnkanaal. Vessels which are sailing over the Nederrijn from the Amsterdam-Rijnkanaal to Amsterdam have to cross two locks in the dammed situation. One lock passage at the Nederrijn could be avoided by replacing the weir 2 to the downstream side of the Amsterdam-Rijnkanaal which is shown in Figure J-28. The shipping intensity at the downstream side of the Nederrijn is much lower compared to the shipping intensity of the upstream side of the Nederrijn, so a better condition for commercial shipping

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is created in this way. The water levels in between weir 2 and 3 remains the same (+3.00m NAP) in order to reduce adaptation costs. However the length of the reach is increased from 30 kilometres to 35 kilometres. This would not be a problem when the dammed water level would be high enough. But now, only a maximum water level of +5.00 NAP could be created which is caused by the limitation of barrier Ravenswaaij. Therefore more river improvement works have to be implemented at the downstream side of weir 1. The impact on the surroundings of this variant is presented in Figure J-29, Figure J-30, and Figure J-31. The amount of adaptation work is presented in Table J-4.

Figure J-28 Variant “3w;Driel&A-R Kanaal&Hagestein;com.ship.” Relocated middle weir (units: kilometres)

Figure J-29 Variant “3w;Driel&A-R Kanaal&Hagestein;com.ship.” Harbour and river characteristics

According to Figure J-29 the following has to be changed/adapted:

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 The Nederrijn has to be deepened in between km 15 and km 30 with 1,5 metres to create sufficient draught.  The Lek has to be deepened in between km 65 and km 85 with 1 metre to create sufficient draught.  1 Large commercial harbour (Wageningen km 25) has to be adapted for a water level decrease of 1 metre.  1 large recreational harbour (Wageningen km 25) has to be adapted for a water level decrease of 1 metre.  3 small commercial harbours (km 18, km 31, km 32) have to be adapted for a water level decrease of 1 metre.

Figure J-30 Variant “3w;Driel&A-R Kanaal&Hagestein;com.ship.” Pumping stations and sluices

According to Figure J-30 the following has to be adapted:  1 pumping station has to be adapted for a water level decrease of 1 metre.  2 inlets have to be adapted for a water level decrease of 1 metre.  1 inlet has to be adapted for a water level increase of 2 metres.

Figure J-31 Variant “3w;Driel&A-R Kanaal&Hagestein;com.ship.” Lock characteristics

The water level at km 48 remains in between the upper and lower boundary according to Figure J-31 so no adaptations have to be implemented.

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J.4.8 OVERVIEW OF THE AMOUNT OF ADAPTATOIN WORK PER VARIANT

The impact on the surroundings of every variant is analysed and presented in Table J-4. The dredging works are a rough estimation of the amount of work. The maximum river bed levels are generated by shoals and not by a continues river bed with a fixed angle as shown in the cross sections. A more detailed bed profile has to be determined in order to estimate the amount of work in a more accurte manner, but this is beyond the scope of this research project due to the time constraint. So, the estimation of the adaptation works of the river bed is executed with use of the maximum bed levels Furthermore a more detailed bank profile of the river has to be determined in order to estimate the amount of bank reinforcement projects in a more accurate manner. An indication of river bank improvements is made with the previous model; adaptations are necessary when the water level ‘overtops’ the banks. This is just a rough preliminary schematisation.

J.5 MULTI CRITERIA ANALYSIS FOR CONFIGURATION DESIGN

J.5.1 EVALUATION CRITERIA

An overview of design criteria for this project is already given in H.2. Not all the criteria are used for the assessment of the configuration variants. A selection of criteria applicable for this design level is already executed in H.2. The results of this selection are shown in Table J-1

Table J-1 Selection of criteria based on Environmental Impact Assessments

Topic Aspect Criteria Application within this level Location of the most upstream weir Diverting enough water into the Water quantity determines the discharge steering IJssel capacity. Modification of the bed profile Implementation of weirs affects the Morphology and Water and the sediment characteristics bottom profile and the sedimentation dredging within the project area. patterns and dredging works. Pumping stations and sluices have to Drainage Capacity of pumping stations be able to fulfil their functions (with or patterns and sluices without adaptations of the design). Natural The operation of the objects affects Maintaining natura2000 goals protection law the (ground)water level of the natural and areas 1998 areas Nature The more objects are built in the river Fish migration Enabling fish to pass the weirs. bed; the more difficult it is to pass the weir. Daily impact on local The impact of changed water levels on Living environment Social impact communities (jobs, economy, the municipalities along the Nederrijn transport, recreation etc.) and Lek has to be taken into account. Sufficient vertical clearance, keel Commercial Safety and capacity. (distances, clearance, and navigational width shipping currents, etc.) should be present in the new situation. Recreational activities on the Usage functions A changed configuration results in a Recreation river banks, recreational shift/change of recreational activities. boating, fishery. Changed groundwater quality or Agriculture and Ground water quality and quantity results in a changed yield of horticulture quantity. agricultures and horticultures. The adaptations of the present Construction Construction Implementation within 10 years configuration must be implemented time within 10 years from now.

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The aspects ‘water quantity’ and ‘drainage patterns’ which are presented in Table J-1 are not taken into account in the evaluation of the variants. The aspect ‘Water quantity’ has the same impact for every variant because the discharge of the IJssel and Nederrijn is determined by the steering efficiency of the variant and steering efficiency is determined by the location of the most upstream weir which is for each variant equal. Therefore, it is not relevant to use this aspect in the assessment. The aspect ‘drainage patterns’ is not taken into account for the determination of the performances because it is already implemented in the adaptation costs of sluices and pumping stations for each variant. This aspect would be taken into account twice when the performances are also determined with this aspect. The aspect of the topic ‘usage functions’ of Table J-1 are assessed separately, because the performances per aspect are not related to each other; a good solution for commercial shipping is not always a good solution for recreational boating. So, the aspects used in the assessment are:  morphology  nature  living environment  commercial shipping  recreation  agriculture and horticulture  construction (time).

J.5.2 WEIGHT REFERENCE

Two methods are available for the determination of the performances of a variant, namely an unweighted and a weighted assessment. Every aspect is equally important for an unweighted assessment, so no weight reference is specified. A weight reference is specified in a weighted assessment; certain aspects are more important with respect to other aspects. An unweighted assessment is generally executed by engineering firms which has to advice the client about the impacts on the surroundings of distinct options. The output of the assessment is a factual overview of the impacts on its surroundings. A weighted assessment is executed by the client (this would have been Rijkswaterstaat for a real design project). They have to determine the importance of certain aspects with respect to other aspects (Oosterwijk, 2012). For example: the yearly yield of the aspect commercial shipping could be larger with respect to the yearly yield of recreational boating in the opinion of the client, so the weight factor of the aspect commercial shipping is larger than the weight factor of recreational boating. The weight reference is determined in Table J-2. The scores are based on the opinion of the author and have been discussed during interim meetings. The aspect and the criteria which are already described in paragraph J.5.1 are presented in the first and the second column. A score of 1 means that the aspect in the first column is more important than the aspect given in the first row and a score of 0 means that the aspect in the first column is less important than the aspect given in the first row. The cell is made black when the aspect presented in the first column is equal to the aspect on the first row. The aspects are the same in this case, so they are equally important. The scores per aspects are summed up (second last column) and divided by the total (total=21) in order to calculate the weight factor. A value of 1 is obtained by summing the distinct weight factors up as a final check. The higher the weight factor the more important a certain aspect is. A weighted and unweighted assessment is performed to determine the influence of the weight reference on the results. A conclusion is made on basis of the two distinct results.

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Table J-2 Determination of the weight reference

Aspect Criteria

Morphology Nature Living environment Professional navigation Recreation Agriculture and horticulture Construction (time) Sum Weight reference Modification of the bed profile and sediment 0 0 1 0 0 1 2 0,10 Morphology characteristics. Enabling fish migration and maintaining 1 0 0 1 1 0 3 0,14 Nature Natura2000 goals. Living Daily impact on local 1 1 0 0 1 0 3 0,14 environment communities Professional Safety and capacity for 0 1 1 1 1 0 4 0,19 navigation navigation. Recreation Recreational activities. 1 0 1 0 1 1 4 0,19 Agriculture and Ground water quality and 1 0 0 0 0 1 2 0,10 horticulture quantity Construction Implementation within 10 0 1 1 1 0 0 3 0,14 (time) years Total 21 1

J.5.3 RATING OF VARIANTS

The rating of the performances per variant is given for each object in Table J-3. A scale from 1 to 5 is used in the assessment; the lowest score is 1 and the highest score is 5. The total weighted score is calculated by multiplying the weight factor by the score and summing the weighted score per aspect up and the unweighted score is calculated by summing the unweighted scores up. The scores per aspects are explained in the following enumeration:  morphology The worst morphologic aspects are allocated to variant “2w;Driel&Lekkanaal;com.ship” and variant “2w;Driel&Culemborg;com.ship”, because large bed profile changes have to be implemented to guarantee sufficient draught. The changes do not affect the behaviour of the system during dammed conditions, but affect a free flowing river. Sediments will be trapped in the the deepened waterway and the dredged triangle in the bottom profile would deform quicker with respect to the other variants, so extra maintenance dredging works have to be implemented to guarantee a sufficient draught (Havinga, 2012). The other variants do have higher scores because a less deepened bed profile is present, so the impact of these variants is smaller compared to variant “2w;Driel&Lekkanaal;com.ship” and “2w;Driel&Culemborg;com.ship”. The score of variant “2w;Driel&Lekkanaal;recr” and “2w;Driel&Culemborg;recr” is equal to “2w;Driel&Culemborg;com.ship&recr” because no adaptation works have to be implemented.  nature Variants which have the least impacts on the fish migration, the existing Natura2000 areas and polluting effects do have the highest score for the aspect nature. Variants composed of 2 weirs (and 2 fish passages) do have a higher score than variants composed of 3 weirs (and 3 fish passages), because it is easier for fish to migrate through the river. Furthermore variants only available for recreational boating do have a higher score because the impact of recreational boating on the surroundings is assumed to be less compared to large vessels. The score of the 3

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weir variants remains higher than variant “2w;Driel&Lekkanaal;com.ship” because the Natura2000 areas are not affected by these variants.  living environment The variant with the lowest impact on the surroundings has the highest score for this aspect. People are less affected by a non-changed situation compared to a changed situation. Variant “2w;Driel&Lekkanaal;recr” and “2w;Driel&Culemborg;recr” do have the lowest score because many people are affected by the downgrading of the river towards a recreational river.  commercial shipping Variants which do have sufficient draught and the lowest amount of obstacles on the reach do have the highest scores. So, variant “2w;Driel&Lekkanaal;com.ship” and “2w;Driel&Culemborg;com.ship” are rated with a 5 and variant “2w;Driel&Lekkanaal;recr” and “2w;Driel&Culemborg;recr” with a 1. Variant “3w;Driel&Amerongen&Hagestein;com.ship” and “3w;Driel&A-R Kanaal&Hagestein;com.ship’ are composed of 3 weirs, so an extra obstacle is present. So these variants are rated with a 4 or 5.  recreation The variants which perform the best on recreation do have the highest score. Recreation consists of recreational boating, scenery, swimming in lakes, etc. The variants only available for recreational boating have the highest score. Variants available for both recreation and commercial shipping do have a lower score. Variant “3w;Driel;Amerongen;Hagestein;com.ship” and “3w;Driel&A-R Kanaal&Hagestein;com.ship” do have a higher score compared to variant “2w;Driel&Lekkanaal;com.ship” and “2w;Driel:Culemborg;com.ship” because less water level changes are present in the new dammed situation and the present recreational areas are less affected.  agriculture and horticulture Agriculture and horticulture located along the Nederrijn and Lek are negatively affected by lowered ground water levels. These ground water levels are lowered by a lower dammed water level with respect to the present situation. The variant which has the largest negative impact on the water levels has the lowest score. The highest score is assigned to the variant which has the lowest impact on the water levels.  construction time Variants which could faster be realised with respect to other variants do have a higher score. Variant “3w;Driel&Amerongen&Hagestein;com.ship” and “3w:Driel&A-R Kanaal&Hagestein;com.ship” do have the highest score because the impacts on the surroundings are the lowest and weirs could easily be executed nearby the existing structures. The score of the separation variants are lower compared to the other variants because the social impact and juridical procedures be longer compared to the other variants due to the stakeholders’ opinions.

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Table J-3 Determination of the performances

;

Weight

Aspect Criteria Kanaal& R

factor -

; ; ; ;

"2w; Driel&Lekkanaal; com.ship” "2w; Driel&Culemborg; com.ship” "2w; Driel&Lekkanaal recr” "2w Driel&Lekkanaal; recr” "2w Driel&Culemborg; com.ship&recr” "2w Driel&Amerongen & Hagestein; com.ship "2w Driel&A Hagestein; com.ship Modification of the bed Morphology profile and sediment characteristics. 0.10 1 1 5 5 2 3 3 Enabling fish migration Nature and maintaining Natura2000 goals. 0.14 2 2 4 4 3 4 4 Living Daily impact on local environment communities 0.14 3 3 1 1 3 4 4 Professional Safety and capacity for navigation navigation. 0.19 5 5 1 1 3 4 5 Recreation Recreational activities. 0.19 1 1 5 5 4 2 2 Agriculture and Ground water quality horticulture and quantity 0.10 3 3 2 2 3 4 4 Construction Implementation within (time) 10 years 0.14 3 3 2 2 3 5 4

Score incl. WF 2.7 2.7 2.8 2.8 3.1 3.7 3.7 Score exl. WF 18.0 18.0 20.0 20.0 21.0 26.0 26.0

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J.5.4 COSTS

Costs are determined using key ratios (Dutch: kentallen). The costs per variant are calculated and presented in Table J-4 . Key ratios used for the determination of building costs and are presented in the sequential enumeration. Maintenance dredging costs are not expressed in costs but included in the performance of the variants, because it would take too much time to quantify these costs per variant. Furthermore the extra sailing costs for commercial shipping for the separation variants are not expressed in money but are taken into account in the performances; it would take too much time for determining these (economic) costs within this graduation research.  The dredging costs (dredging operation, transport, deposit, and BTW) of a cubic metre of soil are €11.33 (Waterschap Hollandsche Delta, 2007). The maximum river bed levels are generated by local shoals and not by a continues river bed with a fixed angle. Therefore, a more detailed bed profile has to be determined to estimate the amount of work in a more accurate, but this is beyond the scope of this research project due to the time constraint. The present bed level is presently too high; a draught limitation is nowadays present. The maximum bed profile has to be lowered with 1 metre over a length of 10 kilometres in order to match the original design depth at weir Driel, 0,5 metres over a length of 5 kilometres at weir Amerongen, and 1 metre over a length of 1,5 metres over a length of 20 kilometres which can be concluded by comparing Figure A-8 with Figure B-4. So, the dredging costs of variant “3w;Driel&Amerongen&Hagestein;com.ship” presented in Table J-4 are costs for overdue maintenance and not extra river improvements caused by an increased draught. The total overdue maintenance costs are presented in Equation J-1.

∑

Equation J-1 Overdue maintenance costs

In which: Volume = 0.5*h*l*b The costs per section are:

o 7.400.000

o 1.800.000

 The costs of raising river banks and summer levees is estimated on €2.500.000 per kilometre. This value is based on a report of Waterschap Hollandse Delta (Ziel, 2009).  The costs for the adaptation of an in or outlet are €500.000 to €3.000.000 according to Waterschap Hollandse Delta (Ziel, 2009). An average per inlet is used for the estimation of project costs  The costs for adapting a pumping station are estimated to be equal to the estimated costs of an inlet. So, the costs are €1.500.000 per pumping station.  Adaptation costs for harbours are not known. Therefore, the assumptions of van der Ziel are also used in this research. The adaptation costs of a large harbour are assumed to be €3.000.000 and the adaptation costs of a small harbour (a quay wall along the river) €500.000 (Ziel, 2009).  The costs of weirs are estimated using Figure J-6. The building costs of a variant composed of 2 weirs is €193.000.000 and the building costs of a variant composed of 3 weirs is €237.000.000 according to Figure J-6.

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Table J-4 Determination of costs per variant

Adapting Adapting Adapting Adapting Overdue Adapting large small large small Amount Total costs Variant & Costs Dredging maintenance Raising banks Adapting sluices pumping commercial commercial recreational recreational of weirs (mio €) at Hagestein stations harbours harbours harbours harbours No Variant Adapting 3 Adapting 1 Lowering the Adapting 1 Adapting 1 adaptations. Raising the Adapting 3 sluices. Adapting pumping Lowering the bed bed with a harbour for harbour for New water banks with a harbours for 2 sluices for a station for with a maximum maximum of a water a water level is equal “2w; maximum of 1,2 a water level difference of -1 a of 1,5 metres over 1,5 metres level level to the water 2 Driel&Lekkanaal; metres over 15 decrease of metre and 1 sluice difference 15 kilometres over 20 decrease of decrease of level in an com.ship” kilometres 1 metre for a difference of of -1 kilometres 1 metre 1 metre 'open' +2 metres. metres configuration Costs (mio €) 17 22 75 3 1,5 3 0 4,5 1,5 193 320 Variant Lowering the bed with a Adapting 3 Adapting 1 maximum of Raising the Adapting 1 Adapting 1 1 harbour Adapting 3 sluices. Adapting pumping Lowering the bed 2,5 metres banks with a harbour for harbour for has to be harbours for 2 sluices for a station for with a maximum over 30 maximum of 0,5 a water a water adapted for “2w; a water level difference of -1 a of 1,5 metres over kilometres metres over a level level a water level 2 Driel&Culemborg; decrease of metre and 1 sluice difference 15 kilometres instead of length of 5 decrease of decrease of decrease of com.ship” 1 metre for a difference of of -1 1,5 metres kilometres 1 metre 1 metre 3,5 metres. +2 metres. metres over 20 kilometres Costs (mio €) 17 55 25 3 1,5 3 0,5 4,5 1,5 193 304 Variant No Adapting 3 1 harbour 3 harbours Adapting 1 No lowering of the Lowering the Raising the Adapting 1 adaptations sluices. Adapting could not could not be pumping bed necessary, bed with a banks with a harbour for New water 2 sluices for a be used used station for because the depth maximum of maximum of 0,5 a water level is equal difference of -2 “2w; anymore anymore a is sufficient for 1,5 metres metres over a level to the water metres and 1 2 Driel&Lekkanaal; due to due to difference recreational over 20 length of 5 decrease of level in an sluice for a recr” insufficient insufficient of -2 boating kilometres kilometres 2 metres 'open' difference of +1 draught draught metres configuration metre Costs (mio €) 0 22 25 0 0 3 0 4,5 1,5 193 249

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Adapting Adapting Adapting Adapting Overdue Adapting large small large small Amount Total costs Variant & Costs Dredging maintenance Raising banks Adapting sluices pumping commercial commercial recreational recreational of weirs (mio €) at Hagestein stations harbours harbours harbours harbours 1 harbour Adapting 3 Variant 1 harbour Adapting 1 No lowering of the Lowering the could not be Adapting 1 1 harbour sluices. Adapting No adaptations. could not pumping bed necessary, bed with a used harbour for has to be 2 sluices for a Water levels be used station for because the depth maximum of anymore; 2 a water adapted for difference of -1.5 “2w; remain in anymore a is sufficient for 1,5 metres harbours do level a water level metres and 1 2 Driel&Culemborg between the due to difference recreational over 20 have a decrease of decrease of sluice for a Recr” present banks. insufficient of -1,5 boating kilometres draught 1,5 metres 3,5 metres. difference of +1,5 draught metres limitation metre Costs (mio €) 0 22 0 3 1 3 0,5 4,5 1,5 193 229 Adapting 3 Variant Adapting 1 Lowering the Adapting 1 Adapting 1 1 harbour sluices. Adapting No adaptations Adapting 3 pumping Lowering the bed bed with a harbour for harbour for has to be 2 sluices for a Water levels harbours for station for with a maximum maximum of a water a water adapted for difference of -1 “2w; remain in a water level a of 1,5 metres over 1,5 metres level level a water level metres and 1 2 Driel&Culemborg; between the decrease of difference 15 kilometres over 20 decrease of decrease of decrease of sluice for a Com.ship&recr” present banks. 1 metre of -1 kilometres 1 metre 1,5 metres 3,5 metres. difference of +2 metres metre Costs (mio €) 17 22 0 3 1,5 3 0,5 4,5 1,5 193 246 Lowering the bed Variant 6 Lowering the with a maximum bed with a of 1 metre over 10 “3w; maximum of kilometres ------Driel&Amerongen& 1,5 metres 3 (overdue Hagestein; over 20 maintenance at com.ship kilometres Driel) Costs (mio €) 7 22 0 0 0 0 0 4,5 0 237 271 No Adapting 3 Variant No adaptation Adapting 1 Lowering the Adapting 1 Adapting 1 adaptations sluices. Adapting works Adapting 3 pumping Lowering the bed bed with a harbours harbour for New water 2 sluices for a necessary. New harbour for station for “3w; with a maximum maximum of for a water a water level is equal difference of -1 water levels a water level a Driel&A-R Kanaal& of 1,5 metres over 1,5 metres level level to the water metres and 1 3 remain in decrease of difference Hagestein; 15 kilometres over 20 decrease of decrease of level in an sluice for a between the 1 metre of -1 com.ship” kilometres 1 metre 1 metres 'open' difference of +2 present banks. metres configuration metre Costs (mio €) 17 22 0 3 1,5 3 0 4,5 1,5 237 289

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J.5.5 SCORE (PERFORMANCE/COSTS)

Finally the score of each variant is calculated. The score is determined by dividing the performances by the costs. The unweighted score is presented in Figure J-32 and the weighted score is included in the main report. The costs are quantified on the horizontal axis in million euros and the dimensionless performances are presented on the vertical axis. Variants equipped with two weirs do have lower score with respect to variants equipped with three weirs. This is caused by the following reasons:  Variants with two weirs and available for commercial shipping (variant “2w;Driel&Lekkanaal;com.ship” and “2w;Driel&Culemborg;com.ship”) do have a larger impact on the surroundings than the three weir variants. Therefore the score of these variants are lower compared to three weir variants.  Two weir variants which are only (variant “2w;Driel&Lekkanaal;recr.” and “2w;Driel&Culemborg;recr”) or partly (variant “2w;Driel&Culemborg;com.ship&recr”) available for recreation do have a lower score because the (negative) impact on commercial shipping is quite large. Therefore the performances of these variants are lower compared to three weir variants.

The costs of two weir variants without commercial shipping are lower with respect to two weir variants with commercial shipping; this is caused by the extra adaptation works of two weir variants with commercial shipping. The score of the two weir variants without commercial shipping and two weir variants with commercial shipping do not differ much for the weighted and unweighted assessment.

Figure J-32 Cost-performances excluding weighting (the variants are described in Table J-5)

The weighted and unweighted results are analysed following the method presented in B.2. The results of this analysis are presented in Table J-5.

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Table J-5 Result of variant assessment

Variant Result assessment No preference. The performance/cost ratio of this variant is smaller “2w; compared to variant “2w;Driel&Culemborg;recr”, Driel&Lekkanaal; “2w;Driel&Culemborg;com.ship&recr”, and com.ship” 3w;Driel&Amerongen&Hagestein;com.ship. This variant has the highest costs and the lowest performance. No preference. The performance/cost ratio of this variant is smaller “2w; compared to variant “2w;Driel&Culemborg;recr”, Driel&Culemborg; “2w;Driel&Culemborg;com.ship&recr”, and com.ship” “3w;Driel&Amerongen&Hagestein;com.ship. This variant has the final last highest costs and the lowest performance. No preference. A much higher performance could be achieved by spending a “2w; small amount of money in order to implement variant “3w; Driel&Lekkanaal; Driel&Amerongen&Hagestein;com.ship. Furthermore variant recr” “2w;Driel&Culemborg;recr” is cheaper and has the same performance for less money. “2w; Preference. This variant has the lowest costs for the weighted and Driel&Culemborg; unweighted assessment. recr” “2w; Preference. This variant has a higher performance with respect to variant Driel&Culemborg; “2w;Driel&Culemborg;recr” according to the weighted assessment, but com.ship&recr” slightly more costs but the same performance/cost ratio. “3w; Preference. This variant has the highest performance AND the highest Driel&Amerongen& performance/costs ratio for the weighted and unweighted assessment Hagestein; com.ship “3w; No preference. Variant “3w;Driel&Amerongen&Hagestein;com.ship is Driel&A-R Kanaal& cheaper and has a higher performance. Therefore this variant is subordinate Hagestein; to the other variants com.ship

J.6 CONCLUSION

A selection of three best variants is made of the 47 variants which are elaborated in the second brainstorm session. Finally a design choice is made between the three best variants. First a description of the remaining variants is given in J.6.1 and finally a choice is made in the main report

J.6.1 DESCRIPTION OF THE REMAINING VARIANTS

The best variants of the two weir variants with sufficient water depth for recreation, the combination variant, and the three weir variants are:  variant “2w;Driel&Culemborg;recr” o A full separation between recreational boating and commercial shipping is present for this variant. The water levels are maintained by two weirs. The upstream weir is located near weir Driel and controls the water distribution of the Nederrijn and IJssel. The downstream weir is located near the village of Culemborg and controls the water levels of the upstream reach. Maintenance dredging costs are lower when no commercial shipping are present at the Nederrijn and Lek, because a limited draught should now be available and less dredging works have to be implemented.  variant “2w;Driel&Culemborg;com.ship&recr.”

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o A partial separation between navigational boating and commercial shipping is present in this variant. The water levels are maintained by two weirs. The upstream weir is located near Driel and controls the water distribution of the Nederrijn and IJssel. The downstream weir is located near the village of Culemborg and controls the water levels of the upstream reach. The waterway in between the Lekkanaal connection and the Amsterdam-Rijnkanaal connection is only navigable for recreational vessels. Commercial shipping is still able to use the Nederrijn and could still access the harbour of Wageningen and Arnhem. In this way, the social impact on the system is limited.  variant “3w;Driel&Amerongen&Hagestein;com.ship.” o The Nederrijn and Lek are fully accessible for commercial shipping. No draught limitations are present. The water levels are maintained by three weirs which are located within a couple of hundred metres from the existing ones (weir Driel, weir Amerongen, weir Hagestein). Minor adaptations on the river have to be implemented and minor changes are present. The social impact is the lowest for this variant.

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K. Verification of the horizontal water level assumption

The weirs of the Nederrijn and Lek have to generate a sufficient water level for the navigation by damming the river. The water levels generated by the weirs for zero discharge are equal to line 0 presented in Figure K-1. Back water curves are present at the river when the discharge is increasing. An asymptotic adjustment of the water level towards the equilibrium depth is present in a steady non uniform flow. The higher the discharge the more water is discharged by the weirs. The water level lowers from line 1 to line 2 and finally to line 3. At a certain moment, the discharge in the river is large enough to generate a sufficient water depth, so the gates of the weirs could be fully removed from the waterway. In this case line 4 of Figure K-1 is present in the river. The impact of the minimum flushing discharge on the water levels has to be determined in order to verify whether or not the schematisation of horizontal water levels is justified. The schematisation of horizontal water levels could be used when the set up of water at the upstream weir is small (a couple of decimetres) with respect to the length of the dammed reach. First a calculation for line number 4 is made in order to determine the characteristics of the reach Nederrijn-Lek. The characteristics of the Nederrijn and Lek are estimated using literature and maps and the values are verified by calculating the discharge corresponding to line number 4.

Figure K-1 Movable weir with varying discharges and water slopes (Gijt & Toorn, 2011)

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K.1 MINIMUM DISCHARGE

The discharge which has to be present to realise line 4 is calculated by the formula of Chézy which is already described in the literature study. This formula is rearranged to calculate the minimum discharge as a function of the depth as presented in Equation K-1.

⁄

Equation K-1 Chézy formula

In which: C = coefficient of Chézy [m1/2/s] B = width of the waterway [m] h = depth [m]

ib = slope of the bed surface [-]

The value of the coefficient of Chézy is based on the Waqua model used by Vincent Hermeling during his master thesis (Hermeling, 2004); the used value is C=45 m1/2/s. The width of the waterway is estimated in 120 metres (Table C-1). The minimum depth of the waterway for navigational purposes is 4.20 metres as stated in H.2. At last the slope of the bed surface has to be determined. The slope of the Rijn is 10-4 according to the lecture notes (Battjes & Labeur, 2009) which is quite in agreement with the average bottom slope derived from Figure A-19 (1.2*10-4) and the bottom slope in between the IJsselkop and weir Hagestein which is 1.1*10-4. A bottom slope of 1,1*10-4 is used in the calculations because this slope is based on the most actual data available. The minimum discharge present in the Nederrijn for a free flowing river

is: [ ⁄ ⁄ ] [ ] [ ] [ ] [ ] ⁄ ⁄ The gates of weir Driel are fully lifted at a discharge at Lobith of 2350 m3/s and for a discharge in the Nederrijn of 412 m3/. So, the schematization of the Nederrijn is quite in accordance with the present dam regime; the assumed characteristics of the river are estimated well and could be used in the determination of the length and water levels of a dammed section.

K.2 LENGTH AND WATER LEVELS OF A DAMMED SECTION

Minimal water levels in between the weirs are generated by zero discharge. In this situation the water levels are horizontal. A zero discharge in the Nederrijn does not occur in the present and new situation because a minimum discharge of 25 m3/s should always be present. This minimum discharge generates a back water curve in the waterway. The water level will adjust asymptotically towards the normal depth. The normal depth is calculated with Equation K-2 . ⁄

( )

Equation K-2 Equilibrium depth according to Chézy (Vriend, et al., 2011)

In which:

he = normal depth [m] q = specific discharge [m2/s]

C = coefficient of Chézy [m1/2/s]

ib = slope of the bed surface [-]

The specific discharge is calculated with Equation K-3.

⁄

Equation K-3 Specific discharge

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In which: Q = discharge [m3/s] B = width [m]

So, the equilibrium or normal depth of the Nederrijn during a flushing discharge of 25 m3/s is:

⁄

[ ⁄ ] ⁄ ( [ ])

[ ] [ ] ⁄

( ) The upstream depth is calculated using the approximation of Bresse which is presented in Equation K-4.

⁄ ( )

Equation K-4 The approximation of Bresse (Vriend, et al., 2011)

In which:

he = normal equilibrium depth [m]

h0 = boundary condition depth [m] x = x coordinate of the river [m]

x0 = x coordinate of the boundary condition [m]

L1/2 = half-length given by Equation A-4. ⁄

⁄ ( )

Equation K-5 The 'half-length'

The reach Driel-Amerongen is used to determine the impact of the minimum flushing discharge on the upstream water level. The length of this reach is 30 kilometres; the water levels are calculated at 10 kilometres, 20 kilometres, 30 kilometres from weir Amerongen. The water level at weir Amerongen is 7.2 metres (based on: (Til, 1961)) and the minimum water depth is 4.2 which is given in Figure K-2. The impact of the back water curve is limited on the water level at the end of the reach as presented in Figure K-2, so it is justified to use the horizontal water level approach for designing a new weir configuration in the Nederrijn.

Figure K-2 Water levels for zero discharge and minimum discharge upstream from Amerongen

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L. Cost estimation of weirs

The estimated costs presented in Figure J-6 are based on the dimensions and the costs of already built weirs which are presented in Table L-1. The values and the costs (which are recalculated to present construction costs) originates from ‘Kostenraming van hoogwaterkeringen op basis van een kental´ (Toorn, sd) and ‘Movable barrier for the 21st century’ (Ziel, 2009).

Table L-1 Dimensions and costs per weir or barrier

Width Gate height Head Costs B2hH Object [m] [m] [m] [million €] [m4] Hartel barrier 170 9,3 5,5 294 1478235 Ramspol barrier 240 9 3,2 173 1658880 Ems barrier 360 8,5 3,8 464 4186080 Nakdong river 200 10 2 125 800000 Kromme Nol 50 8,5 2 39 42500 Hollandsche IJssel 81,2 11,5 2 97 151649 Nederrijn (old weirs) 108 7,5 3 62 262440

The key ratios (Dutch: kentallen) are expressed in the parameter B2hH in which B is the width of the weir, h the height of the weir, and H the head over the weir. The best fit was found by using the square of the width of the weir. This parameter is also used for estimating the gate weights in a pre-design (Erbisti, 2004). An explanation for the square root of the width is the maximum bending moment in the gate generated by the water head for which the weir has to be designed. The bending moment is a function of the load and the width of the gate as given in Equation L-1.

Equation L-1 Bending moment

In which: M = bending moment [kNm] q = load [kN/m] l = width [m]

The head H over the Nederrijn and Lek is roughly 9 metres (Length of the section is 90km; the slope of the bed is 10-4 so the head is 9 metres). The width of the structures is estimated at 100 metres, and the height of a gate is calculated with Equation L-2.

Equation L-2 Estimation of the gate height

In which: h = height of a gate [m] d = minimal water depth (4.2 metres) [m]

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H = water head over the Nederrijn and Lek [m] n = amount of weirs [-]

The key ratios per weir [B2hH] are calculated for 1 weir to 6 weirs and the total construction costs of each configuration are determined. The results are shown in Figure J-6. A linear ratio between costs and amount of weirs has been found. A linear relation is present when the fixed costs are relatively high and the water depth is large compared to the total water level difference (Gijt & Toorn, 2011).

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M. Weir location

The first step in the design process of the weir is the determination of the building location. Some requirements for the location are already described in F.3. The requirements are repeated for convenience in the following enumeration:  The structure should be situated at a straight section in order to provide a straight line of sight with the structure navigational openings to facilitate the incoming vessel to enter and exit the weir without the need to make sharp turns. (IR 4) o A straight channel section should be present for at least 1000 metres.  The structure should be situated in a position that minimizes the cross current in the area where vessels navigate. (IR 5)  The site selection should include an evaluation of the existing geological conditions in order to minimize the foundation works. (PIANC, 2006)  Easy access to the site during the construction period must be available. (RR 2)  The structure should be built at a site where with a natural restriction of the river. (RR 3)  The type of structure should take into consideration space that is available in the area.  The structure must be designed according to the Dutch legislation and the building codes (AR 5 and AR 6)

Seven potential locations are available in between the Amsterdam Rijnkanaal and the village of Culemborg. The figure in which the locations are included is included in the main report. Each location is assessed in the sequential sections.

M.1 LOCATION 1; WEIR NEAR THE AMSTERDAM RIJNKANAAL

The approach channel of the reach which has to be excavated is connected to the crossing of the Amsterdam-Rijnkanaal which results into extra adaptation works with respect to other locations and a changed nautical situation at the crossing. Moreover, this location is located 8 kilometres upstream from the design location which is determined in the configuration design synthesis. This shift to the upstream direction results in extra dredging works to generate sufficient draught for recreation. This results in extra costs compared to the other locations. So, location 1 is less favourable with respect to the other locations and is not an option for the construction of the weir.

M.2 LOCATION 2; WEIR AT THE FLOODPLAINS OF BEUSICHEM (FIGURE M-1)

The reach to be excavated of location 2 is much larger with respect to location 4 and location 7, so more river works have to be implemented which results in higher costs. Furthermore a large farm, a large horse breeding, one campsite, a recreational harbour, and the ferry connection are lost or hampered. So, location 2 is less favourable with respect to the other locations and is not an option for the construction of the weir.

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Figure M-1 Cultivated floodplains of Beusichem (left) and the Lek and ferry connection (right)

M.3 LOCATION 3; WEIR IN THE WATERWAY IN BETWEEN CULEMBORG AND BEUSICHEM

A nearly straight reach is available at location number 3 as presented in Figure M-2, so requirement IR 8) is fulfilled. Furthermore, a local constriction is available at this location so requirement RR 5) is fulfilled. The weir could be built directly in the river bed or could be first constructed at the floodplains and floated into the river and being submerged at the final location. The soil conditions at this location are good as presented in the area analysis. This cross section has been made from the north-east towards south-west direction. The section from 0,1 kilometres till 0,75 kilometres of Figure M-3 represents the flood plains at the north side of the Lek and the section from 0,75 till 1,1 represents the river. A Holocene clay layer is present till a depth of -6 m NAP and sand layers till a great depth. So, a weir could be constructed at location 3 which makes it a feasible location.

Figure M-2 Waterway in between the village of Culemborg and Beusichem

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Figure M-3 Soil layers (DINOloket)

M.4 LOCATION 4; WEIR AT THE FLOODPLAINS IN BETWEEN CULEMBORG AND BEUSICHEM

A straight reach could be excavated in the floodplains (presented in Figure M-4) for the new weir. The weir complex could be constructed ‘in the dry’ at the floodplains. Sufficient space is available for the weir at the floodplains. The lock and fish passage could be realised in the present river bed of the Lek. The soil conditions at this location are equal to location 3 and are presented in Figure M-3. The amount of soil to be excavated for the construction of the reach is presented in Equation M-1. The reach would be 2000 metres long (the length of an approach channel of a weir is 1000 metres), 130 metres wide (present river width) and 7 metres deep (floodplains at +5.00m NAP and the river bed at -2m NAP). [ ] [ ] [ ] [ ]

Equation M-1 calculation of volume of soil to be excavated

The soil could be sold to reduce the project costs or could be dumped in the sand pit to reduce the ground water seepage underneath the levees caused by the sand pit (Bewoners aan de Lek, 2012). Concluding, a weir could be constructed at location 4 which makes it a feasible location.

Figure M-4 Floodplains in between the village of Culemborg and Beusichem

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M.5 LOCATION 5; WEIR LOCATED IN THE SAND PIT

The weir could be constructed in the sand pit at the bend near Culemborg which is presented in Figure M-5. The sand pit could be transformed into a building pit when levees are constructed which surrounds the sandpit and the sandpit is pumped dry. After constructing the weir, a straight reach could be realised by excavating the banks of the sandpit. The depth of the sandpit is about 30 metres and the bottom of the sandpit is located at -25m NAP (Reverda, 2012). The weir has to be founded into the subsoil to transfer the forces towards the subsoil. The distance of the river bed to the bottom of the substructure is 10 metres maximum, which is based on reference weirs presented in Table M-1. The sill of the weir is located around -2m NAP, so the deepest part of the foundation is located at -12m NAP. A minimum layer of sand of 13 metres has to be added to the present bed level. The sandpit has a surface of about 200.000 m2 (measured in Google maps), which results in a volume of 2.5 million cubic metre of sand (200.000m2 x(25m NAP-12m NAP)) to realise a bottom depth of -12m NAP. After the construction of the weir, the river bed has to be raised to -2m NAP in order to connect to the other down and upstream river bed profile. So 2 million cubic metre of sand extra (200.000m2 x(12m NAP-2m NAP)) is needed. 4.5 million cubic metres of sand in total have to be purchased which makes this option more costly with respect to location 4. This results in large costs which makes this option not feasible.

Figure M-5 Sand pit near the village of Culemborg

Table M-1 Foundation dimensions

Distance of river bed- Weir Weir type Foundation type bottom of the weir [m] Amerongen Visor gate Founded on the subsoil 10 Braddock Radial gate Piled foundation 1 Sauer Barrier Flap gate Founded on the subsoil 2 Bremen Weser Weir Flap gate partly piled foundation 7 Ipotesti Dam Radial gate Founded on the subsoil 5

M.6 LOCATION 6; WEIR LOCATED AT THE FLOODPLAINS OF CULEMBORG

The weir is located at the floodplains near the village of Culemborg. A new reach has to be excavated which runs underneath the rail way bridge which is presented in Figure C-13. Navigation is not possible without adaptations of the rail way bridge due to the restricted height as presented in Figure M-6. Furthermore, some pylons of the railway bridge would be located in the new reach which is disadvantageous for navigation. So, this location is not an option.

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Figure M-6 Railway bridge of Culemborg

M.7 LOCATION 7; WEIR LOCATED IN THE RIVER AT CULEMBORG.

The weir is located in a bend when it would be built in the river bed near Culemborg. This bend and the railway bridge hamper the straight line of sight for the shippers as presented in Figure M-7. So this location is not an option according to IR 8).

Figure M-7 Railway bridge of Culemborg and the ‘harbour’ and ferry of Culemborg

M.8 CONCLUSION

A weir complex consists of two parallel channels, namely one channel for the weir and one channel for the locks and if necessary or profitable an energy plant which can be used for the accurate discharge control and minimum discharge. Normally, a new design loop has to be implemented in order to determine the best configuration of the lock channel and the weir channel. However, one configuration is chosen without a new design loop due to the tight planning of the graduation research. Performing an extra design loop would take too much time. Therefore a configuration is chosen by the author without a detailed investigation. The chosen configuration is described in the main report.

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N. Preliminary weir design

The weir is composed of several sub-systems as described in the functional analysis. A start of the structural design of these sub-objects is made in this appendix. First a general description is given for the weir design. Secondly, the width of the weir is determined with use of the minimal dimensions of the gaps and the operational boundaries of the weir. Subsequently weir gate types are described which could be used for the weir. An evaluation has been performed in order to determine the best weir type. Furthermore, a conclusion is drawn about the accurate discharge control.

N.1 GENERAL WEIR DESIGN

A weir consists of numerous objects which are needed to dam the river. The weir has to anticipate at the varying conditions of the surrounding like seasonal variation, variation in water use, and changed water levels. The main objects of a weir, which are indicated in Figure N-1, are (PIANC, 2006) & (Gijt & Toorn, 2011):  the inlet structure o The inlet structure has to defend the bottom for erosion at the upstream side of the structure. Therefore, a bottom protection is placed which has to withstand the flow and has to protect the river bed from erosion. The weir could become unstable and could collapse when the scour holes in the vicinity of the structure become too large.  the substructure o The substructure provides a stable support for the entire structure. The loads at weir have to be transferred via the substructure to the subsoil. The substructure has also to provide resistance to seepage, preventing excessive water losses, and degradation of soil located underneath the substructure.  the superstructure o The superstructure is composed of piers and abutments. The piers and abutments support the gates, bulkheads, gate operation machinery, etc. and have to transfer the loads to the substructure.  the gate o The gates are the water retaining element of the weir complex. The gate(s) have to withstand the water flow and have to regulate the discharge. The gate(s) transfer the hydraulic forces to the superstructure.  the outlet structure. o The outlet structure has to defend the bottom fort erosion at the downstream side of the structure. Furthermore a stilling basing could be located at the outlet structure which has to ‘absorb’ the energy of the water flowing through the gate.

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Figure N-1 Main weir objects (based on: (Gijt & Toorn, 2011))

The substructure could be realised in four different manners as indicated in Figure N-2. The main advantages per option are (PIANC, 2006): a) The construction of the substructure is in one stage. A homogeneous foundation is required which can be built on site or could be floated in and being submerged on site. The concrete works could be executed in one phase or spread over several phases. b) The main piers are constructed separately from the sills. The execution works of the weirs could take place in several phases. In this way, flexibility in schedule of building process is created. Furthermore, this option requires a minimal of reinforcement steel due to the minimal bending moments. The main disadvantage of this option is differential settlement of the piers which could hamper the gates operation. c) The substructures of the weir are separated by expansion joints. Each gate is constructed in a ‘separate’ structure. The main advantage of this system is the safe operation of the gates because the risk of differential settlements is not present. Therefore this system is suitable for weak soils. The objects could also being floated in and submerged like option a). The disadvantages of this solution are: an increased amount of reinforcement with respect to option b) due to the larger bending moments of the structure. Furthermore the total with of the river is reduced due to the two piers located in between each opening. d) This option combines option b) and c). A section of the weir is erected from independent sills and piers and a section of the weir is erected using a monolithic structure.

Figure N-2 Weir foundations (PIANC, 2006)

Several gate types could be placed in between the openings of the weir. Every gate type has its own advantages and disadvantages. In general three distinct principle solutions are available to create a movable closing element namely (Gijt & Toorn, 2011):  translation o vertically o horizontally  rotation

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o horizontal axis o vertical axis  translation and rotation.

N.2 GATES

Several gate types and their operation are summed up in this chapter. The gate description is based on (Erbisti, 2004)& (PIANC, 2006).

N.2.1 HORIZONTAL ROLLER GATE

The roller gate is pushed or towed in its position from a recess at the river banks during high waters. The disadvantages of this gate are the space needed at the floodplains. Furthermore, the crest level of the roller gate is not adjustable, so the discharge cannot be controlled by the gate itself but by a extra structure which makes the complex more costly. Horizontal roller gates are generally not used for weirs but for locks, so they are not an option for weir Culemborg

N.2.2 SECTOR GATE (HORIZONTAL AXIS)

A sector gate is composed of a curved skin plate and a watertight downstream seal as presented in the left figure of Figure N-3. The sector gate is kept in open position by the water pressure on the inner face of the upstream side of the gate. The gate is hinged to the sill at the downstream side at every 1,5 to 3 metres. Debris and ice are able to flow over the gate and are not being stored at the upstream side of the weir like an underflow weir. Sector gates can be made as long as possible because the forces are being transferred to the ground per running metre. However, it is not possible to control a varying water level using a sector gate. Therefore, a sector gate is not an option for this project

Figure N-3 Sector gate with horizontal axis. Left: drawing (Erbisti, 2004); right: application of a segment gate (source: Google Maps)

N.2.3 SEGMENT/RADIAL GATE

A segment gate consists of a curved skin plate supported by radial compressed arms which are fixed to the substructure which is presented in Figure N-4. The gates are rotating around a horizontal axis. Segment gates could also being positioned in a reversed manner. The main advantage of placing the gates in a revered manner is the introduction of loads into the piers;. The loads are transferred to the piers by compression for a reversed segmet gate which is advantageous for the design. The segment gate has to be lifted to let water pass the weir. The gates have to be lifted over a small distance in order to generate large

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discharges. These gates impose a height restriction for the waterway because the hinge of the gate or the gate itself forms an obstruction. Furthermore floating debris and ice are not able to pass the weir when it is (partly) closed. An extra flap gate could be realised on top of the segment gate in order to let floating debris and ice pass the weir, but this is more expensive and needs more maintenance with respect to a normal segment gate. Vessels have to pass the weir according to requirement FR 10), so a weir equipped with a segment/radial gate is not an option.

Figure N-4 Left: cross section of a segment gate (Erbisti, 2004)Right: Haringvliet segment gate (source: http://de- haringvliet.blogspot.nl/)

N.2.4 SUBMERGED SEGMENT/RADIAL GATE

A variation of the above described segment gate is the submersible segment gate. A well-known hydraulic structure equipped with a submerged segment gate is the Thames barrier in London which is presented in Figure N-5. The gate is located at the bottom for high discharges, so navigation is not hindered by a height restriction. The gate is lifted above water level to perform maintenance. The water flows over the gate so floating debris and ice is able to freely flow over the gate but sediments are being trapped behind the gate. The flow of sediments over the gate could impose wear of the gate, so it has to be regularly inspected in order to determine the impact of the wear.

Figure N-5 Thames barrier (source: www.bbc.co.uk/london)

N.2.5 VERTICAL LIFTING GATE

A vertical lifting gate is lowered into the waterway to dam the river as presented in Figure N-6. The gate is positioned by mechanical devises located in the lifting towers. The gate is lowered by cables when the self-

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weight of the gate is larger than the buoyancy, and the gate is lowered by pistons when the self-weight of the gate is smaller than to the buoyancy. The gate is fixed onto a track located in the towers by wheels or a sliding track in order to avoid large friction forces which could damage the gate or could result in blockages of the gate. The gate will be jammed when one side of the gate decreases more slowly with respect to the other side. Generally one side of the gate is covered by a flat skin plate which is stiffened by girders at the backside in order to retain the water. For large spans trusses can be used in order to generate enough strength and stiffness.

Figure N-6 Hartel Canal Barrier

N.2.6 VERTICAL LOWERING GATE

The gate of a weir equipped with vertical lowering gates is lowered into the foundation. The gate is fully pushed up when the river is fully dammed and the gate is partly pushed up when the river is partly dammed. The drawback of this weir type is the sedimentation on top of the gate when the gate has to be lifted. The sediments could block the gate or could result into larger forces in order to push the gate into position. Another drawback is the vertical dimension of the foundation. The gate has to be at least 7,2 metres high, so the foundation should be in the order of 10 metres high which result in higher cost with respect to a more shallow foundation type. Therefore a sector gate or flap gate is a better solution with respect to the vertical lowering gate. So the vertical lowering gate is not an option

N.2.7 FLAP GATE (MECHANICALLY OPERATED)

A flap gate consists of a straight or curved retaining surface which is connected by hinges to the sill foundation. The gate is lifted by pivots located underneath or at the sides of the gate. Flaps are stored submerged and flat to the bottom, so no height restrictions for navigation are present during an ‘open’ river. The disadvantages of the flap gate are the sedimentation at the back side of the flap when the gate is opened and the difficult inspection. The sediments could block the operating mechanisms which is located underneath the gate.

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Figure N-7 Flap gate (source: http://www.cmchydro.es)

N.2.8 FLAP GATE (HYDRAULICALLY OPERATED)

A hydraulically operated flap gate also consists of a straight or curved retaining surface which is connected to the sill. The gate is raised by the buoyancy of the gate and floats into position by reducing the gate weight (for example by pumping out water). The hydraulic operated flap gate is not suitable for the design of weir Culemborg because this design is not able to control the discharge.

N.2.9 DRUM GATE

A drum gate is a horizontal floating vessel, triangular shaped and hinged along the lower upstream edge as presented in Figure N-8. The gate is placed into position by the hydraulic pressure; the pressure underneath the gate generates a moment over the hinges. The gate is held in the maximum position by a flap at the downstream side of the gate. The maximum head controlled by a drum gate should be limited to 4 metres because of the rapid increasing space for the hydrostatic chamber. Furthermore, it is not possible to control a varying water level using a drum gate. The sector gate follows nearly the same principle as the drum gate however it is able to resist a higher head (Erbisti, 2004). Therefore, the sector gate is a better option with respect to the drum gate and a drum gate is not an option for the Culemborg weir.

Figure N-8 Drum gate (Erbisti, 2004)

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N.2.10 INFLATABLES

Inflatables, like the Ramspol barrier presented in Figure N-9, are inflated when needed. Water flows over the weir when the water level at the upstream side becomes higher with respect to the crest level. Discharge and water level regulation of the upstream ‘pool’ is not possible due to the ‘soft’ character of the weir: The water flows at the mid-section of the gate when the intern pressure is decreased; so, a clear control point is not present. Therefore an inflatable weir is not an option and could not be implemented in the Culemborg weir.

Figure N-9 Balgstuw Ramspol (Volkskrant, 2012)

N.2.11 VISOR GATE

The operation of visor gate is already widely discussed in the literature study presented in appendix A.

N.3 VARIANTS

Three major design aspects are available for the weir design, namely the amount of openings, the accurate discharge control, and the gate type. The total width of the opening should be larger than 100 metres and smaller than 140 metres which results from the hydraulic model. The width of a single opening must be at least 29 metres for one lane commercial shipping and 41 metres for two lane commercial shipping. These widths are being used as ‘building stones’ for realising a width which is larger than 100 metres and smaller than 140 metres. Gates must be placed in these openings in order to control the water levels. Several gate types can be applied for weir Culemborg. A preliminary selection is already made in N.2. The use of an accurate discharge control depends on the gate types. An accurate discharge control is necessary when the weir is equipped with underflow gates and no accurate discharge control is needed when overflow gates are being used. A filtering of building stones based on the requirements and the results of the hydraulic model is presented in Table N-1.

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Table N-1 Morphological map of weir design (unfiltered)

Design Options Review Check aspect One large Does meet the requirements if the total width is smaller

opening than 140 metres and larger than 100 metres. 2x 29m Opening is smaller than 100 metres. 3x 29m Opening is smaller than 100 metres. Total width is larger than 100 metres. However this width is 3x 34m not desired because the weir design is based on the design limits. Therefore a larger width is more preferable. 4x 29m Does meet the opening requirement.

Number 5x 29m Opening is larger than 140 metres. of 2x 41m Opening is smaller than 100 metres openings 3x 41m Does meet the opening requirement. 4x 41m Opening is larger than 140 metres. Does meet the requirements when the total width is smaller than 140 metres and larger than 100 metres. But is 3x 29m+1x 41m less advantageous due to different gate sizes. Money can be saved by using equal spans. Total width is equals 140 metres. This size is not desired 2x 29m+3x 41m because the weir design is based on the operational limits. Therefore a smaller width is more preferable. 1x 29m+3x 41m Opening is larger than 140 metres. Using overflow Accurate Able to control the water levels and accurate discharge. gates discharge Using underflow Too sensitivity for small deviations of crest height so not control gates applicable. Horizontal roller Not able to regulate the discharge over the weir. gate Sector gate Not able to regulate the discharge over the weir. Segment gate Significant clearance limitation for navigation. Submerged Able to control discharges and could be lifted out of the

segment gate water for maintenance. Vertical lifting Able to control the discharges and easy to perform

gate maintenance. Vertical lowering Large foundation with respect to other variants, so not the

Gate type gate best option. Flap gate (hydraulically Not able to regulate the discharge over the weir. operated) Flap gate (mechanically Able to control the water levels. operated) Drum gate Not able to regulate the discharge over the weir. Inflatables Not able to regulate the discharge over the weir. Visor gate Able to control the discharges just like the present weirs.

The result of the filtering of the morphological map and the variatns are included in the main report.

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N.4 MULTI CRITERIA ANALYSIS OF THE VARIANTS

N.4.1 WEIGHT REFERENCE

Two methods are available for determining the performance of a variant, namely an unweighted and a weighted assessment. Every aspect is equally important for an unweighted assessment, so no weight reference is specified. A weight reference is used in a weighted analysis; certain aspects are more important with respect to other aspects. The determination of the weight reference is already explained in B.2. The result for this design level is presented in Table N-2. A weighted and an unweighted assessment are performed in order to determine the influence of the weight reference on the results. A conclusion is made on basis of the two distinct results.

Table N-2 Weight reference for weir design

Aspect

Reliability Operation Navigation Maintenance impact Social Sum Outcome weights Reliability 1 1 1 1 4 0.40 Operation 0 1 1 1 3 0.30 Navigation 0 0 0 1 1 0.10 Maintenance 0 0 1 0 1 0.10 Social Impact 0 0 0 1 1 0.10 10 1.00

N.4.2 RATING OF VARIANTS

The rating of the performance per variant is given per aspect in Table N-3. A scale from 1 to 5 is used in the assessment. The total weighted score is calculated by multiplying the weight factor with the score and summing the weighted score per aspect up and the unweighted score is calculated by summing the unweighted scores up. The allocation of the scores per aspects is explained in the following enumeration:  Reliability A variant which is less sensitive for malfunctions, and human errors, less vulnerable for foundation distortions, vibrations, sediments, ice and debris does have the highest score for this aspect. The 120m wide flap gate weir does have the lowest score, because the weir is not able to control the river when one gate fails. For the other weirs, one opening could temporally being blocked when one gate is broken in order to dam the river. The disadvantage of overflow gates are the blocking of sediments and the disadvantages of underflow gates is the blockage of debris and ice. Both types of gates do have the same disadvantages and do have the same score. Vertical lifting gates and visor gates are the most sensitive for foundation distortions. One pylon which settles more than the other pylon can lead into jamming of the gates during lifting operation. Therefore the vertical lifting gate weir has a lower score with respect to the segment gate. The reliability of flap gates is even lower with respect to the vertical lifting gate because the moving elements are placed below water level (Dircke, et al., 2011).  Operation

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A variant which is able to control the river well, is easy to operate, and is limited unavailable for operation during maintenance has a high score for this aspect. A variant equipped with overflow gate is better able to control the river with respect to a gate with underflow gates. Therefore the flap gate weir and the segment gate do have a higher score with respect to the vertical lifting gate and the visor gate. One opening of the weir could be blocked in order to minimise the flow speed or to drain the opening for creating a dry working environment when a gate has to be maintained. Blocking one of the 4 gaps of 29m wide has a lower impact on the operation than blocking one of the 3 gaps of 41m. Therefore the variants with 29m wide gaps do have a higher score with respect to the weirs with 41m wide gaps.  Navigation The ease of passing the weir for navigation is determined by this aspect. A large opening without clearance restrictions does have the highest score. A small opening with clearance restriction does have a low score. Therefore the 120m flap gate weir has a score of 5 and the 29m wide visor gate weir and vertical lifting gate weir a score of 1.  Maintenance A variant with a high score for maintenance has a good maintainability of all details, access to maintenance sensible components, and maintainability during operational conditions. The gates of the submerged segment gate and the vertical lifting gate can be lifted out of the water and are therefore easy to inspect and maintain. Also visor gates can be lifted out of the water, but are harder to maintain due to the cylindrical shape. Also the pivots of the gate are located near or at water surface and are therefore hard to inspect and maintain. Therefore the visor gates have a lower score for maintenance. The flap gates do have the lowest score for maintenance. The hinges are located below water level and therefore hard to inspect during normal conditions. In order to inspect and maintain the hinges in a dry, bulkheads have to be used. Therefore the 41m and 29m gates do have a score of 2. Bulkheads cannot be placed for a weir with on span, so maintenance for this weir is even harder. Therefore the 120m gate does have a score of 1.  Social impact The social impact can be noise, visual hindrance of large lifting towers, or visual hindrance of the waterway. The variant with the least social impact is a weir without mid pylons; this variant has the score ‘5.’ The variant with the highest and the most lifting towers and pylons has the largest impact on the surroundings and has therefore the score ‘1.’ The variant with the most impact are the vertical lifting gate with 4 openings and the visor gate with 4 openings. The submerged segment gate does have only mid pylons and no towers. Therefore this variant has the score 3 for 4 openings and 4 for 3 openings.

Table N-3 Rating of weir variants

Submerged Vertical lifting Visor gates Flap gates

segment gate gates

Aspect Weight factor 1m

3x41m 3x41m 3x4

3x 41m 3x 29m 4x 29m 4x 29m 4x 29m 4x 4x 120m 4x Reliability 0.40 4 4 3 3 4 4 2 3 1 Operation 0.30 4 5 2 3 2 3 2 2 1 Navigation 0.10 4 3 2 1 2 1 4 3 5 Maintenance 0.10 4 3 4 3 3 2 2 1 1 Social impact 0.10 4 3 2 1 2 1 4 3 5 Score incl.WF 4.0 4.0 2.6 2.6 2.9 2.9 2.4 2.5 1.8 Score exl. WF 20.0 18.0 13.0 11.0 13.0 11.0 14.0 12.0 13.0

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N.4.3 COSTS

Costs of the variants are determined using reference barriers/weirs. The used reference structures are the Hartel barrier equipped with vertical lifting gates, the Thames barrier equipped with submerged segment gates, and the flap gates designed by T. Wijdenes in the context of his graduation (Wijdenes, 2010). The flap gate barrier located in Venice could also be used instead of the flap gates designed by T. Wijdenes; however the dimensions and the design conditions differ too much with respect to the Netherlands. A cost reduction of 2.5% could be obtained by implementing visor gates instead of vertical lifting gates. f500.000,- per weir complex was saved of the total costs per weir complex of f20.000.000,- by implementing visor gates for the present weirs (price level 1955) (based on: (Rijkswaterstaat Oost-Nederland, 2005)& (Rijkswaterstaat, 1955)). The costs of the Hartel and Thames barrier originate from ‘Kostenraming van hoogwaterkeringen op basis van een kental´ (Toorn, sd). A key ratio is obtained by dividing the total costs by the height, head and width of the weir of which the result is presented in Table N-4.

Table N-4 Determination of unit costs

Net present height head width BhH key ratio Weir costs [M€] [m] [m] [m] [m3] [M€/m3] Hartel barrier 294 9.3 5.5 170 8695.5 0.034 Thames barrier 2600 7.2 17 530 64872 0.040 97.5% of costs of Hartel barrier is used as 286.65 9.3 5.5 170 8695.5 0.033 reference Flap gate Wijdenes 220 10.4 3.1 226 7286.24 0.030

The costs of each weir are calculated by determining the total width of the weir including the pylons. The width of the pylons is set at 12 metres (Thames barrier 11 metres, present weirs in the Nederrijn 13 metres). The total width is calculated by multiplying the width of the openings by the amount of openings and adding the width of the pylons (number of openings +1). The parameter BhH is calculated by multiplying the width, the height of the gate, and the head over the gate. Now the costs can be calculated by multiplying the BhH parameter with the key ratio of Table N-4 which results in the total costs presented in Table N-5.

Table N-5 Weir costs

width width pylons openings total width BhH € BhH Gate type [m] [m] [m] [m] [m3] [M€] 3x 41m; sub. seg. gate 41 12 3 171 6771.6 271 4x 29m; sub. seg. gate 29 12 4 176 6969.6 279 3x 41m; vert. lift. gate 41 12 3 171 6771.6 229 4x 29m; vert. lift. gate 29 12 4 176 6969.6 236 3x 41m; visor gate 41 12 3 171 6771.6 223 4x 29m; visor gate 29 12 4 176 6969.6 230 3x 41m; flap gate 41 12 3 171 6771.6 204 4x 29m; flap gate 29 12 4 176 6969.6 210 1x 120m; flap gate 120 12 1 144 5702.4 172

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N.4.4 SCORE

An weighted and an unweighted score are determined. The score of an unweighted comparison is presented in Figure N-10 and the score for a weighted comparison is presented in Figure N-11. The variants with the lowest costs are the flap gates; in special the flap gate without mid pylons. The variants with the highest score are the submerged segment gates. The visor and vertical lifting gate are placed in between the two gate types for the costs and performances. The scores taken into account are only implementation costs. Flap gates do have higher maintenance costs which are qualitatively taken into account for the performance. For the life cycle the flap gates can be more costly with respect to gate types which are easier to maintain. Figure N-10 and Figure N-11 can be analysed following the method presented in appendix B.2 resulting into the results presented in Table N-6.

Figure N-10 Cost-performances excluding weighting

Figure N-11 Cost-performances including weighting

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Table N-6 Result of weir variant assessment

Variant Result assessment 3x 41m; submerged Taking into consideration; highest score but also the highest costs. This segment gate variant does have the highest performance/cost ratio for the weighted score Reject; This variant has a lower score for a weighted comparison and an 4x 29m; submerged equal score for a unweighted comparison with respect to the 3x 41m segment gate submerged segment gate for higher costs. 3x 41m; vertical Reject; this variant has a lower performance for the same costs with respect lifting gate to the visor gates 4x 29m; vertical Reject; this variant has a lower performance for the same costs with respect lifting gate to the visor gates Taking into consideration for a weighted comparison; This variant has lower 3x 41m; visor gate costs than the submerged segment gate and a slightly lower performance/cost ratio. Reject; This variant has a lower score for a unweighted comparison and an 4x 29m; visor gate equal score for a weighted comparison with respect to the 3x 41m visor gate for higher costs. Taking into consideration; This variant has a higher score than the visor gates 3x 41m; flap gate for an unweighted comparison and a reasonable performance/cost ratio for a weighted comparison Reject; this variant has a lower score for higher costs for an unweighted 4x 29m; flap gate comparison and a slightly higher score for higher costs for a weighted comparison. Taking into consideration; lowest costs of all variants for weighted and 1x 120m; flap gate unweighted score

N.5 CONCLUSION

A selection of four best variants is made from the the presented variant. A design choice is made between the three variants which is included in the main report.  1x 120m; flap gate o This variant is not chosen for further elaboration unless the low costs. The maintenance disadvantage due to the large span and the impossibility of creating a dry working environment per gate are not desired.  3x 41m; flap gate o Is not a preferable solution due to the hinged connection which is located below water level and the flap which is always located below water level. It is hard to inspect the flap during normal situations. A better variant (visor gate) can be implemented by spending a larger amount of money  3x 41m visor gate o Is considered as a preferable solution. It has a good performance/cost score and it is proven technology. However an extra accurate discharge control is necessary because visor gates are not able to regulate low discharges.  3x 41m submerged segment gate o Is considered as a preferable solution. It has the best performance with respect to the other weir types. Furthermore, no extra discharge control is necessary because a weir equipped with overflow gates is better able to maintain the water level and the gates can be inspected more easily with respect to the other gates.

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O. Hydraulic models of the weir

The functioning of the weir is determined by the boundary conditions of the upstream reach. Three locations impose boundary conditions for the weir, which are:  The crossing of the Amsterdam Rijnkanaal. o Barrier Ravenswaaij closes at a water level of +5.55m NAP. Vessels have to use the locks to cross the Nederrijn and Lek, which results in delay. A maximum water level of +5.00m NAP at the crossing is set as boundary condition to prevent closure of Barrier Ravenswaaij during normal conditions. This water level has to be maintained by the weir; so the dammed water level at weir Culemborg must be lowered for increased discharges to realise a lower water level than +5.00m NAP at at barrier Ravenswaaij. This limitation of water levels is from now on called: ‘the Ravenswaaij limitation.’  The transition of the sloped bottom profile to a horizontal bottom profile of the Nederrijn at Wageningen. o A triangle has to be dredged from the bottom profile in between km 15 and km 30 (Figure O-1) to realise sufficient draught for professional navigation at the Nederrijn. The transition of the horizontal bottom profile and the sloping bottom profile forms a restriction of the depth. The water level at weir Culemborg must be high enough to realise a minimum water level of 4.2 metres at this transition. This transition is located near the city of Wageningen and is called from now on ‘the Wageningen limitation.’  Weir Driel. o Weir Driel has to regulate certain water levels at the IJsselkop to divert sufficient discharge into the IJssel. The functioning of the weir is influenced by the downstream water level (Driel beneden). The water level at ‘Driel beneden’ may not be higher with respect to the present water level during dammed conditions because this influences the steering efficiency of weir Driel. Therefore it has to be verified whether the dam regimes calculated for the Wageningen limitation and the Ravenswaaij limitation exceeds the water levels presented in Figure O-2.

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Figure O-1 Cross section of the Nederrijn and Lek at Q=0m3/s

Figure O-2 Water levels at Driel boven and Driel beneden as a function of discharge (based on: (ARCADIS, 2010))

A hydraulic model is made for calculating the dam regime of weir Culemborg. The dam regime has to maintain the upstream water levels at Ravenswaaij,Wageningen, and Driel. It is verified whether the boundary condition of Driel is exceeded for both dam regimes. The results of the calculations are a

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minimum and maximum dam regime of weir Culemborg. The water levels at weir Culemborg must be maintained in between the two dam regimes. The input parameters of the models are presented in Table O-1. The width of the Nederrijn is set as a constant to simplify the model However the width is not constant over the reach; a width of 95 metres is present at weir Driel and a width of 140 metres is present at weir Culemborg. Therefore, two calculations are made for both widths to verify the Driel boundary condition. A width of 140 metres is used for the resulting dam regime of weir Culemborg because the width of the last 20 kilometres equals 140 metres and represents the behaviour of the present weirs the best.

Table O-1 Input parameters of the hydraulic model

Parameter Symbol Unit Value Wet cross section of the waterway A m2 variable

Width of the Nederrijn and Lek Bnr m 90 or 140 Location of weir Culemborg xcul m 0 Location of Ravenswaaij xrav m 6000 Location of transition xtrans m 27500 Location of weir Driel xdri m 40000 Depth of the Nederrijn and Lek d m variable

Depth of transition near Wageningen dtrans m -0.8 Depth at Culemborg measured from +NAP dcul m 2.2 Water level at the Nederrijn and Lek h +m NAP variable

Water level at weir Culemborg hcul +m NAP variable Maximum water level at Ravenswaaij hrav +m NAP 5.00 Maximum water level at transition htrans +m NAP 5.00 -4 Bottom slope ib - 1,1*10 Flow speed u m/s variable Specific discharge q m2/s variable Discharge Q m3/s variable 3 Maximum discharge for realising the minimal depth Qmax m /s 568 Chézy coefficient Ch m1/2/s 45 Gravitational constant g m/s2 9.81

O.1 GENERAL MODEL FOR THE NEDERRIJN AND LEK WATER LEVELS

O.1.1 MODEL DESCRIPTION FOR THE WATER LEVELS IN THE NEDERRIJN AND LEK

Based on: (Voortman, 2011).

The hydraulic model is based on the following assumptions:  The Nederrijn and Lek are schematised as prismatic channels.  The slope of the Nederrijn and Lek is constant til the transition to the horizontal part at Wageningen.  The flow is stationary; water levels and discharges do not change over time.  The downstream water level of weir Culemborg is equal to: o An average discharge of 0.55m +NAP for low discharges. This water level is set by the discharge of the Waal and the influence of the tide which is determined during the RINK project (Dongen, 2012). o The equilibrium water level for higher discharges.

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The input parameters of the hydraulic model are drawn in Figure O-3 and Figure O-4. The water levels and bed levels of the river are measured with respect to the reference level NAP. The bed levels are positive, so a bed level of -2.2m NAP (which is present at weir Culemborg) is a positive value for the depth.

Figure O-3 Wet cross section of the Nederrijn and Lek

Figure O-4 Longitudinal cross section of the Nederrijn and Lek

The hydraulic model is based on the formula of Bernoulli which is presented below:

In which: H = energy head [m] U = flow speed [m/s] h = water level [m] g = gravitational constant [m/s2]

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An expression for the water level could be obtained by rearranging the Bernoulli formula:

The course of the water level is obtained by differentiating the rearranged Bernoulli formula to the x- coordinate of which the result is presented in Equation O-1.

Equation O-1Course of water level over x

The formula of Chézy is used to obtain an expression for The Chézy equation holds:

√ In which: Ch = Chézy coefficient [m0.5/s] R = Hydraulic radius [m] i = Bottom slope [-]

The formula of Chézy is rewritten by assuming the hydraulic radius to be equal to depth and the bottom slope equal to the course of hydraulic head resulting in:

√

In which the flow velocity is equal to:

This results by combining both formulas in:

√

Equation O-2 is obtained by rearranging this expression. Equation O-2 can be substituted in to Equation O-1.

Equation O-2 Course in head over x

An expression for is obtained by using the continuity equation:

An expression for is obtained by differentiating the continuity equation to x.

The flow is stationary so . Therefore the differentiated continuity equation changes into Equation

O-3.

Equation O-3 Course of velocity

The area A is a function of bottom depth, water level and width as presented below:

This equation is differentiated to x in order to obtain the course of the wet cross section A. The result is shown below:

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( )

The width is uniform so . Equation O-4 is obtained by substituting this in the course of wet cross section.

( )

Equation O-4 Course of wet cross section

Now Equation O-1 is elaborated further. First the flow velocity is rewritten for the discharge divided by the wet cross section of which the result is presented in Equation O-5.

Equation O-5 Course of water level over x

Now Equation O-2 and Equation O-3 are substituted into Equation O-5 which results in Equation O-6.

( ) ( )

Equation O-6 Course of water level over x

Subsequently the course of wet cross section (Equation O-4) is substituted in into Equation O-6 of which the result is presented in Equation O-7.

( ) ( )

Equation O-7 Elaborated course of water level over x

Equation O-7 has to be rearranged to use the expression for in the model. The resulting function of the course of water level has been found and presented in Equation O-8.

( ( )) ( ) ( ) ( ) ( )

Equation O-8 Final expression for the course in water level

The water levels at the reach Nederrijn and Lek are calculated for a range of x and Q, so the differential equation is solved for two variables using Runge-Kutta 4

O.1.2 VERIFICATION OF THE MODEL FOR A (PARTIAL) DAMMED RIVER

To verify whether Equation O-8 represents the behaviour of the river well, several verification calculations are performed for a ranging discharge. The results of these calculations are compared with the results of a modal based on the Bresse’s formulas which are described in A.6. The results of the calculations of Equation O-8 and Bresse’s formula are presented in Figure O-5 and Table O-2. The water levels at the Wageningen limit have been calculated using the minimal dam regime of weir Culemborg. The water levels at the Wageningen limit calculated with the formula of Bresse should be equal to 7.2m (bottom level at -2.2m NAP, water level at 5m NAP, resulting in a water depth of 7.2 m) when corrected for the bottom angle. The difference between both models is not larger than 5% as presented in the last column of Table O-2. The models nearly equal each other, so Equation O-8 and model the behaviour of the river well.

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Figure O-5 Water levels for minimum dam regime and discharges lower with respect to the openings discharge

Table O-2 Water levels at the Wageningen limit (27.5km from weir Culemborg) calculated with Bresse Specific Normal Half Depth z(27.5km) Differ- Discharge discharge depth length at x=0 (Bresse) h(27.5km) z(27.5 km) ence [m3/s] [m2/s] [m] [m] [m] [m] [m] [m] % 25 0.2 0.5 37626 7.20 7.54 5 7.2 4.78 200 1.4 2.1 22933 7.02 7.24 5 7.2 0.54 400 2.9 3.3 17141 6.34 7.31 5 7.2 1.56 500 3.6 3.9 13539 5.51 7.26 5 7.2 0.82 569 4.1 4.2 9187 4.21 7.20 5 7.2 0.01

O.1.3 VERIFICATION OF THE MODEL FOR AN OPEN RIVER

The water levels at the sloped river bed should be parallel to the bottom for an undammed river, because a ‘normal depth‘ is present for a river in equilibrium state. Furthermore, the water levels at the transition from the sloped river bed to the horizontal part should follow a back water curve. The calculated water

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levels of the sloped part of the Nederrijn-Lek are parallel to the bottom as presented in Figure O-6. So, the model models the behaviour of the river well for an free flowing river.

Figure O-6 Water levels for minimum dam regime and discharges higher with respect to the openings discharge

O.1.4 COMPARISON WITH THE PRESENT DAM REGIMES

The present control range of weir Driel spans from 0m3/s to 412 m3/s (based on (ARCADIS, 2010)). The control range of a weir located in the Nederrijn for a width of the Nederrijn of 95 metres determined with the hydraulic model is 0m3/s till 386m3/s which approaches the control range of the weir Driel. The stage duration graph of a schematised Nederrijn for a width of 95 metres is presented in Figure O-7. The present weirs are 29% a year open (ARCADIS, 2010) and the weirs for the minimum dam regime generated by the Wageningen limit for a width of the Nederrijn of 95 metres are 122 days a year open, which is 33% a year. So, the schematisation of the Nederrijn for a width of 95 metres matches the operation of weir Driel. The control range of weir Amerongen and Hagestein spans from 0m3/s to 635m3/s (based on: (ARCADIS, 2010)). These weirs are opened as soon as possible for navigational purposses. Therefore the control range of weir Culemborg is calculated for the minimum dam regime which is generated by the Wageningen limit. The control range of the minimum regime is 0m3/s to 568m3/s as presented in Figure O-8 for a width of the Nederrijn-Lek of 140 metres. The river is dammed for 310 days a year and undammed for 55 days a year for the minimal dam regime, which is 15% of a year. The present weirs are open for 13% a year

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(ARCADIS, 2010) which nearly matches the calculated result. The control range of the present weirs is larger because the width of the river at Hagestein is larger with respect to the width of the river at the upstream side of weir Culemborg. A higher discharge is needed to generate more depth for a wider river, so the control range of the present weirs has to be larger with respect to weir Culemborg. From this comparison, it is concluded that the schematisation of the Nederrijn for a width of 140 metres matches the functioning of weir Amerongen, Hagestein, and could also be used for the new weir located near Culemborg.

Figure O-7 Stage duration curve for Bnr=95m

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Figure O-8 Stage duration curve for Bnr=140m

O.1.5 RESULT OF THE HYDRAULIC MODEL FOR THE NEDERRIJN AND LEK

Now the water levels at the Nederrijn are calculated for the set limits at barrier Ravenswaaij and the transition of the sloped bottom and the horizontal bottom profile at Wageningen. Furthermore, it is verified whether the water levels at Driel exceeds the present ‘normal water levels’ at weir Driel which are given in Figure O-2.

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The model is executed for a width of 95 metres and 140 metres as lower and upper boundary. The resulting dam regimes of weir Culemborg are presented in Figure O-9 for a width of 95 metres and in Figure O-10 for a width of 140 metres. The red continuous line presented in Figure O-9 and Figure O-10 represents the maximum dam regime at weir Culemborg which is generated by the Ravenswaaij limit. The blue dotted line in Figure O-9 and Figure O-10 represents the minimum dam regime which is generated by the Wageningen limit. The green dashed line represents the downstream water level at weir Culemborg. For low discharges, the water level is constant at +0.55m NAP which is generated by a combination of tide and the runoff of the Waal. The downstream water level increases according to the ‘normal water level’ for higher discharges for which the water level of +0.55m NAP is exceeded. The upstream and downstream water levels at the weir are equal to each other when the blue dotted line intersects with the green dashed line. At this discharge the weir can be opened. A higher upstream water level can be maintained by damming the river for higher discharges which could be used for hydro power for example. The weir has to be opened to prevent Ravenswaaij from closing when the red continuous line intersects the green dashed line. A dam regime which meets the requirements is obtained when the water levels are in between or equal to the red continuous line and the blue dotted line. Sufficient water depth for navigation is available when the blue dotted line intersects the dashed-dotted purple line which represents the minimal draught for navigation at the weir.

Figure O-9 Dam regime for Bnr=95 metres

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Figure O-10 Dam regime for Bnr=140 metres

Still it has to be verified whether the maximum water level at weir Driel is exceeded. Therefore the water levels at weir Driel are calculated for a width of 95 metres and 140 metres. The water levels presented in Figure O-11 nearly equals the water levels of ‘Driel beneden’ and the weir can be opened according to the present dam regime of weir Driel with some small adjustments. The water levels calculated for a width of 140 metres are lower with respect to the water levels of ‘Driel beneden’ as presented in Figure O-12. So, it can be concluded that the water levels of the Driel limit are not exceeded for all widths and discharges.

O.1.6 SUITABLE DAM REGIME FOR WEIR CULEMBORG

The dam regime calculated for a width of 140 metres is used for the elaboration of weir Culemborg. A width of 140 metres of the Nederrijn models the behaviour of the river the best for the downstream region and equals the behaviour of weir Hagestein which is situated nearby Culemborg. The result of the calculations for a width of the Nederrijn of 95m approaches water levels at weir Driel and is used for the verification of the dam regime of weir Driel. More accurate results have to be obtained by using a model with a variable width but this is beyond the scope of this graduation research.

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Figure O-11 Water levels at Driel for Bnr=95 metres

Figure O-12 Water levels at Driel for Bnr=140 metres

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O.2 LOCAL MODEL FOR THE FLOW AT THE WEIR

A local model is made for the calculation of the flow characteristics at the contraction of the weir. The description of the local model is given in O.2.1 and the results are presented in O.2.2.

O.2.1 MODEL DESCRIPTION FOR THE WATER LEVEL AND SPEED AT CONTRACTION OF THE WEIR

Used literature: (Battjes & Labeur, 2009) & (Nortier, 1989).

The width of the waterway decreases from the width of the Nederrijn (Bnr) to the width of the weir gates

(Bgate) as presented in Figure O-13. The contraction of the waterway takes place at a limited distance, so the river bed is assumed to be horizontal which is justified for the small river bed angle. Two sub models are made for the calculation of the local water levels and flow speeds. The parameters used in the flow model are presented in Table O-3.

Table O-3 Input parameters for the flow speed calculation

Parameter Symbol Unit Value Width of the Nederrijn and Lek Bnr m 140 Width of the gates Bgate m variable Depth of the Nederrijn and Lek d m variable Water level at the Nederrijn and Lek h +m NAP variable

Water level at the downstream boundary hds +m NAP variable Water level at the upstream boundary hup +m NAP variable Flow speed u m/s variable

Flow speed at the upstream side of the weir uup m/s variable Flow speed at the downstream side of the weir uds m/s variable Specific discharge q m2/s variable Discharge Q m3/s variable

Energy head at the upstream side of the weir Hup +m NAP variable Energy head at the downstream side of the weir Hds +m NAP variable Discharge coefficient for weirs and locks col - 0.7≤col<1.4 Gravitational constant g m/s2 9.81

Figure O-13 Contraction of the waterway

: ARCADIS & TUDelft 215

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The first model uses a discharge formula for open weirs and lock flow which is presented in Equation O-9 and calculates the water level difference. This equation for the flow through a contraction is solved using

Mathcad for the downstream water level zds.

√ ( )

Equation O-9 Submerged weir flow (Nortier, 1989)

The second model uses the energy balance which is applied to the upstream side of the weir for a normal width and at the weir for the contracted width. The flow speed increases, so no energy losses are present and an energy balance can be applied as presented in Equation O-10; the energy at the normal waterway is equal to the energy of the contracted waterway.

Equation O-10 Energy balance

Equation O-12 is obtained by filling in the Bernoulli equations at on both sides of Equation O-10.

Equation O-11 Energy head

In which: z = h+d

Equation O-12 Elaborated balance equation

Two unknown variables are present in Equation O-12 which are the water column at the weir zw and the

flow speed at the weir uweir, so an extra equation is necessary for solving Equation O-12. The extra equation is obtained by introducing the volume balance which is presented in Equation O-13.

Equation O-13 Volume balance

Equation O-14 is obtained by elaborating and rearranging Equation O-13. The result is filled in into Equation O-12 which results in Equation O-15. Equation O-15 is solved for the flow speed at the contracted part of the waterway at the weir.

Equation O-14

Equation O-15

The result for uweir is used for determining the water column at the contracted waterway with use of Equation O-14.

O.2.2 APPLICATION OF THE FLOW MODEL FOR DETERMINATION OF THE WIDTH OF THE WEIR

The width of the weir is determined by the number of openings, the minimal width per opening, the flow velocity through the gaps when the weir is open, and the design discharge of the river.

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Three widths of the openings are applicable according to the requirements, which are:  a minimal width of 29 metres for one lane traffic for navigational class Va (FR 10)  a minimal width of 41 metres for two lane traffic for navigational class Va (FR 10)  a minimal width of 25 metres for recreational class AM (FR 11)  one undisturbed opening. The weir must be realised in a constriction of the river (PIANC, 2006). The width of the river is 140 metres, so a weir complex which nearly equals the width of the river is not a good solution. Furthermore, the flow speed at the weir must be lower than 2m/s for the passage of navigation (Meijer, 2012). Therefore the constricted waterway may not be too constricted because the flow speed at the weir becomes too. The flow speed in the summer river bed has been calculated till a discharge of 1500m3/s. Higher discharges causes flooding of the floodplains which can be concluded on from Figure O-14. The continuous brown ‘saw tooth’ line represents the top of the summer levees along the river (source: Table B-13) the water levels for Q=1500m3/s, 1600m3/s, 1700m3/s, and 1800m3/s are also indicated in Figure O-14. Water levels generated by discharges larger than 1500m3/s exceeds the summer levees. More area is present for the flow and the discharge spreads over the floodplains. Therefore, it is assumed that the maximum rate of flow in the main summer river bed is reached for a discharge of 1500m3/s.

Figure O-14 Maximum discharge for the summer river bed

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The maximum flow speed for a ranging opening is indicated by a horizontal red line in Figure O-15. Two distinct regimes are identified namely for the dammed Nederrijn and the open Nederrijn. The flow speed for the dammed Nederrijn increases exponentially for the first part because the water level decreases from +5.0m NAP for zero discharge to +2m NAP for the maximum discharge and the rate of flow increases. So more water flows through a smaller wet cross section for higher discharges which causes the flow speed to increase rapidly. After lifting of the gates, the water level increases according to the normal depth and a larger wet cross section is present for higher discharges. Therefore the flow speed increases less rapidly with respect to the first part of the graph. It can be concluded on basis of Figure O-15 that the weir must be larger than 95 metres in order to maintain a maximum flow speed of 2m/s.

Figure O-15 Flow speed at the weir

Another restriction is the set-up of water generated by the presence of the contracted waterway. The water level difference at the upstream side of the waterway and at the contracted waterway are calculated till a discharge of 1500m3/s. The results are presented in Figure O-16; the set up may not be larger than 0.1 metres in order to reduce the set up for high water levels (Voortman, 2012). Based on Figure O-16 can be

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concluded that the width of the gate must be larger than 100 metres in order to maintain a smaller set up than 0.1 metres.

Figure O-16 Water level set up

O.3 MODEL FOR AN OVERFLOW GATE

A sub model is made for the calculation of the characteristics of an overflow gate. The model is described in O.3.1 and the results are presented in O.3.2.

O.3.1 MODEL DESCRIPTION FOR A WEIR WITH OVERFLOW GATES

Used literature: (Ankum, 2002), (Cruise, et al., 2007) and (Voortman, 2009).

The upstream water levels which have to be maintained are already determined in the dam regime calculation. The weir and in specific the gates of the weir have to control these upstream water levels. The crest of the gate could be lowered to increase the discharge over the gate and to lower the water level at the reach. The parameters used in this section are presented in Table O-4.

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Table O-4 Input parameters used for overflow calculation

Parameter Symbol Unit Value Wet cross section of the waterway A m2 variable

Width of the Nederrijn and Lek Bnr m 140 Width of the gates Bgate m variable Water level at the Nederrijn and Lek h +m NAP variable

Water level at the downstream boundary hds +m NAP variable Water level at the upstream boundary hup +m NAP variable

Crest level of a structure controlled overflow gate hcr_sc +m NAP variable Crest level of a tail controlled overflow gate hcr_tc +m NAP variable Crest level hcr +m NAP variable Tail controlled flow speed utc m/s variable Structure controlled flow speed usc m/s variable Discharge Q m3/s variable

Discharge coefficient of an overflow weir Cgate - 1.9 Discharge coefficient for tail controlled weir Ctc - - Discharge coefficient for structure controlled weir Csc - - Gravitational constant g m/s2 9.81

Two distinct flow regimes are identified for an overflow weir, namely:  Tail controlled flow, presented in Figure O-17 (Dutch: onvolkomen overlaat) o The downstream water level is high enough to influence the flow over the weir.  Structure controlled flow, presented in Figure O-18 (Dutch: volkomen overlaat) o The downstream water level is too low to be able to influence the upstream water level.

Figure O-17 Tail controlled weir flow

Figure O-18 Structure controlled weir flow

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The downstream water level is divided into regions:  The downstream water level is set constant at +0.55m NAP. This value is determined during the RINK project using SOBEK as a tide average water level for low discharges.  The downstream water level is equal to the equilibrium water level calculated with Equation O-16. In which d(0) is the bottom level at weir Culemborg. Equation O-16 is used when the result is higher than +0.55m NAP. ⁄

( )

Equation O-16 Downstream water level of weir Culemborg

O.3.1.1 TAIL CONTROLLED FLOW

The flow speed for tail controlled flow is given by Equation O-17 and the cross sectional area for the flow is given by Equation O-18.

√ ( )

Equation O-17 Flow speed for tail controlled weir flow

Equation O-18 Flow area for tail controlled weir flow

The discharge over the weir is obtained by combining Equation O-17 and Equation O-18 and introducing a discharge coefficient; the result is shown in Equation O-19.

√ ( )

Equation O-19 Tail controlled discharge

The parameter to be solved is the crest level hcr for a variable discharge Q. Therefore Equation O-19 is rewritten. First the brackets of Equation O-19 where the crest level is included must deleted of by multiplying. The result is presented below:

√ ( ) √ ( )

The result is rearranged and written as a function of hcr as presented in Equation O-20.

√ ( )

Equation O-20 Crest level for tail controlled flow

O.3.1.2 STRUCTURE CONTROLLED FLOW

The flow speed for structure controlled flow is given by Equation O-21 and the cross sectional area for the flow is given by Equation O-22.

√ ( )

Equation O-21 Flow speed for structure controlled weir flow

( )

Equation O-22 Flow area for structure controlled weir flow

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The discharge formula for structure controlled weir flow is found by combining Equation O-21 and Equation O-22 and introducing a discharge coefficient.

⁄ ( )

Equation O-23 Structure controlled discharge

The parameter to be solved is the crest level hcr for a variable discharge Q. Therefore, Equation O-23 is rewritten. First the power of 3/2 of must deleted for tail controlled flow by raising the equation to the power of 2/3; the result is presented below:

⁄ ⁄ ( √ ) ( )

Now the brackets where the crest level is in included must be deleted by multiplying which results in the following expression:

⁄ ⁄ ⁄ ( √ ) ( √ )

This can be rearranged and written as a function for hcr of which the result is presented in Equation O-24.

⁄ ⁄ ( √ )

⁄ √ ( )

Equation O-24 Crest level for structure controlled flow

O.3.1.3 DISCHARGE COEFFICIENTS

The discharge coefficients for tail and structured controlled flow are presented in Equation O-19 and Equation O-23. Nortier made a expression to relate both coefficients to each other. The expression is presented in Equation O-25.

√

Equation O-25 Related discharge coefficients

The resulting expression for a structure controlled discharge using a tail controlled discharge coefficient is presented in Equation O-26. Both discharges can be determined using the discharge coefficient for tail controlled flow only.

⁄ √ ( )

Equation O-26 Structure controlled discharge

O.3.1.4 TRANSITION OF TAIL AND STRUCTURE CONTROLLED FLOW

A transition limit between structure controlled flow and tail controlled flow is present; a structure controlled flow becomes tail controlled flow when the downstream water level influences the rate of flow over the weir. The limit for tail and structure controlled flow is presented in Equation O-27. The flow becomes tail controlled when the downstream water level is equal or larger with respect to the right hand side of Equation O-27.

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( )

Equation O-27 Transition between tail and structure controlled flow

It has to be verified whether the discharge equations for tail controlled and structure controlled flow and the crest level equations for tail controlled and structure controlled flow are continuous differentiable. This means that the graphs intersects at the same angle and a continuous transition from structure controlled flow to tail controlled is present. Two sub-models are made in order to verify this.

The upstream water level (1.5m above crest level) and the crest level (0m above crest level) are set as constants for the discharge model. The downstream water level varies from 0m to 1.5m measured from the crest level. The structure controlled discharge is constant according to Figure O-19 because the downstream water level does not influence the discharge of a structure controlled weir. The angles of both graphs are equal at the intersection, so the discharge formulas are continuous differentiable.

Figure O-19 Verification of the continuous differentiability of the discharge equations

The upstream water level (1.5m above crest level) and the discharge (500m3/s) are set as constants for the crest level model. The downstream water level varies from 0m to 1.5m measured from the same reference as the upstream water level. The structure controlled crest level is constant according to Figure O-20 because the downstream water level does not affect the discharge over a structure controlled weir. Also in this case, the angels of both graphs are equal at the intersection, so the crest level equations are continuous differentiable.

Figure O-20 Verification of the continuous differentiability of the crest level functions

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O.3.2 RESULTS OF THE HYDRAULIC MODEL OF A WEIR WITH OVERFLOW

The crest levels as a function of discharge for a 100 metre wide weir are presented in Figure O-21. The green dotted line represents the crest level for structure controlled flow and the red line for tail controlled flow. The crest height is +5m NAP for zero discharge and decreases for an increasing discharge. The flow becomes tail controlled at discharges higher than 550m3/s. The capacity of the weir decreases rapid due to the lower discharge capacity of a tail controlled flow and the limited head over the weir caused by the rapid decrease of the upstream dammed water level. The gate position is theoretically lower with respect to the bed level for discharges just before opening. This is caused by the expression under the root of the tail controlled flow equation (Equation O-28) which approaches zero. Therefore the formula reaches a limit which causes the rapid decrease of the crest level.

√ ( )

Equation O-28 Crest level for tail controlled flow

The flow characteristics for larger widths are determined in order to solve this rapid decrease. Therefore, the crest levels are calculated for a weir with of 100 metre, 110 metre, 120 metre, 130 metre and 140 metre for which the results are presented in Figure O-22. Also the upstream water level, downstream water level, and the sill level are presented in Figure O-22. The crest height decreases rapid for every widths, so another solution has to be found for the rapid decrease of crest level.

Figure O-21 Crest level of an overflow weir for a 100m wide opening for the minimal dam regime

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Figure O-22 Crest levels and water levels for different opening widths for the minimal dam regime

A solution is the extra set up of the upstream water level. The upstream water level has to be set up higher in order to maintain sufficient head over the weir for discharges larger than 550m3/s to prevent the gate from rapidly lowering. A calculation is made with a dam regime which is equal to the minimal dam regime increased with 30% of the difference of the minimal and maximal dam regime which is determined using Equation O-29. The minimum dam regime, maximum dam regime, and the new dam regime are presented in Figure O-23.

( )

Equation O-29 New dam regime

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Figure O-23 New dam regime

The resulting crest levels are presented in Figure O-24. The gates have to be lowered less rapid with respect to the calculated crest levels presented in Figure O-22 and the weir is able to control the water levels without reaching a level lower than the sill level. So, the mentioned problem is solved by creating a larger set up of water at the upstream side of the weir. The steep decrease at the end of the control range represents the lowering of the gates in order to restore an open river for discharges which are higher than 568m3/s.

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Figure O-24 Crest levels for the new dam regime for different gate widths and the increased dam regime

O.4 MODEL DESCRIPTION FOR A WEIR WITH UNDERFLOW GATES

Used literature: (Ankum, 2002), (Cruise, et al., 2007) and (Nortier, 1989).

Water flows underneath the gate from the upstream to downstream side of the weir for an underflow gate. At the downstream side of the weir a hydraulic jump is generated which is highly instable. By application of an (underflow) weir the hydraulic jump is usually located against the downstream side of the weir. Stilling basins are used to ‘push’ the hydraulic jump against the weir if the hydraulic jump is located too far away from the weir without application of a stilling basin (Nortier, 1989). This is also in favour of the vertical stability of the weir because a larger layer of water is present at the downstream side of the weir which counterbalances the uplift force generated by the groundwater pressure.

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The gates of the weir are lifted to let a certain discharge pass the weir. The size of the ‘gap’ underneath the gate is equal to µa in which µ is a contraction coefficient. The flow accelerates from the upstream side of the weir till the minimum gap size and the flow decelerates from the minimum gap size. The discharge of a lifting gate is calculated using the energy equation presented in Equation O-30 prior to the jump and the momentum equation presented in Equation O-31 across the jump. Also another method is available for determining the gap size for underflow gates. This method is described in ‘handleiding voor het ontwerpen van granulaire bodemverdedigingen achter tweedimensionale uitstromingsconstructies’ (Rijkswaterstaat, 1995). This method is used in order to verify the results of the first model. This is necessary because numerical instabilities are present for solving the first model for the gap sizes. The parameters used for both model are presented in Table O-5 and Figure O-25.

Table O-5 Parameters for an underflow gate

Parameter Symbol Unit Value Wet cross section of the waterway A m2 variable

Energy level at upstream side of the weir Hup m variable Flow speed at the gate Ugap m/s variable Width of the gates Bgate m variable Water level at the weir zw m variable Water level at the downstream boundary zds m variable Water level at the upstream boundary zup m variable Gap underneath the underflow gate a m variable Specific discharge q m2/s variable

Momentum at the upstream side of the weir Fup N variable Momentum at the constriction Fds N variable Energy head at the upstream side of the weir Hup +m NAP variable Energy head at the downstream side of the weir Hds +m NAP variable Discharge coefficient of an underflow weir µ - variable Gravitational constant g m/s2 9.81

Figure O-25 Submerged underflow weir

O.4.1 HYDRAULIC MODEL BASED ON ENERGY AND MOMENTUM EQUATION

The energy level at the upstream side of the weir is equal to the energy level at the minimum cross section at the weir which is presented in Equation O-30.

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Equation O-30 Energy equation

In which:

Hup = the energy level at the upstream side of the weir

Hw = the energy level at the maximum contraction of the flow

The rate of momentum at the maximum contraction of the flow is equal to the rate of momentum at the downstream side of the weir which is expressed in Equation O-31.

Equation O-31 Momentum equation

In which:

Fup = the momentum at the maximum contraction of the flow

Fds = the momentum at the downstream side of the weir with ‘normal’ flow

Equation O-32 is obtained by substituting the Bernoulli equation into Equation O-30 and Equation O-33 is obtained by substituting the expression for momentum into Equation O-31.

Equation O-32 Filled in energy equation

Equation O-33 Filled in momentum equation

Two unknowns are present in Equation O-32 and Equation O-33 which are the gap size ‘a’ and the water level at the hydraulic jump. Two equations are available, so this system can be solved. The system is solved by rewriting the energy equation and substituting the rearranged energy equation into the momentum equation of which the result is presented in Equation O-35. Equation O-35 is solved for ‘a’ using Mathcad.

Equation O-34 Energy equation expressed for hw

( )

Equation O-35 Final equation to be solved

O.4.2 HYDRAULIC MODEL BASED ON THE RWS MANUAL

Based on: (Rijkswaterstaat, 1995).

This model is based the flow out of a reservoir through a small opening using the energy equation and neglecting the upstream velocity for which the formula is presented in Equation O-36.

√

Equation O-36 Flow speed through a small opening (Battjes & Labeur, 2009)

The specific discharge q is calculated by introducing in which µ=0.8 which results in Equation O-37.

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√ ( )

Equation O-37 Expression for specific discharge

The gap size of the weir is calculated by rearranging Equation O-37. The rearranged formula is presented in Equation O-38. This formula is solved for ‘a’ for a varying discharges using Mathcad.

√ ( )

Equation O-38 Gap size underneath the gate

O.4.3 RESULTS OF THE HYDRAULIC MODEL OF A WEIR WITH UNDERFLOW GATES

The gate levels of weir equipped with underflow gates are calculated for a total opening width of 100m, 110m, 120m, 130m, and 140m to determine the gap size. But first the results of both methods are compared of which the result of a 100m wide opening is presented in Figure O-26.

Figure O-26 Comparison of both methods for Bweir=100m and the minimal dam regime

Both methods do have nearly the same result for the gate levels of an underflow gate. Therefore one of the methods can be used for the calculation of the gate levels for different widths of the weir. The result of this calculation is presented in Figure O-27. At zero discharge the gate level is equal to the level of the sill (- 2.2m NAP). The gate is lifted for higher discharges to let water pass the weir. At higher discharges the gate has to be lifted rapidly in order to let sufficient discharge pass the gate. The theoretical gate level is larger with respect to the water level for realising sufficient discharge for discharges higher than 550m3/s. This is caused by the limited head and the large discharge which has to be generated. Therefore an increased minimal dam regime is maintained to solve this. The same increased new dam regime as used for the overflow gate is applied for the gate level calculation of an underflow gate. The resulting gate levels are presented in Figure O-28. The gate levels do not exceed the water levels, so the weir is able to regulate the water levels and discharges for the increased dam regime. The steep increase at the end of the control range represents the lifting of the gates in order to restore an open river.

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Figure O-27 Gate level for different opening widths and the minimal dam regime

Figure O-28 Gate levels for different widths and increased dam regime

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O.5 ACCURATE DISCHARGE CONTROL

The controllability of an underflow and overflow gate is determined in this section. The crest levels of an underflow gate are determined in O.5.1 and for an overflow gate in O.5.2.

O.5.1 UNDERFLOW GATE

The minimum flushing discharge which has to be regulated by the weir is 25m3/s. The gap sizes necessary for the minimum flushing discharge and higher discharges for weir openings from 0m to 140 metres are presented in Figure O-29. The step sizes for large openings are very small. The gates have to be lifted over 3 centimetres for a total width of 100 metres to realise a discharge of 25m3/s which is undesired because small openings of the gate results in undesired vibrations (PIANC, 2006). Therefore a smaller opening is necessary to enlarge the gap size for the minimum discharge. The gates have to be lifted over 11 centimetres for an opening width is 29 metres and in order to realise a discharge of 25m3/s which is also very small. Therefore a separate accurate discharge control is necessary in order to let pass the minimum discharge of 25m3/s.

Figure O-29 Gap sizes of an underflow weir

O.5.2 OVERFLOW GATE

The gate levels of an overflow gate for a ranging width and a variable discharge are presented Figure O-30. The gates have to be lowered over 18 centimetres in order to generate the minimum discharge of 25m3/s. For a width of 29 metres, the gates have to be lowered over 42 centimetres in order to generate the minimum discharge.

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Figure O-30 Gate position of an overflow gate

O.5.3 CONCLUSION

The head is raised to the power of 1.5 for a sharp crested weir of which the equation is presented in Equation O-39. So a slightly higher water level results in a much larger discharge. The head of an underflow gate is raised to a power of 0.5, so an equal head for an underflow gate results in less discharge with respect to an overflow gate. So the application of overflow gates results in a better controllable upstream water level so an accurate discharge control is not strictly necessary (Erbisti, 2004).

⁄ √ ( )

Equation O-39 Discharge formula of an overflow gate

√ ( )

Equation O-40 Discharge formula of an underflow gate

The distance for which an overflow gate has to be lowered to realise a certain discharge is much larger with respect to the distance an underflow gate has to be lifted. Therefore it is better to use an overflow gate for the accurate discharge control with respect to an underflow gate. Furthermore an underflow gate cannot be used for the minimum discharge because the gap size is too small. Therefore a weir complex equipped with underflow gates has to be equipped with an accurate

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discharge control which has to let through the minimum discharge and which has to fine tune the discharge.

O.6 MATHCAD SHEETS

The calculations are performed using the program Mathcad. However the Mathcad sheets are not printed but digitally provided on a CD or USB drive. Also a pdf file for the Mathcad sheets is provided at the repository of the TU Delft (repository.tudelft.nl)

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P. Load definitions

P.1 LOADS AT THE GATE

Permanent loads, variable loads, and special loads are acting at the gate. An overview of the permanent loads, the variable loads, and the special loads are given a handbook and enumerated below (Technische adviescommissie voor de waterkeringen, 2003). A permanent load is:  self-weight. The variable loads are:  hydrostatic pressure  waves  loads caused by rotation of the gate  currents. Special loads are:  collision  earthquake  explosion  ice  current for non-closure of a gate  sabotage  installation.

P.1.1 PERMANENT LOADS

Self-weight The composite is composed of resin (venylesther) and reinforcement (glassfibre). The weight of venylesther is 1089kg/m3 and the weight of glassfibre is 2443kg/m3 (FiberCore-Europe, 2011). The chosen volume content of the laminate for hydraulic structures is 50% (Stichting CUR, 2003). The specific density of the laminate is calculated with Equation P-1.

( )

Equation P-1 volumetric density of the laminate

In which:

Vf = glassfibre content [vol%] 3 = specific density of glassfibre [kN/m ] 3 = specific density of venylesther [kN/m ]

The specific density of the laminate is calculated by filling in this formula:

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[ ] [ ⁄ ] [ ] [ ] [ ⁄ ] [ ] ⁄ ⁄ [ ] ⁄

P.1.2 VARIABLE LOADS

Sediments The gate is positioned at the bottom of the waterway when the weir is open. A layer of sediments of 0.1meters is assumed to be present at the gate which is also used for a weir calculation of flap gates in California (ARCADIS, 2010). The specific density of the sediments is assumed to be 20kN/m3 whis results in a load of 2kN/m2 on top of the gate.

Hydrostatic pressure The hydrostatic pressure is caused by the water column. Two maximum configurations are present for which the hydrostatic pressure has to be taken into account:  The weir is fully open and is located in its recess. So the pressure at the topside of the gate is equal to the water column at the gate, and the pressure at the downside of the gate is equal to the water column at the gate plus the gate thickness.  The weir is closed. The topside of the gate is located at +5.00m NAP and is equal to the maximum dammed water level. The bottom side of the gate is located at -3m NAP which corresponds to a water column of 8 metres. So the hydrostatic pressure ranges from 0kN/m2 at the topside of the gate till 80kN/m2 at the bottom side of the gate.

Waves The height and period of the waves acting at the gate are determined for the maximum fetch and the maximum depth using the program Hydra-R and the formulas of Bretschneider. The waves acting on the gate are 0.22m high according to appendix Q.1. The hydraulic pressure generated by the waves is determined using the formula of Sainflou of which the result is presented in appendix Q.2.

Loads caused by rotation of the gate The loads caused by rotation are transferred by a mechanism placed on a pylon to the disks. The disks have to bear the loads caused by rotation of the gate.

Currents A load at the gate is present for an open situation, and for party closed operation. Minor currents are present at the gate for the maximum head over the gate Therefore, forces created by the currents are neglected for a fully closed oparation.

P.1.3 SPECIAL LOADS

Collision A vessel could collide with the gates. The gate should be able to absorb the energy of a colliding ship. An indication of the frequency of a colliding ship is determined below (Technische adviescommissie voor de waterkeringen, 2003). The amount of commercial vessels moving through weir complex Hagestein is 8297 per year (Table C-12). For the new configuration, the weir is not accessible for 15% of a year resulting in 7052 passages a year. The probability of failure per vessel passage is 3E-5 for locks. The collision of a vessel with a weir gate is less probable than a collision of a vessel with a lock door. Therefore, it is assumed that the probability of failure per vessel can be lowered by a factor 100 resulting in a probability per vessel of 3E-7. The collision

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frequency is determined by: per year, which is very low. Therefore, the collision force is not taken into account for the design of the gate in this research. For a real project, this has to be taken into account.

Earthquake Hydraulic structures are generally not designed for earthquakes in the Netherlands. Therefore, earthquakes are not taken into consideration for the design of the weir.

Explosion An explosion of a vessel or pipe line has to be taken into account for the design of the weir. The probability of an explosion of a vessel is limited due to the limited amount of passages a year. The load caused by an explosion has to be determined for a more detailed design of a gate. This is beyond the scope of this graduation research and therefore not performed.

Ice The gates have to be kept free from ice in winter. Otherwise, the gates are not able to rotate anymore. The gates have to be closed when it is not possible to keep the gates free from ice.

Sabotage People can deliberately damage the structure and the gates. Implementing the load caused by sabotage is beyond the scope of this graduation research but should be considered for a real design project.

Dragging anchor An anchor which is dragged forward by the ship could be hooked at the tip of the gate when the gate is positioned in the recess of the foundation. The load of a dragging anchor can be minimized by shaping the sill and gate in such a way that an anchor would not be hooked at the tip of the gate. This has to be elaborated in further detail.

P.2 LOAD FACTORS

Two limit states are defined which have to be checked for the design, namely for the serviceability limit state (SLS) and the ultimate limit state (ULS). The structure must remain functional for its intended use subjected for normal loadings for SLS. The ULS represents the load for which the structure must not collapse. The load factors for SLS are:  for all loads. The load factors for the ULS are:  Permanent loads: o 1.35 for self-weight when this is the only present load o =1.2 for self-weight, ground pressure, and ground water pressure (unfavourable) o =0.9 for self-weight, ground pressure, and ground water pressure (favourable)  Variable loads: o =1.5 for extreme wind, temperature, and traffic loads o =1.3 for current, ship induced waves, and ship induced current o =1.25 for water level differences and wind generated waves.  Special loads: o =1 for all special loads.

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P.3 LOAD COMBINATIONS

Four different load combinations are identified which are an open, a closed, a maintenance, and installation load case. The loads which act at the gate are enumerated per load combination:  open position o water pressure o self-weight o sediments o current  closed position o self-weight o water pressure o current  maintenance position o self-weight o wind  installation of the gate o self-weight o accelerations o wind.

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Q. Waves and hydrostatic pressure

Q.1 WAVE HEIGHT

Based on: (HKV Lijn in water, 2007)& (Rijkswaterstaat RIZA, 2007).

Wave heights and periods generated by the wind are calculated for the load and strength calculation of the weir and the weir gates. The formulas of Bretschneider are used for the determination of the significant wave height and significant wave period. The applied formulas are presented in Equation Q-1 and Equation Q-2.

( ( ) )

Equation Q-1 Calculation of the significant wave height using Bretschneider

( ( ) )

Equation Q-2 Calculation of the significant wave period using Bretschneider

In which:

Hs = significant wave height [m] u = wind speed at 10 metres height [m/s] g = gravitational constant [m/s2] F = effective fetch [m] d = water depth [m]

v1 = auxiliary constant without physical meaning [-]

( ( ) )

v2 = auxiliary constant without physical meaning [-]

( ( ) )

The fetches are determined for one location at the bank for one wind direction. The central direction is chosen along the wind direction, so the angle between the wind direction and the measured fetch is 0°. The wind does not blow constant from the same direction but deviates from the central direction of 0°. An overview of the input parameters for the calculation is given in Figure Q-1 and the equation for the calculation of the effective fetch is given in Equation Q-3. This method is applied for 3 locations, namely at the middle of the weir and at the two corners of the weir as presented in Figure Q-2, Table Q-1, Table Q-2, and Table Q-3. The largest effective fetch is determined for the south west corner of the weir and has a length of 381 metres. The wave height at the weir for this fetch and for a design wind speed of 14m/s is 0.2 metres and the wave period is 1.68 seconds. The wave heights are calculated using Mathcad.

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Also a calculation is made using the program Hydra-M which calculates the significant wave height and significant wave period for a high water conditions for which the floodplains are flooded. A significant wave height of 0.27 metres and a significant wave period of 2.01s are determined using this program. The gates are lifted during high water and do not experience a wave of 0.27m height. Therefore a significant wave height of 0.22 metres is used for the design.

Figure Q-1 Determination of the fetch

∑

Equation Q-3 Calculation of the effective fetch (HKV Lijn in water, 2007)

Figure Q-2 Determination of fetches

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Table Q-1 Fetch calculation at the middle of the weir

Wind direction South-East location 1 R(α) x cos2 2 Α cos α cos α R(α) α Fe [°] [m] [m] -42 0.743 0.552 90 50 -36 0.809 0.654 103 67 -30 0.866 0.750 122 91 -24 0.914 0.835 149 124 -18 0.951 0.904 197 178 -12 0.978 0.956 292 279 -6 0.995 0.990 581 575 0 1.000 1.000 1730 1730 6 0.995 0.990 581 575 12 0.978 0.956 292 279 18 0.951 0.904 197 178 24 0.914 0.835 149 124 30 0.866 0.750 122 91 36 0.809 0.654 103 67 42 0.743 0.552 90 50 13.512 4461 330

Table Q-2 Fetch calculation at the north east corner of the weir

Wind direction South-East location 2 2 R(α) x cos Fe Α cos α cos2 α R(α) α [°] [m] [m] -42 0.743 0.552 182 100 -36 0.809 0.654 207 135 -30 0.866 0.750 243 182 -24 0.914 0.835 299 250 -18 0.951 0.904 394 356 -12 0.978 0.956 585 560 -6 0.995 0.990 1520 1505 0 1.000 1.000 1830 1830 6 0.995 0.990 0 0 12 0.978 0.956 0 0 18 0.951 0.904 0 0 24 0.914 0.835 0 0 30 0.866 0.750 0 0 36 0.809 0.654 0 0 42 0.743 0.552 0 0 13.512 4919 364

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Table Q-3 Fetch calculation south west corner of the weir

Wind direction South-East location 3 R(α) x cos2 2 Α cos α cos α R(α) [m] α Fe [°] [m] [m] -42 0.743 0.552 0 0 -36 0.809 0.654 0 0 -30 0.866 0.750 0 0 -24 0.914 0.835 0 0 -18 0.951 0.904 0 0 -12 0.978 0.956 0 0 -6 0.995 0.990 0 0 0 1.000 1.000 1602 1602 6 0.995 0.990 1987 1967 12 0.978 0.956 584 559 18 0.951 0.904 393 355 24 0.914 0.835 299 250 30 0.866 0.750 243 182 36 0.809 0.654 207 135 42 0.743 0.552 182 100 13.512 5151 381

Q.2 WAVE PRESSURE

The wave pressure is determined by using Sainflou’s method (Molenaar, et al., 2008). Sainflou’s method is an approximation of the Stokes’ second order wave theory for complete reflection. The parameters used

for the Sainflou calculation are presented in Figure Q-3. The water level at the structure increases with h0 which is calculated using Equation Q-4.

Equation Q-4 Height increase of the middle level

In which:

h0 = height increase of the middle level [m]

Hi = wave height of an incoming wave [m] k = the wave number of the incoming wave [m-1] d = water depth [m]

The Sainflou method and the Stokes’ second order theory results to the same maximum pressure at middle water level and at the river bed. Equation Q-5 is used for the determination of the wave pressure at middle water level and Equation Q-6 is used for the determination of the wave pressure at the river bed.

Equation Q-5 Wave pressure at the middle water level of a wave

Equation Q-6 Wave pressure at the river bed

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Figure Q-3 Sainflou wave pressure

The values of the water level increase, p0, and p1 are calculated using a Mathcad model. The results of this calculation are shown for a depth of 7.2 metres and 8 metres in Table Q-4.

Table Q-4 Wave pressure at the gate

Parameter Result d=7.2m Result d=8m Dimension h0 0,01 0,01 m p0 0,16 0,13 kN/m2 p1 2,18 2,18 kN/m2

Q.3 HYDROSTATIC PRESSURE

The upstream and downstream water pressure linearly increases with the water depth. The design water level is +5.00m NAP and a the foundation is located at -2.2m NAP resulting in a water column of 7.2 metres. The bottom of a 8 metres high gate is located at -3m NAP resulting in a water column of 8 metres. A depth of 7.2 metres is used for the preliminary design and a depth of 8 metres is used for the weir gate calculations. A water level of -0.5m NAP is used for the downstream side of the weir. This water level is present for a discharge lower than -700m3/s at Lobith (Rijkswaterstaat, 2012) which occur less than 0.01% a year (ARCADIS, 2010). The mentioned water levels are presented in Figure Q-4. The hydrostatic pressure is calculated using Equation Q-7 and resulting pressures are presented in Table Q-5

Equation Q-7 Hydrostatic pressure

In which: r = density of water [kg/m3] g = gravitational constant [m/s2] d = depth [m]

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Figure Q-4 Design water levels

Table Q-5 Hydrostatic pressure at the gate

Result d=7.2m Result d=8m Side kN/m2 kN/m2 Upstream 72 80 Downstream 17 22

Q.4 LOAD AT THE GATES

The wave pressure, hydrostatic pressure, and the total pressure at the gates over depth are presented in Figure Q-5. The load per running metre is determined by integrating the pressure over the depth. Safety factors are included to account for determining the design load. The resulting loads are presented in Table Q-6 which are obtained by integrating the pressures over the depth as presented in Equation Q-8. Torsion and moments are neglected for the preliminary design and included in the gate design (appendix U)

∫ ∫ ∫

Equation Q-8 Load at the gate

In which:

pup = upstream hydrostatic pressure [kN/m2]

pds = downstream hydrostatic pressure [kN/m2]

pwave = wave pressure [kN/m2] d = depth [m]

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Figure Q-5 Pressures at the gate

Table Q-6 Loads at the gate

Load for Load for Safety Design load Design load Load d=7.2m d=8m factor d=7.2m d=8m kN/m2 kN/m2 [-] kN/m2 kN/m2 Waves 8 9 1,25 11 12 Upstream water 254 314 1,25 318 392 level Downstream 14 31 0,9 13 28 water level

Q.5 MATHCAD SHEETS

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R. Gate weight and gate costs

The total width of the opening of the gate should be at least 100 metres and smaller than 140 metres which is determined in appendix O. The total width could be divided in several smaller openings. The opening should be larger than 25 metres for recreational boating, larger than 29 metres for one lane commercial shipping, and larger than 41 metres for two lane commercial shipping. A model is made for comparing the different opening widths and the gate weight of a steel and fibre reinforced polymer (FRP) gate which is described in this appendix.

R.1 ESTIMATION OF GATE WEIGHT (GIRDER ON TWO SUPPORTS)

The gate is assumed to be pin supported at both abutments. The load resulting from the water level difference and the waves is schematized as a distributed line load q as presented in Figure R-1. The load q is calculated in Q.4. The distributed load is applied at the middle of the gate to simplify the calculation. This shifted resultant force causes a torsional moment at the gate. For now, the torsional moment is not taken into consideration for the estimation of gate weight because a sufficient accurate calculation is obtained without torsion for this design level (Kooij, 2012). Therefore, the torsional moment is taken into account for the more detailed calculations presented in appendix U.

Figure R-1 Estimation of gate weight (girder on two supports)

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R.1.1 CALCULATION METHOD OF THE GATE WEIGHT

Two calculations methods are used for the determination of the gate weight. The first calculation uses the maximum stresses for SLS and ULS and the second calculation uses the deflection limit. The calculation method for the maximum stresses is presented in R.1.1.1 and the calculation method for the maximum deflection is described in R.1.1.2.

R.1.1.1 CALCULATION FOR MAXIMUM STRESSES

The maximal moment for the gate is located at mid span and is equal to Equation R-1 for a simply supported beam.

Equation R-1 Maximum moment for a simply supported beam

Figure R-2 Gate schematisation

The gate has to be strong enough to resist this moment. Therefore, the upstream and downstream slab with a width ‘a’, thickness ‘t’, and the height ‘d’ needs to be large enough to bear the tensile (downstream slab) and compressive (upstream slab) forces. The variables ‘a’, ‘t’, and ‘d’ are presented in Figure R-2. The tensile and compressive forces for these slabs are calculated using Equation R-2. In which ‘a’ represents the distance between both slabs as indicated in Figure R-1 and Figure R-2.

Equation R-2 Bending forces

The minimum steel or FRP surface area for a design stress sd is calculated using Equation R-3.

Equation R-3 Minimum surface area of a slab

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The weight of the slab is estimated by multiplying the minimum surface area by the density of the material. The gate has two slabs (one upstream and one downstream) and trusses which connects both slabs and a percentage of 30% is used for taking into account the weight of the trusses. The total weight of the gate is calculated using Equation R-4.

( )

Equation R-4 Calculation of the gate weight

R.1.1.2 CALCULATION FOR MAXIMUM DEFLECTION

The second limit used for the calculation of the mass of the gate is the deflection limit. Two deflection limits are included in the calculation, namely:

 which is used for FRP lifting gates and lock doors

 which is used for deflections of bridges.

A sub-model is made using Mathcad for determining the minimal moment of inertia. The dimensions of the gate are calculated using the minimal moment of inertia using Steiner’s rules; the moment of inertia of the web is neglected in the calculations. The deflection at mid-span of a FRP gate is presented in Equation R-5. The derivation of this formula is presented in appendix T.6. The deflection due to shear is about 6% of the total deflection for a simply supported beam as presented in appendix T.6. Equation R-5 is simplified by replacing the expression for shear deflection by an extra bending moment deflection of 6%. A formula for the moment of inertia is obtained by rearranging Equation R-5 and using this assumption. The result is presented in Equation R-6.

Equation R-5 Deflections of a two sided supported FRP gate

In which: v = deflection [m] q = distributed load [kN/m] l = span [m] E = tensile modulus [kN/m2] I = moment of inertia G = shear modulus [kN/m2] A = cross sectional area [m2]

Equation R-6 Minimal moment of inertia for a given deflection

The dimensions of the upstream and downstream plate are determined using the Steiner’s rule which is presented in Equation R-7.

( ( ) )

Equation R-7 Steiner’s rule

In which: t = plate thickness [m] d = plate height [m] a = gate width [m]

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An expression for the minimal plate thickness is obtained by rewriting Equation R-7. The result is presented in Equation R-8.

Equation R-8 Minimal plate thickness for a given moment of inertia

The volume of the gate is calculated by multiplying the plate thickness by the height and the span and adding a 30% of extra mass for taking into account the weight of the trusses. The used formula for the mass is presented in Equation R-4.

R.1.2 STEEL GATE WEIGHT

The used parameters for the calculation of a steel gate are presented in Table R-1.

Table R-1 Parameters for the calculation of a steel gate

Parameter symbol Unit Value 3 Density of steel steel kg/m 7850

Height of gate agate m 3 2 maximum steel stress max N/mm 200

Tensile modulus Esteel GPa 210

The weight of the gate is calculated using the presented parameters and equations for a ranging span. The results of this calculation is presented in Figure R-3. A start weight of 50 tons is used for including the weight of the connection of the gate to the abutments, which stems from the calculation of the vertical lifting gates of the new sluice complexes of the Afsluitdijk (ARCADIS, 2010). The result represented by the red line is verified using the design formulas presented in the book: ‘Design of hydraulic gates’ (Erbisti, 2004). The weight of three gate types are calculated using the head over the gate (H), the height of the gate

(h), and the span (Bgate). The formula for a fixed wheel gate is presented in Equation R-9, the formula for a submerged segment gates is presented in Equation R-10, and the formula for a flap gate is presented in Equation R-11.

( )

Equation R-9 Fixed wheel gate weight (Erbisti, 2004)

( )

Equation R-10 Submerged segment gate (Erbisti, 2004)

Equation R-11 Flap gate weight (Erbisti, 2004)

The formula approaches the estimated gate weight for the range 20meters-50meters which is within the application range of the new gates of the weir. Therefore the estimated values for gate weight are used for further design given the match of the formulas of Erbisti.

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Figure R-3 Steel gate weight as a function of span

R.1.3 FRP GATE WEIGHT

Based on: (Stichting CUR, 2003), (SMOZ Project "Kunststof Sluisdeur", sd), (Kolstein, 2008), (FiberCore- Europe, 2012), and (Snijder, 2012).

Now the weight of a FRP gate is calculated. The heaviest parts of a FRP gate are the laminates which bear the tensile and compressive forces generated by the moment. A sandwich structure is equipped with a lower density core which is not able to bear the tensile and compressive forces; the core is only used as spacer to increase the moment of inertia of the sandwich and for bearing the shear forces. The weight of the core is 10 times lower with respect to the laminate and is therefore neglected for the determination of the gate weight. An appropriate laminate for gate design is an anisotropic laminate with 50 vol% glass fibre for which 55% of the fibres are aligned with the main load direction and 15% in other directions as presented in Figure R-4. The used parameters for the calculation are presented in Table R-2.

The tensile modulus of the main load bearing direction (E1) is used for the estimation of the gate weight of a FRP gate. Maximum strains are defined for the determination of the maximum stresses. A strain limit of 1.2% is set for ‘normal’ structures which are not in contact with water for SLS and ULS calculations. A strain limit of 0.27% is set for structures which are in contact with water in order to avoid micro cracking (Stichting CUR, 2003).

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Figure R-4 Fibre distribution of a 055/9015/+4515/-4515 laminate

Table R-2 Nominal stiffness properties of the used laminate (Stichting CUR, 2003)

Parameter Value Unit

E1 25.8 GPa

E2 15.9 GPa

G12 5.6 GPa 0.32 -

Table R-3 Parameters for the calculation of the weight of a FRP gate

Parameter symbol Unit Value 3 Density of the resin venylesther kg/m 1089 3 Density of glass fibre glassfibre kg/m 2443

Volume fraction Vf % 50 Height of gate a m 3

Serviceability limit strain SLS % 0.27 Ultimate limit strain % 1.2

The stresses linearly increases with strain and therefore Hooke’s law is applied for the determination of the maximum stresses for SLS and ULS. The nominal stress for SLS is presented in Equation R-12 and the nominal stress for the ULS is presented in Equation R-13.

69.66

Equation R-12 Stress for SLS

Equation R-13 Stress for ULS

The stresses have to be divided by a material factor and a conversion factor to determine the design value. The following holds for the material factor:

In which:

= 1.35 [-] Partial material factor for taking into account the uncertainties for producing the material with the exact material properties.

Dependents on the production methods presented in Table B-17. A post-cured vacuum injection laminate is used for the fabrication of the gate

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Table R-4 Partial material factor m,2

Partial material factor m,2 Production method Post cured laminate Non-post cured laminate Spray up 1.6 1.9 Hand laminating 1.4 1.7 Vacuum or pressure injection 1.2 1.4 Filament wounding 1.1 1.3 Prepreg 1.1 1.3 Pultrusion 1.1 1.3

The following holds for the conversion factor:

In which:

= 1.1 [-] Partial material factor for taking into account the temperature effects.

= 1.0 [-] for dry environment, 1.1 [-] for changing humidity conditions, and 1.3 [-] for prolonged contact with humid environment (surface water, groundwater, etc.).

= Creep factor in which: t = the time span of the load in hours (50 years). n = exponent dependent on the fibre reinforcement. 0.01 for UD lamella, 0.04 for woven lamella, and 0.1 for mat lamella.

= 1.1 [-] For fatigue.

Table R-5 gives an overview of the application of the different conversion factors for six limit states. The factors are calculated using Mathcad. The results per load case are presented in the last row of Table R-5.

Table R-5 Conversion factors

The design stresses for SLS and ULS are presented in Equation R-14 and Equation R-15.

[ ] [ ] [ ]

Equation R-14 Design stress for the first crack calculation (SLS)

[ ] [ ] [ ]

Equation R-15 Design stress for the strength calculation (ULS)

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The maximum design stress for the first crack calculation (SLS) is very low and question marks have been raised about the application of the material factor and the conversion factor. This has been discussed with Kasper Snijder, composite structures engineer of the company FiberCore. He stated that the material factor and conversion factor are included in the strain reduction and the nominal stresses calculated for the 0.27% strain limit can be used as design strain. A safety factor of 13 would otherwise be taken into account by including the strain reduction and the material and conversion factor which is indicated in Equation R-16. The maximum stress determined for the 0.27% strain without material and conversion factor is also used for the design of the Spieringsluis (SMOZ Project "Kunststof Sluisdeur", sd). This has to be verified by experiments and a more detailed literature research because this assumption is in contrast with the recommendations of the CUR96.

Equation R-16 Safety factor for stress reduction, material factor, and conversion factor

The calculation described in appendix R.1 is performed using the stresses presented in Equation R-12, Equation R-14, and Equation R-15 for the mentioned strain limits. The resulting gate weight for the SLS tensile stress with material and conversion factor, SLS stress without material and conversion factor, ULS stress with material and conversion factor, and for the deflection limits are presented in Figure R-5 for a gate width of 3 metres (a=3m).

Figure R-5 FRP gate weight depending on the span

The mass of a FRP gate calculated for the ULS maximum stress and the mass for a FRP gate calculated for the SLS maximum stress without material and conversion factor is lower with respect to the mass of a FRP gate calculated for the deflection limits according to Figure R-5. A lower mass corresponds with a lower moment of inertia, so the deflections of the FRP gate calculated for the ULS maximum stresses and the FRP

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gate for the SLS maximum stresses are higher and are therefore not a suitable solution. The gate mass of a FRP gate calculated for the SLS with material and conversion factors is larger with respect to the gate mass calculated for the deflection limits and has therefore smaller deflections with respect to the gates calculated for the deflection limits and could be a suitable solution.

R.1.4 COMPARISON OF A STEEL AND FRP GATE

The weight of the gate is calculated for a gate width of 3 metres. The results for a varying span of a FRP gate and a steel gate are presented in Figure R-6. The weight of a FRP gate for a deflection limit of 1/150 is the lowest for every span and the weight of a FRP gate for a deflection limit of 1/300 is lower for spans smaller than 40 metres with respect to a steel gate. The weight of a steel gate is the highest for spans smaller than 40 metres.

Figure R-6 Gate weights of steel and FRP gates

The costs of a gate are determined using the estimated mass of a gate and key ratios. The used key ratios are based on a consult with a contracting expert (Schoonhoven, 2012). The costs of a steel gate and FRP gate for a variable width are presented in Figure R-7; the following key ratios are used (price level September 2012):  €6.5 per kg for a steel gate  €12 per kg for a FRP gate

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Figure R-7 Costs of steel and FRP gates

R.2 MATHCAD SHEETS

The calculations are performed using the program Mathcad. The Mathcad sheets are not printed but digitally provided on a CD or USB drive. Also a pdf file for the Mathcad sheets is provided at the repository of the TU Delft (repository.tudelft.nl)

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S. Weir foundation

Two foundation configurations are designed for weir Culemborg which are described in the following enumeration:  The sill and the pylons are connected to each other for the first configuration. The loads at the gate and the pylons are transferred via a continuous foundation plate to the subsoil. The continuous foundation transfers the loads acting at the pylons and gates and the own weight to the subsoil. An impression of this foundation type is presented in the left figure of Figure S-1.  The hydraulic load at the gates is transferred via the bearings to the pylons for the second foundation type. The pylons bears the forces acting on the gates and the forces acting on the pylons. The foundation of the pylons transfers these loads and the own weight into the subsoil. The horizontal load at the sill generated by the difference in groundwater pressure and the load generated by the own weight of the sill which is transferred by the sill foundation into the subsoil. An impression of this foundation type is presented in the right figure of Figure S-1.

Figure S-1 Continuous foundation (left) and a separate foundation of the pylons and sills (right) (PIANC, 2006)

The ground body on which the weir is founded consists of mainly sand as presented in the CPT tests in appendix C.4.5. A minimum CPT value of 10MPa is obtained at a level of -7m NAP on which the structure is founded. Therefore no piles are needed for obtaining enough bearing capacity of the weirs. However, when the dimensions of the weir becomes large, it is more preferable for founding the weir on piles (Voortman, 24). The weir foundation has several limit states which are elaborated in S.1. The results of the calculations based on the design limits are presented in S.2.

S.1 LIMIT STATES

Based on: (Molenaar, et al., 2008) & (Voorendt, et al., 2009)

Sliding stability First of all, the weir may not slide away due to the horizontal load acting at the structure. Therefore, the frictional force has to resist the resulting horizontal load. The frictional force is calculated by the sum of the vertical forces multiplied by a friction coefficient as presented in Equation S-1. The resulting horizontal force must be smaller than the friction coefficient multiplied with the total vertical force.

Equation S-1 Sliding stability

In which:

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= Resulting horizontal force [kN] = Friction coefficient [-]

= ( )

In which ϕ is the angle of internal friction. = Resulting vertical force [kN]

Overturning stability The forces acting at the structure generates an overturning moment. The adhesive and cohesive properties of the sand body are poor and therefore, no tensile forces can be present under the foundation. Therefore, the resultant force has to remain within the core of the structure to prevent tensile stresses. The core is defined as the area extending over 1/6th of the width from the centre of gravity. So, the check presented in Equation S-2 must be executed to determine whether a tensile force is present.

Equation S-2 Overturning stability

Soil stress The vertical effective stresses should not exceed the maximum bearing capacity of the subsoil. An exceedance of the maximum effective stress results in a collapse of the structure. The maximum stress acting on the soil is calculated using Equation S-3.

Equation S-3 Maximum stress

In which: = Resulting vertical force [kN] = Area of the bottom plate [m2] = Total overturning moment [kNm] W = Width of the considered element [m] H = Height of the considered element [m]

The maximum allowable stress is calculated using the Brinch Hansen method. This method includes the cohesion, surcharge including soil coverage, and the capacity of the soil below the foundation as presented in Equation S-4. A value of 500kN/m2 could be used as a first estimate.

Equation S-4 Maximum allowable stress Brinch Hansen (Molenaar, et al., 2008)

In which: = Cohesion [kN/m2] = Effective stress at the depth at the foundation surface [kN/m2] = Effective volumetric weight [kN/m3]

= Factors for the bearing force [-]

= Factors for the shape of the foundation [-]

= Factors for the horizontal load [-]

Piping A groundwater flow underneath the structure can develop when a head over a structure is present and the seepage distance is too small. Two empirical formulas are used to calculate the minimum seepage

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distances which are the formulas of Bligh and Lane. The formulas of Bligh and Lane are presented in Equation S-5 and Equation S-6.

Equation S-5 Bligh piping criteria

Equation S-6 Lane piping criteria

In which:

= Horizontal seeping distance [m]

= Vertical seeping distance [m]

= Bligh piping coefficient = 12 (course sand; formation of Kreftenheye)

= Lane piping coefficient = 5 (course sand; formation of Kreftenheye) = Water level difference [m]

S.2 FULLY DAMMED OPERATION

An upstream water level of +5.00m NAP is present for a fully dammed operation (appendix J.6) in combination with a wave height of 0.2m (appendix Q.1). The most unfavourable downstream water level is -0.5m NAP (Table C-5) for which the weir has to remain stable. The maximum value for unity checks is 1. However, a maximum of 0.8 is used in order to take into account the rough schematisation of the weir for the unity check for each limit state which is presented in Equation S-7. The weight of the foundation or pylon could decrease when a stilling basin or a more detailed design of the pylon is made in a sequential design level.

Equation S-7 Unity check for sliding stability

S.2.1 CONTINUOUS FOUNDATION SLAB

The continuous foundation slab is checked for the limit states described in S.1. The foundation slab is schematised as a rectangular plate without recesses. A stilling basin and recess for the gate have to be included in more detailed stability calculations which have to be performed in a sequential design level. A sheet at the upstream side of the weir is included in the design which reduces the upward water pressure and increases the piping length. The stability of the structure is verified per running metre. The weight of the pylon is equally spread over the total width of the opening and pylon. The top of the pylon is located at +7.5m NAP which is 1 metre higher than the design water level for weir Hagestein (Rijkswaterstaat, 2012). A factor 0.5 of the total volume is taken into account for the weight of the pylon, so

the weight of the pylon is: . A drawing of the weir and the forces acting at the weir are presented in Figure S-2. The calculation of the forces and moments are included in the Mathcad sheets. The results of a continuous foundation slab are presented in Table S-1.

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Figure S-2 Continuous foundation

Table S-1 Dimensions of a continuous foundation slab

Length Element [m] Width of the foundation 45 Thickness of foundation 3 Length of sheet pile 15 Location of the gate from the middle of the foundation (negative value means -2.5 left from the middle; positive value means right from the middle)

S.2.2 SEPARATE FOUNDATIONS FOR PYLONS AND SILLS

Two calculations are performed, namely one stability calculation for the sills and one stability calculation for the pylons. First the results of the sill beam are presented. The forces taken into account for this calculation are presented in Figure S-3. The sill is schematised as a rectangular box. The sheet pile is schematised as positioned at the tip of the foundation, so the upward hydrostatic pressure at the upstream side of the foundation is he reduced hydrostatic pressure. The loads at the gates are transferred via the hinges to the pylons, so the sill does not have to bear the forces generated by the water level difference. The resulting dimensions are presented in Table S-2.

Figure S-3 Sill foundation

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Table S-2 Dimensions of the sill beam Length Element [m] Width of the foundation 25 Thickness of foundation 4 Length of sheet pile 20 Location of the gate from the middle of the foundation (negative value -3.5 means left from the middle; positive value means right from the middle)

The pylon of the weir transvers the horizontal forces acting at the weir and the horizontal forces acting at the gates to the subsoil. The pylon is placed on a foundation slab. The foundation slab transvers the horizontal and vertical loads to the subsoil. A weight reduction factor of 0.5 is taken into account for determining the weight of the pylon which is placed on the foundation slab. The geometry of the pylons is presented in Figure S-4. The calculations of the forces and the moments are presented in S.4. The resulting dimensions are presented in Table S-3.

Figure S-4 Pylon foundation

Table S-3 Dimensions of the pylon

Length Element [m] Width of the foundation 55 Width of the pylon 45 Thickness of foundation 3 Length of sheet pile 20 Location of the gate from the middle of the foundation (negative value -3 means left from the middle; positive value means right from the middle)

S.2.3 COMPARISON WITH THE PRESENT WEIRS

The weirs of Driel, Amerongen, and Hagestein are built according to the second foundation technique. A sill beam is aligned from span to span following the radius of the visor gate. The pylons bears the forces acting on the pylons and the gates. The width of the weir pylon differs from the design of weir Hagestein due to the absence of the accurate discharge control in the pylons. However the length is of the same order as the present weirs in the Nederrijn. The sill beam is 3 metres thick and about 15 metres wide.

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S.3 CONCLUSION

The stiffness of the foundation slab of the part near the pylons of a continuous foundation slab differs considerably with respect to the stiffness of the foundation in between the pylons when an equal reinforcement ratio is used. The difference in stiffness causes the pylon foundation to transfer more load to the subsoil than the foundation located in between the pylons. The stiffness of the foundation located in between the pylons has to be increased with respect to the stiffness of the pylon to obtain a continuous foundation slab which transfers the same amount of load per running metre, which is costly. Therefore the second foundation system is a better option with respect to the continuous foundation slab. The dimensions the pylon foundation are quite large. Therefore, one could choose for a piled pylon foundation to reduce the dimensions of the structure. This option is more cost effective as a large foundation as presented in Figure S-4 (Voortman, 24). This is also applied for the Ems barrier submerged segment gate which is presented in Figure S-5. The elaboration of the piled foundation is beyond the scope of this graduation research, so only a recommendation is done for the application of a piled foundation for the pylon.

Figure S-5 Cross section of the Ems barrier

S.4 MATHCAD SHEETS

The calculations are performed using the program Mathcad. However the Mathcad sheets are not printed but digitally provided on a CD or USB drive. Also a pdf file with Mathcad sheets is provided at the repository of the TU Delft (repository.tudelft.nl)

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T. Global weir gate design

T.1 ROTATABILITY OF THE GATE

The gate rotates from the bottom into upright position to control the discharge and water level. The upstream water could be stored at the flat side of the gate or the convex side as presented in Figure T-1. The best option is storing the upstream water at the convex side, because:  The resultant force from the water pressure goes through the point of rotation, so the moment of the gate is zero (Erbisti, 2004).  The convex side is under compression which is more stable for local buckling due to the convex shape with respect to a flat plate under compression (Kooij, 2012).

Therefore the convex side of the gate has to be rotated in the upstream direction as presented in the lower left figure of Figure T-1.

Figure T-1 Rotating directions of a submerged segment gate

T.2 MAXIMUM GATE DEFLECTIONS

No deflection limits are developed for hydraulic structures, because every structure is unique. The deflection limit developed for the project ESA (vertical lifting gate) is 1/150 of the span; which results in a maximum deflection of 0.27 metres for a gate with a span of 41 metres (ARCADIS & DHV, 2010). The deflection limits for a submerged segment gate are more stricter because a watertight connection between the rotating gate and the foundation has to be obtained. Normal profiles can cope with a deflection till 100mm in the middle of the span (Sloten, 2012). Therefore a deflection limit of 100mm is used for design.

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T.3 GATE GEOMETRY

The upstream water level which the gate has to retain is +5.00m NAP and the sill level is located at -2.2m NAP which results in a minimal height of 7.2 metres. However, an extra height needs to be present for the watertight closure of the gate. Therefore a gate height of 8 metres is used. The radius of the gate is 6 metres, which is determined using the formula for the gate height presented in Figure T-2.

Figure T-2 Determination of the gate height and radius

The width (distance between the upstream and downstream plate) of the gate should be in the order of 3 metres for remaining within the maximum deflection limit as presented in Figure T-3. The deflections presented in Figure T-3 are calculated for a simply supported beam for a maximum bending moment.

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Figure T-3 Deflection of FRP gates

T.4 GATE CROSS SECTIONS

The gate cross section can be composed in different manners. In this section, a description is made for several cross sections and one gate cross section is chosen for further design

A homogeneous cross section The space in between the front and the back side of the gate could be filled up with the same material creating a homogeneous cross section as presented in Figure T-4. However, producing a 3 metre thick FRP beam is not possible and a lot of material is needed. The material placed in between the front and the back side is located close to the neutral axis and does not significantly contribute to the moment of inertia which makes it an inefficient design. Therefore, a homogeneous cross section is not a suitable option.

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Figure T-4 Solid gate

A gate with one front flap stiffened with beams A large stiffness has to be obtained to minimize the deflections to 0.1 metres. Therefore a height of about 3 metres is needed. Applying I beams with a maximum width of 3 metres results in the gate geometry presented in Figure T-5. However this is not an optimal solution because:  It is not practical to create I beams for FRP structures.  Sediments will be trapped in between the I beams which is unfavourable.  The flow at the backside of the gate is hampered by the beams, which is unfavourable. Therefore, a plate stiffened with I beams is not a suitable option.

Figure T-5 Gate stiffened with I beams

A gate with one front slab and stiffened with a truss (Figure T-6) A truss could be applied at the downstream side of the gate to increase the stiffness. However, the connections of a FRP truss are costly and hard to fabricate using FRP. Therefore a closed geometry with fewer connections is preferred above a truss for a FRP gate. Furthermore, the flow at the backside of the gate is more streamlined for a closed geometry with respect to this type. So a gate with one front slab which is stiffened with a truss is not a suitable option

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Figure T-6 Gate stiffened with a truss

A gate with a front and a back flap The disadvantages of a gate with a front slab with I beams and the disadvantage of a homogeneous cross section are solved for this geometry. A closed cross section is obtained which is presented in Figure T-7. The flow is not hampered by trusses or beams and sediments are not trapped anymore in between the profiles. Furthermore the excessive material for a homogeneous cross section is saved. However the cross section is not able to bear the shear force. So, this cross section is an improvement of the previous cross sections but can be further improved.

Figure T-7 Gate composed of a front and back flap

A gate with ‘stiffening ribs’ Stiffening ribs are added to the cross section to increase the shear capacity. The stiffening ribs connect the front and the back side of the gate as presented in Figure T-8, and for the Nakdong weir in Figure T-9. The front and the back sides bears the tensile and compressive forces generated by moment and the stiffening ribs bears the shear. So, this gate type is a suitable solution and is used for further design.

Figure T-8 Gate with stiffening ribs

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Figure T-9 Nakdong weir equipped with submerged segment gates strengthened with shear elements (kgal, 2012)

T.5 LOADS AT THE GATE

The gate has to resist the upstream hydrostatic pressure, the downstream hydrostatic pressure, and the wave pressure which are presented in Figure T-10. The loads are acting at the curved upstream side and at the flat downstream side.

Figure T-10 Gate schematisation

The upstream and downstream loads are schematised as a resultant upstream and a resultant downstream load acting at 1/3th of the height of the gate. The upward hydrostatic pressure is neglected for these calculations because the horizontal load has a larger impact on the design because it acts in plane with the weakest axis of the gate. Therefore the middle figure of Figure T-10 is used as schematisation of the weir gate.

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A resulting force is obtained by summing up the downstream and upstream resultant forces. The force is shifted to the middle of the gate to simplify sequential calculations. A resultant torsional moment appears due to the shifted force, which has to be beard by the gate cross section. It is assumed that the torsional moment is of minor importance with respect to the resultant force because the gate has a large torsional shear stress capacity due to the enclosed profile. So, the gate is designed for the maximum moment generated by the total resultant force acting as a line load at the middle of the gate. The resulting design is checked for the torsional moment capacity in U.1.4. The torsional moment is calculated below.

The upstream resultant distributed load is:

[ ] [ ] [ [ ]]

The downstream resultant distributed load is:

[ ] [ ] [ [ ]]

The working line of the resultant force is: [ ] [ ]

The working line has to be shifted over =1.1m which results in a torsional moment for SLS of:

[ ]( )=332 kNm

T.6 MECHANICAL SCHEME OF THE GATE

The loads have to be transferred from the gate to the rotational disks and to the pylons of the weir complex. Several options are possible for the load transfer from the gate to the rotational disks and to the pylons. The mechanical schemes for the gate, which could be applied are discussed in T.6.1 . The deflection of the gate for every mechanical scheme is discussed in T.6.2. No detailled design is yet made of the gate. Therefore, the calculations are performed using the properties of a gate for a span of 41 metres, a height of 7.2 metres, and a width of 3 metres which are determined in appendix R for the determination of the weight and the costs.

T.6.1 MECHANICAL SCHEME OF THE GATE

Four mechanical schemes are identified which are also included in the main report. The mechanical schemes are:  a pin supported connection at both sides of the gate o Both disks do have a pin supported connection with the abutments.  a pin supported and fixed connections of the gate o Both disks do have a pin supported connection with the abutments.  a fixed connection at both sides of the gate o Both disks do have a pin supported connection with the abutments.  a fixed connection at both sides of the gate. o One disk is pin supported, the other disk sliding supported.

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Figure T-11 Mechanical schemes of the gate

The mechanical schemes of the gates are presented in Figure T-11. The deflections of the disks are neglected with respect to the deflection of the gates. This is justified due to the small radius of the disk (~6 metres) with respect to the large span of the gate (~41 metres). Therefore, the gate is modelled as a simply supported beam, a beam which is pin supported at one side and fixed to the other side, and a beam which is fixed at both supports as presented at the right side of Figure T-11. The connections of the gate and the rotating disks of the 4th gate type transfers the shear force only. No moments are present due to the sliding support. Therefore, the behaviour of the span is modelled as a simply supported beam. The verification of this assumption is presented in T.6.3

T.6.2 DEFLECTIONS OF THE GATE

The deflections of the gate could be calculated with use of the ‘Forget-Me-Nots’ for bending beams. These Forget-Me-Nots are applicable for bending beams with a negligible shear deformation like a steel beam or a concrete beam. However, the shear deformation of a fibre reinforced beam is not negligible (Stichting CUR, 2003). Therefore extended Forget-Me-Nots are derived for the mechanical schemes presented in Figure T-11. The standard Forget-Me-Nots are derived for an Euler-Bernoulli bending beam, the extended Forget-Me-Nots have to be derived for a Timoshenko bending beam. The following paragraphs are based on ‘an introduction to the analysis of slender structures’ (Simone, 2011) and ‘Toegepaste Mechanica Deel 2 Spanningen, vervormingen en verplaatsingen’ (Hartsuijker, 2007).

T.6.2.1 EULER-BERNOULLI & TIMOSHENKO BEAMS

The sign convention for an Euler-Bernoulli beam and a Timoshenko beam is given in Figure T-12. The positive x-axis is directed to the right and the positive y axis is directed downward. The positive moments, shear, curvature, deflection, and load are also presented in Figure T-12. The beam problems for an Euler- Bernoulli beam are solved with the assumption that the shear strain is negligible. Furthermore the beam has a straight longitudinal axis with a cross section which is symmetric over the y axis. The only unknown of an Euler-Bernoulli beam is the deflection v.

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The assumption that a plane section remains plane and normal to the deformed centreline is relaxed for the Timoshenko beam theory. Two unknown independent variables are available for this theory which are the transverse deflection v and the cross sectional rotation ϕ. ϕ is the rotation of the cross section with respect to the vertical axis. The extended sign convention is presented in Figure T-13.

Figure T-12 Sign convention for beam Euler-Bernoulli beam problems

Figure T-13 Extended sign convention for Timoshenko beam problems

An impression of the difference of shear deformation and bending deformation is presented in Figure T-14. The upper figure presents the shear deformation and the lower figure presents the bending deformation.

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Figure T-14 Shear deformation and bending deformation

The deflection for an Euler-Bernoulli bending beam is calculated using the 4th order differential equation presented in Equation T-1.

Equation T-1 Euler-Bernoulli beam differential equation

In which: E = Tensile modulus of the material [N/m2] I = Moment of inertia of the beam [m4] v = Deflection [m] q = Distributed load [N/m]

Four boundary conditions are needed to solve this differential equation. The boundary conditions available for the beam problem are:  the deflection v  the rotation ϕ o in which  the moment M o in which  the shear force V

o in which

The deflections for a Timoshenko beam are calculated with use of the 4th order differential equation presented in Equation T-2.

Equation T-2 Timoshenko beam differential equation

In which: E = Tensile modulus of the material [N/m2] I = Moment of inertia of the beam [m4] G = Shear modulus [N/m2] A = Cross sectional area [m2] v = Deflection [m] q = Distributed load [N/m]

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Four boundary conditions are needed to solve the differential equation. The boundary conditions available for the beam problem are:  the deflection v  the rotation ϕ

o in which

 the moment M

o in which ( )

 the shear force V

o in which

The deflections of an Euler-Bernoulli beam and a Timoshenko beam are calculated using Equation T-1 and Equation T-2. The results for the four mechanical schemes are presented in T.6.2.2.

T.6.2.2 DEFLECTIONS OF TIMOSHENKO BEAMS

The deflections of a simply supported beam, a beam which is pin supported at one side and fixed at the other side, and a beam which is fixed at both sides are calculated for an Euler-Bernoulli and Timoshenko beam. The derived formulas are determined and plotted for:  a span of 41 metres  a width of the gate of 3 metres  a height of the gate of 7.2 metres  a moment of inertia of 3.26 m4  a tensile modulus of 25.8 GPA  a cross sectional area of 1.45 m2  a distributed load of 248kN/m which is the SLS load used for the estimation of gate weight.

Simply supported beams The boundary conditions applicable for a simply supported beam are:  v=0m at x=0m for an Euler-Bernoulli and Timoshenko beam  M=0Nm at x=0m o which results in for an Euler-Bernoulli beam

o which results in for a Timoshenko beam

 v=0m at x=l for an Euler-Bernoulli and Timoshenko beam  M=0Nm at x=l o which results in for an Euler-Bernoulli beam

o which results in for a Timoshenko beam.

The deflection as a function of x for an Euler Bernoulli beam is given in Equation T-3 and for a Timoshenko beam in Equation T-4

( )

Equation T-3 Deflection as a function of x for a simply supported Euler-Bernoulli beam

( ) ( )

Equation T-4 Deflection as a function of x for a simply supported Timoshenko beam

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The maximum deflection is obtained at which results in the Forget Me Not presented in Equation

T-5 for a Timoshenko beam. The first term of Equation T-5 represents the bending deformation and the second term represents the shear deformation.

Equation T-5 Maximum deflection of a Timoshenko beam

This extended ‘Forget-me-Not’ equals the standard ‘Forget-Me-Not’ when the shear deformation is neglected. The CUR96 also proposed a deflection formula for a simple supported beam. However an error is made in the CUR96, because the units of the presented formula do not match. A quadratic relationship between the span and the shear deformation should be present according to Equation T-5, but the CUR96 gives a linear relationship between the span and the shear deformation. By replacing the linear relationship between the span and the shear deformation for a quadratic relationship the units matches and the same result as Equation T-5 is obtained. The deflections of an Euler-Bernoulli beam, the deflections of a Timoshenko beam, and the ‘Forget-Me-Not’s’ for a simply supported beam for the described geometry and loads are presented in Figure T-15.

Figure T-15 Deflections of a simply supported Euler-Bernoulli and Timoshenko beam

A beam which is pin supported at one side and fixed at the other side The boundary conditions applicable for a beam which is pin supported at one side and fixed at the other side are:  v=0m at x=0m for an Euler-Bernoulli and Timoshenko beam  M=0Nm at x=0 o which results in for an Euler-Bernoulli beam

o which results in for a Timoshenko beam

 v=0m at x=l for an Euler-Bernoulli and Timoshenko beam

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 ϕ=0Nm at x=l o which results in for an Euler-Bernoulli beam

o which results in for a Timoshenko beam

. in which ∫ for small angular rotations (Zenkert, 1995).

The deflection as a function of x for an Euler-Bernoulli beam is given in Equation T-6 and for a Timoshenko beam in Equation T-7.

( )

Equation T-6 Deflection as a function of x for a ‘pin supported-fixed’ Euler-Bernoulli beam

( ) ( )

Equation T-7 Deflection as a function of x for a ‘pin supported-fixed’ Timoshenko beam

Not the maximum but the deflection at mid-span is calculated using Equation T-8. The first term of Equation T-8 represents the bending deformation and the second term represents the shear deformation.

Equation T-8 Deflection at mid span for a’ pin supported-fixed’ Timoshenko beam

This extended ‘Forget-Me-Not’ equals the standard ‘Forget-Me-Not’ when the shear deformation is neglected. The deflections of an Euler-Bernoulli beam, of a Timoshenko beam, and the ‘Forget-Me-Not’s’ for a ‘pin supported-fixed’ beam for the prescribed geometry and loads are presented in Figure T-16.

Figure T-16 Deflections of a ‘pin supported-fixed’ Euler-Bernoulli and Timoshenko beam

A beam which is fixed at both sides The boundary conditions applicable for a beam which is fixed at both sides are:  v=0m at x=0m for an Euler-Bernoulli and Timoshenko beam

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 ϕ=0Nm at x=0m o which results in for an Euler-Bernoulli beam

o which results in for a Timoshenko beam

. in which ∫

 v=0m at x=l for an Euler-Bernoulli and Timoshenko beam  ϕ=0 Nm at x=l o which results in for an Euler-Bernoulli beam

o which results in for a Timoshenko beam

. in which ∫

The deflection as a function of x for an Euler-Bernoulli beam is given in Equation T-9 and for a Timoshenko beam in Equation T-10.

( )

Equation T-9 Deflection as a function of x for a ‘fixed-fixed’ Euler Bernoulli beam

( ) ( )

Equation T-10 Deflection as a function of x for a ‘fixed-fixed’ Timoshenko beam

The maximum deflection is obtained at which results in Equation T-11 for a Timoshenko beam. The first term of Equation T-11 represents the bending deformation and the second term represents the shear deformation.

Equation T-11 Deflection at mid span for a ‘fixed-fixed’ Timoshenko beam

The extended ‘Forget-Me-Not’ equals the standard ‘Forget-Me-Not’ when the shear deformation is neglected. The deflections of an Euler-Bernoulli beam, of a Timoshenko beam, and the ‘Forget-Me-Not’s’ for a ‘fixed-fixed’ beam for the prescribed geometry and loads are presented in Figure T-17.

Figure T-17 Deflections of a ‘fixed-fixed’ Euler-Bernoulli and Timoshenko beam

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T.6.3 VERIFICATION OF THE BEHAVIOUR OF THE 4TH MECHANIC SCHEME

The assumption that the fourth mechanical scheme behaves like a simply supported beam is verified using the program MatrixFrame. The deflection, the moments, and the shear forces are calculated using the differential equation and MatrixFrame for verifying this assumption. The calculation is performed for the same parameters of the previous section.

Figure T-18 The fourth mechanical scheme

The results of the MatrixFrame calculation are presented in Figure T-19. Only a maximum bending moment at mid span and maximum shear forces at the connections are present. The results of the Mathcad calculation are presented in Figure T-20 and Figure T-21. The shear forces and bending moments resulting from the MatrixFrame calculations equals the shear forces and bending moments resulting from the differential equation solved in Mathcad. The deflections of the gate calculated with MatrixFrame are presented in Figure T-19 and the deflections of the span calculated with the Mathcad sheet are presented in Figure T-22. The deflections calculated with MatrixFrame equals the deflections of an Euler-Bernoulli bending beam. So, the deflections calculated by MatrixFrame are smaller with respect to the deflections calculated using the differential equation of a Timoshenko beam which is indicated in Figure T-22. The horizontal displacement of the sliding support is 0.102 metres according to Figure T-19 which is also calculated using the Mathcad sheet by multiplying the radius of the disk with the rotation at the supports. The rotation at the supports is calculated using Equation T-12 for an Euler-Bernoulli beam and Equation T-13 for a Timoshenko beam.

Equation T-12 Rotation of a beam for an Euler-Bernoulli beam

Equation T-13 Rotation of a Timoshenko beam

The formula for the displacement of the sliding bearing is presented in Equation T-14. A factor 2 is taken into account for including the rotation of both rotational disks.

Equation T-14 Horizontal displacement of the sliding support

The horizontal displacement of the sliding bearing of an Euler-Bernoulli beam is 0.109 metres and the horizontal displacement of a Timoshenko beam is 0.117 metres which is determined using the Mathcad model.

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Figure T-19 MatrixFrame calculation; Loads, Moment line, Shear force line, and deflections.

Figure T-20 Bending moments for a simply supported beam

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Figure T-21 Shear forces for a simply supported beam

Figure T-22 Deflections of an simply supported beam

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T.6.3.1 CONCLUSION

The results of the MatrixFrame calculation equals the results of the differential equation analytical solved using Mathcad. So, it can be concluded that the span of the 4th mechanical scheme presented in Figure T-11 can be modelled as a simply supported beam. The moments and the shear forces are equal for the Euler- Bernoulli beam and the Timoshenko beam. The deformations are larger for the Timoshenko beam and have to be calculated with the differential equation.

T.7 MATHCAD SHEETS

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U. Gate design

U.1 GLOBAL DESIGN OF THE GATE

The thickness of the gate should be in the order of 3 metres and the radius of the disk 6 metres according to appendix T.2. The deflection of the gate must be equal or lower than 100 millimetres. The hydrostatic pressure and the wave load which are calculated in appendix Q are schematised as a line load acting at the centre of the gate. The cross section of the gate, presented in Figure U-1 is checked for the ultimate limit state, the service ability limit state, and the maximum deflections.

Figure U-1 Cross section of the gate (determined in appendix T.4)

U.1.1 SOLID PLATES

The plates are schematized as solid FRP plates for this calculation. The gate widths are varied from 2.5 metres to 3.5 metres and the wall thickness from 0 metres to 0.2 metres. The moment of inertia, the cross sectional area, and the position of the centre of gravity measured from the downstream side are determined for the different cross sections using the program AutoCAD. A cross section of the gate which is analysed by AutoCAD is presented in Figure U-2. The centre of gravity, the cross sectional area, and the moment of inertia are interpolated as a function of plate thickness using Mathcad. The parameters used for the determination of the stresses are presented in Table U-1.

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Figure U-2 Gate cross section; solid plates

Table U-1 Parameters used for the cross section calculation

Parameter Symbol value unit Based on 2 Maximum stress for SLS σd_SLS 69 N/mm Equation R-12 2 Maximum stress for ULS σd_ULS 117 N/mm Equation R-15

Maximum deflection wd 0,1 m (Sloten, 2012)

Distributed SLS load qSLS 295 kN/m Mathcad sheets

Distributed ULS load qULS 376 kN/m Mathcad sheets

E modulus for the main load bearing direction based on the CUR96 E1_CUR 25,8 GPa (Stichting CUR, 2003) (Stichting CUR, 2003) & (Snijder, G-modulus G12 5,6 GPa 2012) Span L 41 m Appendix N.4 and N.5 Moment of inertia I Variable m4 - Cross sectional area A Variable m2 - Gate height H 8 m Appendix T.3 Plate thickness t variable m - Centre of gravity from the downstream side z Variable m - Gate width W variable m -

The stresses are determined using Equation U-1 and the deflections using Equation U-2 using the interpolated values of the centre of gravity, cross sectional area, and the moment of inertia.

Equation U-1 Maximum stress as a function of the plate thickness

In which:

Md = design bending moment [kNm] z(t) = centre of gravity measured from the downstream side of the gate [m] I(t) = moment of inertia [m4]

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Equation U-2 Deflection at mid span for a Timoshenko beam as a function of the plate thickness

In which:

qd = design distributed load [kN/m] l = span of the gate [m] I(t) = moment of inertia as a function of plate thickness [m4]

E1 = tensile modulus for the main load bearing direction [N/mm2]

G12 = shear modulus of the laminate [N/mm2]

A(t) = cross sectional area as a function of the plate thickness [m2]

The cross sectional properties of the gates are presented in four sub tables in Table U-2. The cross sectional properties for a ranging wall thickness are determined for a gate width of 2.5 metres, 3 metres, 3.25 metres, and 3.5 metres. The first column of a sub table gives the wall thicknesses, the second column gives the distance from the downstream side of the gate till the centre of gravity, the third column gives the cross sectional area, and the fourth column gives the moment of inertia. The location of the centre of gravity is nearly constant. Therefore, a constant value is used for the sequential cross sectional analysis.

Table U-2 Cross sectional properties of a gate composed of massive plates

Gate width of 2.5 metres Gate width of 3.0 metres t z A I t z A I [m] [m] [m2] [m4] [m] [m] [m2] [m4] 0 1,00 0 0 0 1,25 0 0 0,05 1,00 0,88 0,86 0,05 1,26 0,93 1,32 0,1 1,00 1,76 1,64 0,1 1,26 1,85 2,54 0,15 1,00 2,62 2,35 0,15 1,26 2,75 3,65 0,2 1,01 3,46 2,97 0,2 1,26 3,64 4,67

Gate width of 3.25 metres Gate width of 3.5 metres t z A I t z A I [m] [m] [m2] [m4] [m] [m] [m2] [m4] 0 1,40 0 0 0 1,52 0 0 0,05 1,40 0,95 1,59 0,05 1,52 0,97 1,90 0,1 1,40 1,88 3,06 0,1 1,52 1,93 3,67 0,15 1,40 2,80 4,41 0,15 1,52 2,88 5,31 0,2 1,40 3,71 5,66 0,2 1,52 3,80 6,83

The maximum stresses of the four gate widths as a function of the plate thickness are presented in Figure U-3, Figure U-4, Figure U-5, and Figure U-6. The maximum deflections of the four gate widths as a function of the plate thickness are presented in Figure U-7, Figure U-8, Figure U-9, and Figure U-10. The stresses for all gate widths are below the ULS limit and the SLS limit proposed by FiberCore which are used as design value (stated in appendix R.1.3). The deflection is governing for the design according to Figure U-7, Figure U-8, Figure U-9, and Figure U-10. The following wall thicknesses per gate width are needed in order to comply with the deflection limit:  width of the gate is 2.5 metres

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o a wall thickness larger than 25 centimetres  width of the gate is 3 metres o a wall thickness of 17 centimetres  width of the gate is 3.25 metres o a wall thickness of 15 centimetres  width of the gate is 3.5 metres o a wall thickness of 12 centimetres.

The thickness of FRP plates which is normally produces is 5 centimetres and could be stretched to 7 centimetres (Snijder, 2012). The resulting wall thicknesses are larger with respect to the present and future production limit of 5 and 7 centimetres, so a solid FRP plate is not an option. An option is the application of a sandwich panel instead of a solid FRP plate. Two FRP plates of 5 or 7 centimetres thick could be used as outer faces of a sandwich panel which is elaborated in U.1.2. Another improvement of the presented design is an increase of the longitudinal tensile modulus. The company FiberCore has designed a glass fibre reinforced laminate with a higher E-modulus in the main load bearing direction. The transverse properties of the FiberCore laminate are comparable with the laminate proposed by CUR (Snijder, 2012). Therefore, the E-modulus for the longitudinal direction (E1) proposed by CUR is replaced by the E-modulus of FiberCore for the calculation of the sandwich panel.

Figure U-3 Stresses for a gate width of 2.5 metres

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Figure U-4 Stresses for a gate width of 3 metres

Figure U-5 Stresses for a gate width of 3.25 metres

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Figure U-6 Stresses for a gate width of 3.5 metres

Figure U-7 Deflections for a gate width of 2.5 metres

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Figure U-8 Deflections for a gate width of 3 metres

Figure U-9 Deflections for a gate width of 3.25 metres

Figure U-10 Deflections for a gate width of 3.5 metres

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U.1.2 SANDWICH PANELS

Sandwich panels are used to generate a larger FRP area per running metre and to increase the moment of inertia of the panel. Instead of one plate of 5 centimetres thick (which is the present production limit), two faces of 5 centimetres are combined into one panel realising a FRP thickness of 10 centimetres. The present production limit could even be stretched to a plate thickness of 7 centimetres resulting in a FRP thickness of 14 centimetres for a sandwich panel (Snijder, 2012). The FRP faces have to be positioned as far as possible from the centre of gravity to generate a large moment of inertia of the gate. Therefore, a small core thickness of 0.15 metres is used.

The tensile modulus for the longitudinal direction E1 could be increased by applying more unidirectional fibres in the main load bearing direction with respect to the proposed laminate designed by the CUR; the transverse properties of the CUR laminate remain in the same order; the properties of this laminate proposed by FiberCore are presented in Table U-3 (Snijder, 2012).

Table U-3 Laminate properties for a changed E1

Stiffness properties Unit CUR laminate Adjusted laminate

E1 GPa 25,8 31

E2 GPa 15,9 15,9 G12 GPa 5,6 5,6 v12 - 0,32 0,32

U.1.2.1 INFRACORE® PANELS

The solid plates are now replaced by InfraCore® Panels. The outer and inner faces are composed of FRP laminate and a non-constructive core which is located in between the faces. Shear walls interconnects the outer and inner face. A cross section of the InfraCore® panel which is based on a consult with FiberCore is presented in Figure U-11.

Figure U-11 Cross section of an InfraCore® panel

The maximum stresses for the service ability limit state and the ultimate limit state are calculated using Equation U-1 and deflections of the gate are calculated using Equation U-2. Cross sectional properties like the moment of inertia, cross sectional area, and the centre of gravity are determined using AutoCAD. A cross section composed of InfraCore® panels which is analysed is presented in Figure U-12. The width ’W’ presented in Figure U-12 is varied from 3 metres till 3.75 metres and the thickness of the faces ‘t’ is varied from 0.05 metres till 0.07 metres. A thickness of 0.05 metres is equal to the present maximum thickness and could be stretched to a thickness of 0.07 metres. A smaller thickness is not preferable because a large thickness is needed according to U.1.1. The results of the cross sectional analysis are presented in Table U-4. The deformations are governing according to U.1.1, so a choice for the width and FRP thickness is made on basis of the deformation limit which is performed in the main report.

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Figure U-12 InfraCore® panel; gate cross section

Table U-4 Cross sectional properties of a InfraCore® panel gate

Gate width of 3 metres Face thickness 0,05m Face thickness 0,06m Face thickness 0,07m FRP I A z FRP I A z FRP I A z face [m4] [m2] [m] face [m4] [m2] [m] face [m4] [m2] [m] Outer Outer Outer 1,35 0,95 1,61 1,14 1,86 1,33 shell shell shell Inner Inner Inner 0,94 0,89 1,10 1,06 1,25 1,23 shell shell shell Total 2,29 1,84 1,24 Total 2,71 2,20 1,24 Total 3,10 2,56 1,24

Gate width of 3,25 metres Face thickness 0,05m Face thickness 0,06m Face thickness 0,07m FRP I A z FRP I A z FRP I A z face [m4] [m2] [m] face [m4] [m2] [m] face [m4] [m2] [m] Outer Outer Outer 1,65 0,98 1,96 1,18 2,27 1,37 shell shell shell Inner Inner Inner 1,18 0,92 1,38 1,10 1,57 1,27 shell shell shell Total 2,83 1,90 1,37 Total 3,34 2,27 1,37 Total 3,84 2,64 1,37

Gate width of 3,5 metres Face thickness 0,05m Face thickness 0,06m Face thickness 0,07m FRP I A z FRP I A z FRP I A z face [m4] [m2] [m] face [m4] [m2] [m] face [m4] [m2] [m] Outer Outer Outer 1,98 1,01 2,36 1,21 2,73 1,41 shell shell shell Inner Inner Inner 1,46 0,95 1,71 1,13 1,94 1,31 shell shell shell Total 3,44 1,96 1,50 Total 4,06 2,34 1,50 Total 4,67 2,72 1,50

Gate width of 3,75 metres Face thickness 0,05m Face thickness 0,06m Face thickness 0,07m FRP I A z FRP I A z FRP I A z face [m4] [m2] [m] face [m4] [m2] [m] face [m4] [m2] [m] Outer Outer Outer 2,33 1,04 2,77 1,24 3,21 1,44 shell shell shell Inner Inner Inner 1,75 0,97 2,05 1,16 2,34 1,35 shell shell shell Total 4,07 2,01 1,62 Total 4,82 2,40 1,62 Total 5,55 2,79 1,62

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U.1.2.2 ADJUSTMENT TO THE INFRACORE® PANEL CROSS SECTION

The cross section presented in Figure U-12 has a flat top and bottom side which is disadvantageous for the flow of water. Instabilities like vortex shedding and an undefined separation points could be present for this cross section. These instabilities could result in vibrations of the gate. Therefore, the flat top and bottom side are replaced for a curved top and bottom side as presented in Figure U-13. This is advantageous for the flow of water over the gate for a fully dammed and a partly dammed operation. These curved top and bottom side are also applied for the submerged segment gate of Nakdong which is presented in Figure T-9 (kgal, 2012). Further investigations have to be implemented for the flow patterns over the gate to determine the vibrations and the impact of the vibrations caused by vortex shedding and the undefined separation point. A solution for the undefined separation point could be the installation of a ‘mesh’ at the top side of the gate in order to fix the separation point. The moment of inertia, the cross-sectional area, and the deflection of the cross sections of Figure U-12 and Figure U-13 are presented in Table U-5. The deflection of the adjusted cross section remains within the 0.1m limit. So, this cross section is used for further elaboration.

Figure U-13 Adjusted cross section of the gate

Table U-5 Cross sectional parameters and deflection of the original and adjusted gate

I A v Gate cross section [m4] [m2] [m]

Cross section presented 4,077 2,34 0.090 in Figure U-12

Cross section presented 3,92 2,26 0.093 in Figure U-13

U.1.3 DIMENSIONS OF THE WEBS FOR SHEAR

The presented cross section has to bear the moment. The shear force has to be transferred to the bearings by shear webs located in between the upper and lower InfraCore panel. The shear webs are also needed to limit the vertical span of the panels as presented in Figure U-14. A choice is made for InfraCore® panels or sandwich panels instead of a solid FRP plate. The out of plane stiffness of InfraCore® panels or sandwich panels are larger with respect to a solid FRP plate resulting in a more robust design. The faces of the shear panel bear the shear force. The maximum shear force is present at the bearings and is equal to . The panels are calculated for SLS and ULS. A strain of 1.2% is used for ULS and a strain of 0.27% is used for SLS. The maximum angular rotation of the laminate for ULS and SLS is

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presented in Equation U-3 for ULS and Equation U-4 for SLS. The maximum shear forces are calculated by

multiplying the maximum angular rotations by the shear modulus G12. The material factors and conversion factors presented in R.1.3 are used for the determination of the design shear stress. The representative values of this laminate are determined in B.5.2 and are presented in Table U-6. The shear forces for SLS and ULS are presented in Equation U-5 and Equation U-6.

Equation U-3 Maximum angular rotation for ULS (based on (Nijhof, 2006))

Equation U-4 Maximum angular rotation for SLS (based on (Nijhof, 2006))

[ ] [ ]

Equation U-5 Shear force for SLS

[ ] [ ]

Equation U-6 Shear force for the ULS

Table U-6 Design parameters for the shear web design

parameter Symbol Value Unit

G-modulus G12 7.0 GPa Angular rotation SLS γmax_SLS 0.54 % Angular rotation ULS γmax_ULS 2.4 % 2 Representative shear stress SLS τrep_SLS 37.8 N/mm 2 Representative shear stress ULS τrep_ULS 168 N/mm Material factor γm 1.62 - Maximum conversion factor SLS γc_SLS 1.791 -

Maximum conversion factor ULS γc_ULS 1.628 - 2 Design shear stress SLS with factors τd_SLS 13.0 N/mm 2 Design shear stress SLS without factors τd_SLS_without 37.8 N/mm 2 Design shear stress ULS τd_ULS 63.7 N/mm

The maximal shear stress for SLS is calculated with material and conversion factor and without material and conversion factor. The CUR96 proposes the application of the material and conversion factor. However the company FiberCore does not use these factors for the actual design of hydraulic structures. They state that the application of the material factor and the conversion factor is too conservative nowadays. Tests have shown a higher strength than proposed by CUR96. Therefore the shear stress determined without application of material and conversion factor is used. The maximum shear stress for a cross section is calculated using Equation U-7. This equation is rearranged to obtain an expression for the maximum shear force which the cross section can bear. This rearranged equation is presented in Equation U-8.

Equation U-7 Calculation of shear stress

In which: Τ = shear stress V = shear force S = area moment

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B = width of the cross section I = moment of inertia

Equation U-8 Expression for the maximum shear force per shear web

The moment of inertia of a single web is calculated using Equation U-9 and the area moment is calculated using Equation U-10.

Equation U-9 Moment of inertia of a shear web

( )

Equation U-10 Area moment of a shear web

In which: b = total width of the faces per shear web [m] h = height of a shear panel [m]

The shear webs are uncoupled from the outer plates for the shear force calculation. So, every shear web acts separetly. The sum of the shear force capacities of the distinct webs must be equal or larger than the applied shear force. Two designs are elaborated, namely a design composed of 4 shear webs and a design composed of 5 shear webs presented in Figure U-14. The face thickness is fine tuned in order to realise an economic design.

Figure U-14 Gate cross section for 4 shear webs and 5 shear webs

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The thickness is varied from 0.01 to 0.04 metres for the design of the shear webs. This thickness represents the total thickness of both faces of the shear panel. The shear capacity is calculated for the maximum shear stress for SLS and ULS. The results and the unity check are presented in Table U-7 and Table U-8. The presented values are calculated using MathCad.

Table U-7 Unity checks for a design for four shear webs

Face thickness of a shear wall [m] Face thickness of a shear wall [m] 4 shear t=0.01 t=0.02 t=0.025 t=0.03 t=0.01 t=0.02 t=0.025 t=0.03 webs Shear force SLS Shear force ULS design [kN] [kN] Web 1 622 1245 1556 1867 1049 2097 2622 3146 Web 2 728 1457 1821 2185 1227 2454 3068 3681 Web 3 738 1477 1846 2215 1244 2488 3110 3732 Web 4 658 1315 1644 1973 1108 2216 2770 3324 Total shear 2746 5494 6867 8240 4628 9255 11570 13883 force Design 5995 7713 shear force Unity check 2,2 1,1 0,9 0,7 1,7 0,8 0,7 0,6

Table U-8 Unity checks for a design for five shear webs

Face thickness of a shear wall [m] Face thickness of a shear wall [m] 5 shear t=0.01 t=0.02 t=0.025 t=0.03 t=0.01 t=0.02 t=0.025 t=0.03 webs Shear force SLS Shear force ULS design [kN] [kN] Web 1 589 1179 1474 1769 994 1987 2484 2981 Web 2 703 1406 1758 2109 1185 2369 2961 2554 Web 3 746 1492 1865 2238 1257 2513 3142 3770 Web 4 723 1446 1808 2170 1219 2437 3046 3656 Web 5 635 1270 1588 1905 1070 2140 2675 3210 Total shear 3396 6793 8493 10191 5725 11446 14308 16171 force Design 5995 7713 shear force Unity check 1,8 0,9 0,7 0,6 1,3 0,7 0,5 0,5

A minimum thickness of 2.5 centimetres is needed for a cross section composed of 4 shear webs and a minimum thickness of 2 centimetres is needed for a cross section composed of 5 shear webs according to Table U-7 and Table U-8. The thickness is divided over the two faces of the panel for calculating the face thicknesses. So a minimum thickness of 1.25 centimetres for four shear panels and 1 centimetre for 5 panels should be sufficient. However bending moments and forces generated by the connection are also present. Therefore a thickness of 2 centimetres is chosen for the design composed of 4 shear webs.

U.1.4 CHECK FOR TORSION

The gate has also to bear a torsional moment which is generated by the hydrostatic pressure acting at the gate. The torsional moment generates a shear force in the faces of the cross section. The shear stress is calculated using Equation U-11.

Equation U-11 Torsional shear stress (Hartsuijker, 2007)

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In which:

τtorsion = torsional shear stress [N/mm2]

Mt = torsional moment [Nm]

Am = area enclosed by the central axis of the faces which is indicated in Figure U-15 [m2] t = plate thickness [m]

Figure U-15 Torsion definitions (Hartsuijker, 2007)

The maximum shear stress is checked for SLS and ULS. The strain limit of 0.27% for SLS and the strain limit of 1.2% for ULS are used for the calculations. The faces of the sandwich of the outer plates bear the shear forces generated by the torsional moment. The laminate of the faces is a anisotropic laminate with a higher glass fibre content in the main direction and a lower glass fibre content in the transverse direction. The shear modulus of the laminate is lower with respect to the shear modulus of the shear elements presented in U.1.3. The material parameters used for the calculation are presented in Table U-9. The thicknesses of the faces of the sandwich are 6 centimetres which is determined in U.1.2. The sandwich is composed of two faces and has a total FRP thickness of 12 centimetres.

The area which is enclosed by the centre line of the sandwich Am is marked in Figure U-16 and is calculated using the massprop function of AutoCAD. The torsional moments for SLS and ULS are equal to:  332kNm for SLS  450kNm for ULS.

The shear stresses generated by the torsional moment are calculated using Equation U-12 and Equation U-13 for SLS and ULS; the results are well below the design shear stresses. So, the gate is able to bear the torsional moment. The stress is very low so it was justified to neglect the torsional moment in the pre design of appendix T and U.1. [ ]

[ ] [ ]

Equation U-12 Torsional shear stress for SLS [ ]

[ ] [ ] [ ]

Equation U-13 Torsional shear stress for ULS

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Table U-9 Design parameters for the torsional stress check

parameter Symbol Value Unit

G-modulus G12 5.6 GPa Angular rotation SLS γmax_SLS 0.54 % Angular rotation ULS γmax_ULS 2.4 % 2 Representative shear stress SLS τrep_SLS 30.2 N/mm 2 Representative shear stress ULS τrep_ULS 134.4 N/mm Material factor γm 1.62 - Maximum conversion factor SLS γc_SLS 1.791 - Maximum conversion factor ULS γc_ULS 1.628 - 2 Design shear stress SLS with factors τd_SLS 10.4 N/mm 2 Design shear stress SLS without factors τd_SLS_without 30.2 N/mm 2 Design shear stress ULS τd_ULS 50.9 N/mm

Figure U-16 Cross sectional area Am for torsion calculation

U.1.5 DESIGN CHECK OF THE PANELS

The shear webs support the outer panels on which the hydrostatic force acts. The hydrostatic force causes local moments, local shear forces, and deflections in the outer plates in the direction with the smallest span. The maximum hydrostatic pressure acts at the lowest part of the gate for an open configuration which is presented in Figure U-17. The panel is able to bear the forces for a dammed configuration when it is strong and stiff enough to bear the presented load case, because the loads for a dammed configuration are lower with respect to this load case.

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Figure U-17 Pressures for the shear force calculation of the panels

The panels are schematised as non-curved panels spanning in between the shear walls. The field moment of the panel is calculated using Equation U-14 and the support moment of the panel is calculated using Equation U-15.

Equation U-14 Field moment for the panel

Equation U-15 Support moment of the panel

The outer plates of the panel bear the tension and compressive forces in the main and transverse direction.

The tensile modulus E2 is used for the design check for the transverse loads. Also a local shear force is present at the panel at the shear web connections. The vertical shear elements which are located in between the two faces bears the shear force. The shear force at the connections at the shear webs is presented in Equation U-16. The laminate applied for the shear webs is the same laminate as used for the shear elements. This laminate has a higher G-modulus with respect to the anisotropic laminate used for the outer faces.

Equation U-16 Shear force at the shear wall

Local deflections of the plates are calculated using the extended ‘Forget me Not’ which are derived in appendix T.6. The ‘Forget me Not’ for a two sided supported beam which is fixed at both sides is presented in Equation U-17.

Equation U-17 Deflection at mid span for a fixed-fixed plate

The design properties used for the check for moment capacity, check for shear force capacity, check for deflections, and a check for the combined stress capacity are presented in Table U-10.

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Table U-10 Parameters used for local strength calculation parameter Symbol Value Unit E-modulus of the plate laminate for the transverse direction E2 25,8 GPa

G-modulus of the shear elements G12-w 5,6 GPa Distributed load SLS qlocal_SLS 127 kN/m Distributed load ULS qlocal-ULS 152 kN/m Span between shear webs lwebs 1.5 m Spacing between shear elements dshpl To be determined m Thickness of a shear element tshpl To be determined m Core thickness bcore 0.15 m Face thickness of outer plates tplate 0.06 m 2 Design stress for transverse direction (SLS) σd2_SLS 43 N/mm 2 Design stress for transverse direction (ULS) σd2_ULS 72 N/mm 2 Design shear stress of the webs (SLS) τd_SLS_w 38 N/mm 2 Design shear stress of the webs (ULS) τd_ULS_w 64 N/mm

U.1.5.1 CHECK FOR MOMENT CAPACITY

The transverse moments at the supports for SLS and ULS are:

The arm in between the faces is equal to the core dimension and two times the half of the thickness of the FRP faces. This results in a thickness of 0.21m as indicated in Figure U-18. The transverse tensile and compressive forces in the FRP faces in the transverse direction are:

The transverse tensile and compressive stresses per running metre are:

The maximum stress for the transverse direction for SLS is 43 N/mm2 and the maximum stress for the transverse direction for ULS is 72N/mm2, so the actual stresses are well below the maximum stresses.

Figure U-18 Plate cross section

U.1.5.2 CHECK FOR SHEAR CAPACITY

The design shear stress at the shear webs for SLS and ULS are:

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A spacing of 15 centimetres is chosen in between the shear elements, so the shear force per shear element for SLS and ULS are:

( ) ( )

The moment of inertia and the area moment of a shear element located in between the faces of the sandwich are:

Now the shear stress is calculated for the shear elements. The shear stresses for SLS and ULS are:

The maximum shear stress for SLS is 38N/mm2 and the maximum shear stress for ULS is 64N/m2, so the shear stresses of the laminate are lower with respect to the maximum shear stress.

U.1.5.3 CHECK FOR MAXIMUM DEFLECTIONS

At last the deflection of the span in between the shear webs is calculated. The moment of inertia and the cross section per running metre of the panel presented in Figure U-19 are:

( ( ) )

( ) The deflection of the panel at mid span is:

The deflection at mid span is low and would not hamper the functioning of the weir and the water tight connection for fully dammed operation.

Figure U-19 Parameters for deflection calculation

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U.1.5.4 CHECK FOR COMBINED STRESS CAPACITY

Design checks are performed for the main direction and transverse direction but separate design checks only are not sufficient for the design. The combined stress capacity check for the tensile and compressive forces needs to be executed to verify whether the maximum combined stress is exceeded. The combined tensile stress is determined using Equation U-18 and the combined compressive stress is determined using Equation U-19. This design check is executed for SLS and ULS. The representative values for the combined stress check are presented in Table U-11. The SLS stress is determined without conversion and material factor as proposed by FiberCore.

( ) ( ) ( ) ( ) ( )

Equation U-18 Combined stress for tensile forces (Stichting CUR, 2003)

( ) ( ) ( ) ( ) ( )

Equation U-19 Combined stress for compressive forces (Stichting CUR, 2003)

Table U-11 Representative values for the combined stress capacity check

parameter Symbol Value Unit Source Representative tensile and compressive stress for the σ 84 / 372 N/mm2 Appendix B.4.6 main direction (SLS / ULS) 1,t,R Representative tensile and compressive stress for the σ 43 / 191 N/mm2 Appendix B.4.6 transverse direction (SLS / ULS) 2,t,R 2 Design shear stress (SLS / ULS) τ12,R 30 / 134 N/mm Appendix B.4.6 Material factor (SLS / ULS) γm 1 / 1.62 - Appendix B.4.6.2 Conversion factor (SLS / ULS) γc 1 / 1.79 - Appendix B.4.6.2 Load factor γf 1.25 - Appendix P

The service ability load which is presented in Figure U-20 is used for the check for the combined stress capacity. The hydrostatic and wave load are schematised as a distributed line load q which is present at the middle of the cross section of the gate with a torsional moment caused by the shift of the resultant force. This distributed load is used for the calculation of the maximum tensile and compressive stress generated by the bending moment for the main load bearing direction and the shear force caused by the torsional moment. The hydrostatic pressure which is present at the bottom of the gate is used for the transverse stress calculation. The maximum pressure of 80kN/m2 is uniformly applied for a span of 1.5 metres. The maximum bending moment at the shear web connection for the main direction and transverse direction and the maximum shear force at the shear web support are calculated using Mathcad.

The load factor γf is included for SLS and ULS stresses and not at the right hand side of Equation U-18 and Equation U-19 for the combined stress capacity checks. So Equation U-18 and Equation U-19 changes into Equation U-20 Equation U-21.

( ) ( ) ( ) ( ) ( )

Equation U-20 Combined stress capacity check for service ability limit states (SLS)

( ) ( ) ( ) ( ) ( )

Equation U-21 Combined stress capacity check for ultimate limit states (ULS)

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Figure U-20 Hydrostatic pressures for a fully dammed operation

The calculations of the combined stress check are performed using Mathcad. A unity check is performed by dividing the left hand side of Equation U-20 and Equation U-21 by the right hand side. The results of the unity check are:  tensile stress check for SLS o 0.08 which is lower than 1  tensile stress check for ULS o 0.05 which is lower than 1  compressive stress check for SLS o 0.154 which is lower than 1  compressive stress check for ULS. o 0.09 which is lower than 1.

U.2 DESIGN CHECKS

Based on: (Timoshenko & Gere, 1963) and (Zenkert, 1995).

More detailed design checks are performed to verify the capacity and performance of the designed gate. Eight failure modes for sandwich panels are identified by Professor Zenkert of the University of Stockholm. Not every failure mode presented in Figure U-21 is applicable for the design of the InfraCore® panel because the foam core is replaced by FRP shear elements. The eight failure modes of sandwiches are listed in this section. It is mentioned when a failure mode is not applicable for the InfraCore® panel per sub section. For now the walls of the gate are schematised as non-curved panels. Design checks for non- curved sandwich plates are performed in U.3.

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Figure U-21 Failure modes of sandwich beams. (a) Face yielding/fracture, (b) core shear failure, (c) and d) face wrinkling, (e) general buckling, (f) shear crimping, (g) face dimpling and (h) local indentation (Zenkert, 1995).

U.2.1 YIELDING OR FRACTURE OF THE FACE IN TENSION OR COMPRESSION (FAILURE MODE A)

The yielding or fracture failure of a panel depends on the proposed strain. The failure strain for SLS originates from the strain for micro cracking and the failure strain for ULS originates from fracture. The micro-cracking strain and fracture limit is used for the design of the panel for the combined stress capacity check in U.1.5.4. The compressive and tensile stresses are well below the limit for SLS and ULS. Therefore, no micro-cracking and fracture failure occurs for a fully dammed operation.

U.2.2 CORE SHEAR FAILURE (FAILURE MODE B)

The core is not able to resist the shear force for core shear failure, which is presented in figure b of Figure U-21. The check proposed by Zenkert cannot be performed because a non-constructive core combined with FRP shear elements are used instead of a constructive core. The shear elements located in between the faces are designed for the maximum shear stress and are therefore able to resist the SLS and ULS shear force. Therefore, no check is performed for core shear failure.

U.2.3 FACE WRINKLING (FAILURE MODE C AND D)

For face wrinkling, the core is crushed or the face separates from the core due to a compressive force and a moment. The check proposed by Zenkert cannot be performed because a non-constructive core and shear elements are used. Instead of face wrinkling, face dimpling has to be checked because FRP shear elements are used.

U.2.4 GENERAL BUCKLING (FAILURE MODE E)

The gate is subjected to bending and not to compression. However the panels of the compressed part of the structure could buckle due compressive stress generated by the moment. Therefore local buckling checks are performed for the compressed panels instead of a general buckling check. Schematising the panels as two sided supported beams and using the standard buckling formula gives a wrong indication

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of the buckling stress because the panels are clamped at four sides. Therefore the plate buckling stress is calculated which is executed in U.2.6 and U.3.2.

U.2.5 SHEAR CRIMPING (FAILURE MODE F)

Shear crimping occurs when an out of plane load and an in plane compressive force are present. It looks like a local mode of failure, but it is a form of general buckling. The critical buckling load is used for the determination of the critical face stress. The critical face stress for shear crimping is calculated using Equation U-22. This failure mode is checked after the determination of the bucking load performed in appendix U.3.2.1.

Equation U-22 Shear crimping

In which:

σf;sc = failure stress for shear crimping [N/mm2]

Ff = general buckling load [N]

tfaces = thickness of the faces [mm]

U.2.6 FACE DIMPLING (FAILURE MODE G)

The faces are buckling locally for the failure mode ‘face dimpling’ which is presented in failure mode ‘g’ of Figure U-21. The critical face dimpling stress is calculated in U.2.6.1 for sandwiches composed of corrugated cores and in U.2.6.2 for a honeycomb core. The top view of a honeycomb core and the cross section of a corrugated core are presented in Figure U-22. The critical face dimpling stress for a corrugated core is comparable for a InfraCore® panels with shear elements aligned in one direction and the critical face dimpling stress for a honeycomb core is comparable for an InfraCore® element with shear webs in two directions. It is not clear whether FiberCore uses shear webs in one direction or two. Therefore both face dimpling stresses are calculated.

Figure U-22 Top view of a honeycomb sandwich (left) and cross section of a corrugated sandwich (right) (Zenkert, 1995)

U.2.6.1 FACE DIMPLING OF A CORRUAGED CORE

The critical face dimpling stress is calculated using Equation U-23.

Equation U-23 Face dimpling (Zenkert, 1995)

In which:

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σf;fd = failure stress for face dimpling [N/mm2] k = buckling coefficient which depends on the size of the plate inscribed by the corrugation [-]

Ef = modulus of elasticity which is approximated for orthotropic faces by: 2 √ [N/mm ]

vf = poisson ratio [-]

tf = thickness of the faces [m] l = length of the plate [m]

A value for the buckling coefficient is not given by Zenkert. Zenkert refers to a study performed by Timoshenko about the elastic stability of beams and element including the buckling theory for an ordinary homogeneous plate (Timoshenko & Gere, 1963). The buckling load of a plate inscribed by the corrugation depends on the ratio of span of the plate ‘a’ and the length of the compressed side ‘b’ which are presented in Figure U-23 and the number of half waves of the post buckled shape of the plate. The number of half waves of the buckling mode (‘m’) as a function of the ratio ‘a’/’‘b’ is given in Figure U-24.

Figure U-23 Plate dimensions (Timoshenko & Gere, 1963)

Figure U-24 Buckling coefficients as a function of the number of half waves (m) and the ratio a/b (Timoshenko & Gere, 1963)

The shear elements for the outer sandwich plates are spaced at every 15 centimetres and the span between the shear webs is 1.5 metres. So ‘a’ is equal to 0.15 centimetres and ‘b’ is equal to 1.5 metres resulting in an a/b ratio of:

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[ ] [ ] [ ] A buckling mode consisting of 1 half wave is present for this plate dimension according to Figure U-24 which is in accordance with the sketched face dimpling deformation of in Figure U-21. The corresponding buckling coefficient ’k’ is very large and not given in Figure U-24. Timoshenko & Gere proposes also a general buckling formula for a buckling mode consisting of one half wave (m=1) which is presented in Equation U-24.

( )

Equation U-24 Critical force for one half wave (m=1) (Timoshenko & Gere, 1963)

In which:

Nf;fd = buckling force [N] D = modulus of rigidity of the face [Nm]

= ( )

In which: E = tensile modulus [N/mm2] b = width of the element [m]

tf = face thickness [m] b = width of the compressed side (Figure U-24) [m] a = span of the element (Figure U-24) [m]

The critical buckling stress is calculated by dividing the buckling force by the cross sectional area of the considered element. The buckling force of the corrugated laminate is:

( )

Now the face dimpling stress for a sandwich/InfraCore® panel with a corrugated core is calculated. The critical face dimpling stress is:

U.2.6.2 FACE DIMPLING OF A HONEYCOMB CORE

The critical face dimpling stress of a honeycomb core is calculated using Equation U-25.

( )

Equation U-25 Critical face dimpling stress for a honeycomb core (Zenkert, 1995)

In which: s = radius of a honeycomb which is equal to 15 centimetres.

The face wrinkling stress of a sandwich with a honeycomb core is: [ ] [ ] ( ) [ ]

U.2.6.3 CONCLUSION

The design compressive stresses for SLS and ULS are 43 N/mm2 and 72 N/mm2. These stresses are well below the calculated face dimpling stresses of an honeycomb core and a corrugated core. Therefore, no face dimpling is present.

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U.2.7 CORE INDENTATION (FAILURE MODE H)

Core indentation occurs for concentrated loads at fittings, corners, or joints. Core indentation is solved by spreading the applied load over a sufficient large area (Zenkert, 1995). The minimal area is estimated using Equation U-26.

Equation U-26 Core indentation

In which:

Amin = the minimal area P = the localised load

σcz = compressive strength of the core material.

The localised load is applied at the face, and the face acts as a plate on an elastic support representing the core. The face bends due to the load and a compressive stress is applied to the core material. The core fails when the applied stress is larger than the maximum compressive stress of the core. The structures or elements are bolted together for large FRP structures. This is needed because the area which has to be glued together becomes too large. The glue is already hardened at one side of the gate when the glue at the other end of the element is just applied. The used bolts are fixed to the plate by a chemical anchor in a massive block of FRP as presented in Figure U-25. The thickness of the surrounding FRP block must be equal or larger as 3 times the bolt diameter to prevent the sandwich from failing. The steel bolt is the weakest part of the connection for this configuration (Snijder, 2012). Therefore, core indentation at fittings, corners, or joints is not present for the described bolted connection.

Figure U-25 Bolt connection proposed by FiberCore Figure U-25 (based on: (Snijder, 2012))

U.3 PLATE CALCULATIONS

The previous calculations are based on a beam schematisation; the spans are modelled as a two sided supported beam. Now the maximum deflections and the buckling load of the sandwich/InfraCore® plates are checked. In order to do so, a grid of the elements located in between the upper and lower plate must be defined. The shear webs are spaced at 1.5 metres from each other in the transverse direction. In order to increase the stiffness of the gate, plates in the longitudinal direction are placed. It is assumed that the plates are spaced at every 5.23 metres as presented in Figure U-26. The maximum plate deflections are calculated in U.3.1 and the critical buckling force of a plate is determined in U.3.2.

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Figure U-26 Cross section in the span direction of the gate

U.3.1 MAXIMUM PLATE DEFLECTIONS

The maximum plate deflection of four sided clamped orthotropic plates is calculated with Equation U-27. The ‘x’ represents the main direction and the ‘y’ represents the transverse direction as presented in Figure U-27.

( )

[ ( ) [ ( ) ] ( ) ] [ ( ) ]

Equation U-27 maximum deflection in the middle of the plate (Zenkert, 1995)

In which: q = pressure [N/mm2]

νxy = poisson ratio xy [-]

νyx = poisson ratio yx [-]

Dx = flexural rigidity for the x direction [Nmm]

Dy = flexural rigidity for the y direction [Nmm]

Dxy = torsional stiffness [Nmm]

Sx = shear stiffness for the x direction [N/mm]

Sy = shear stiffness for the y direction [N/mm] a = span in the main (x) direction [m] b = span in the transverse (y) direction [m]

Figure U-27 Rectangular sandwich plate definitions (Zenkert, 1995)

The maximum pressure of 127kN/m2 which is presented in Figure U-17 is used for the calculations of the plate deflection. The xy poisson ratio is given by the CUR96 and is equal to 0.32 (Stichting CUR, 2003).

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However the yx Poisson ratio is not given. Zenkert relates the Poisson ratios with the flexural rigidities for the x and y direction which is presented in Equation U-28. This relation is used for the determination of the yx Poisson ratio. The flexural rigidity in the x direction and y direction are determined using Equation U-29 and Equation U-30.

Equation U-28 Relation between the Poisson ratio and the flexural rigidities in the x and y direction (Zenkert, 1995)

( [ ])

Equation U-29 flexural rigidity for the x direction (Zenkert, 1995)

( [ ])

Equation U-30 Flexural rigidity for the y direction (Zenkert, 1995)

In which:

E1 = tensile modulus in the main direction [N/mm2]

E2 = tensile modulus in the transverse direction [N/mm2]

bcore = thickness of the core [m]

tf = thickness of the faces [m]

The torsional stiffness is determined using Equation U-31:

( )

Equation U-31 Torsional stiffness (Zenkert, 1995)

In which:

G12 = shear modulus of the laminate [N/mm2]

tf = thickness of the faces [m]

The shear stiffness is normally given for a constructive solid core which is not present for the InfraCore panel. Therefore, the shear stiffness of a shear element is smeared out over the thickness of a shear element and the span in between the shear elements. The ‘smeared out’ shear stiffness for the x and y directions are presented in Equation U-32 and the same shear modulus is used for both direction. The weighted shear modulus is determined using Equation U-33. The number of shear elements per metre is multiplied by the thickness of the shear elements and divided by the width in which the shear elements are located. In this way, an average shear modulus is determined. This is not exact but an approximation of the real shear stiffness.

( )

Equation U-32 Shear stiffness (Zenkert, 1995)

[ ]

Equation U-33 Weighted shear modulus

In which:

G12 = shear modulus [N/mm2]

bspacing = span in between shear elements [m]

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tsw = thickness of a shear element [m]

The calculation is performed using Mathcad with the parameters presented in Table U-12.

Table U-12 Input parameters for the plate deflection calculation parameter Symbol Value Unit Pressure q 127 N/m2

Tensile modulus for the main direction E1 31 GPa Tensile modulus for the transverse direction E2 15.9 GPa Shear modulus of the shear elements G12 7.0 GPa Thickness of the faces tf 6 mm Thickness of the core bcore 15 cm Thickness of the shear elements tsw 5 mm Spacing of the shear elements bspacing 15 cm xy Poisson ratio νxy 0.32 - Span in the main direction a 5 m Span in the transverse direction b 1.5 m

The maximum plate deflection at mid span is 0.61mm which is larger than the deflection of 0.12mm which was determined in U.1.5.3. The difference is explained by the manner the plate is modelled. In U.1.5.3 the faces are taken into account for the shear deflection. But in this section, the shear elements have to bear the shear force and deflects more due to the shear force. Both deformations remain within 1mm and are acceptable for the design of the gate.

U.3.2 PLATE BUCKLING

The critical buckling load is calculated for a four sided fixed plate which is presented in Figure U-28. A fixed configuration is chosen because the continuous plate is connected to the other faces.

Figure U-28 Plate buckling; four sided fixed

The critical buckling force of the plate is calculated using Equation U-34. The critical buckling force is defined per running metre of plate.

Equation U-34 Buckling force per running metre of plate (Zenkert, 1995)

In which: P = Total buckling force [N/m]

Pb = Plate buckling force for bending contribution [N/m]

Ps = Plate buckling force for the shear contribution [N/m]

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The critical buckling force for the bending contribution is calculated with Equation U-35 and the critical buckling force for the shear contribution is calculated with Equation U-36.

( )

[ ( ) ( ) [ ( )]]

Equation U-35 Bending buckling (Zenkert, 1995)

( )

Equation U-36 Shear buckling (Zenkert, 1995)

In which:

νxy = Poisson ratio xy [-]

νyx = Poisson ratio yx [-]

Dx = Flexural rigidity for the x direction [Nmm]

Dy = Flexural rigidity for the y direction [Nmm]

Dxy = Torsional stiffness [Nmm]

Sx = Shear stiffness for the x direction [N/mm]

Sy = Shear stiffness for the y direction [N/mm] a = Span in the main (x) direction [m] b = Span in the transverse (y) direction [m]

Equation U-34, Equation U-35, and Equation U-36 are rewritten as a buckling coefficient K and a buckling

force Px as presented in Equation U-37. The effects of the plate dimensions and the boundary conditions are taken into account in the buckling coefficient which is calculated using Equation U-38.. The expression in front of the buckling force is a general expression which is not influenced by the dimensions and the boundary conditions.

( )

Equation U-37 Relation between the buckling coefficient and the buckling force (Zenkert, 1995)

( ) [ ( ) ] [ ( ) ( ) ( )]

( )

[ ( ) ] ( ) [ ( ) ( ) ( )]

Equation U-38 Buckling coefficient

In which:

x = shear factor [-] which is calculated in Equation U-39

Equation U-39 Shear factor (Zenkert, 1995)

The critical buckling load is calculated with Equation U-40.

( ) [ ]

Equation U-40 Critical plate buckling load

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The critical buckling load is calculated using Mathcad. The buckling coefficient which is calculated is equal to 3.89[-] which results in a critical buckling load of kN/m. The critical buckling stress is calculated by dividing the critical buckling load by the face thickness. The critical buckling stress is equal to 6.15 kN/mm2 for face thicknesses of 6 centimetres. The critical buckling stress is significant higher than the maximum compressive stress of the laminate. Therefore, the plate does not buckle.

U.3.2.1 CHECK FOR SHEAR CRIMPLING

Shear crimpling was already mentioned in U.2.5 but is not calculated because no critical buckling load was defined. The critical stress using the result for buckling is given in Equation U-41. [ ]

[ ] [ ]

Equation U-41Critical shear crimpling stress

This stress is equal to the buckling stress presented in U.3.2 which agrees with the statement that shear crimpling has the same limit as for buckling.

U.4 CONNECTIONS

The distinct panels of the FRP gates have to be connected to each other. The connections could be realised by glue or bolts. The lengths of the strips which have to be glued together are large (, so the first part of the glue is already hardened before the glue is applied at the other side of the gate. Therefore glued connections are not an option for the FRP gate of weir Culemborg (Snijder, 2012). Another option is a bolted connection. The bolt configuration proposed by FiberCore is already described in U.2.7 and applied for the gate design of weir Culemborg. The connections of the webs with the upstream and downstream panels are described in U.4.1 and the connections of the FRP gate with the rotation disks are described in U.4.2. The connections are described using the top view of the weir gate of Culemborg presented in Figure U-29. In consultant with the graduation committee, it is decided that the connections are only described and not calculated due to the time limit. Therefore, the described connections are only an indication how the bolts would be arranged.

Figure U-29 Topside of the submerged segment gate

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U.4.1 CONNECTION OF THE WEBS TO THE FRONT AND BACK WALL

The connections of the webs with the front and the back panels are presented in Figure U-30. The bolts are connected to the sandwich plates by chemical anchors. A RVS corner strip connects the distinct sandwich plates to each other. A close up of the connection of a shear web with the outer faces is presented in Figure U-31. The given configuration of the bolts and the dimensions of the connections are just an impression. Further investigation has to be executed in order to calculate the exact number of bolts.

Figure U-30 Cross section BB'

Figure U-31 Shear web connection; Detail B

U.4.2 CONNECTIONS OF THE FRP GATE TO THE ROTATION DISKS

The connection of the gate with the rotation disk has to be rigid. Therefore, bolts are applied at the downstream side of the gate and at the end sides of the gate as presented in Figure U-32. 10 Bolts are drawn at the end side of the gate and 16 bolts are drawn for the connection at the flat downstream side of the gate. The given configuration of bolts and the dimensions of the connections are just an impression. Further investigation has to be executed to calculate the exact number of bolts.

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Figure U-32 Cross section DD' and EE'; connection of the gate with the rotation disk

U.5 NATURAL FREQUENCIES

At last the natural frequencies of the designed gates of weir Culemborg are calculated. Two calculations are performed namely a calculation for a one degree of freedom system (1 DOF system) without damping and a calculation for a bending beam without damping. The gate is placed in water and a part of the water body which is in contact with the gate vibrates due to the vibrations of the gate. The mass of the vibrating water body is called the ‘added mass.’ The determination of the quantity of the added water mass is described in U.5.1. The vibration of the 1 DOF system is described in U.5.2 and the vibrations of a beam are described in U.5.3. It has to be verified whether the natural frequencies of the gate corresponds to the frequencies of the loads. (Large) unwanted vibrations are present when the frequencies of the loads equal the natural frequencies of the gate. Only the frequency of the waves is known, but the frequencies of the vortex shedding and the vibrations caused by overflow are not known. Therefore, the natural frequencies of the gate of weir Culemborg are compared with the dynamic behaviour of gates which are presently used for existing hydraulic structures. The dynamic calculations are performed for the FPR gate which is designed in U.1 and the steel gate which is designed in the calculation of gate weight for a span of 41 metres and a width of 3.25 metres which is described in R.1.2.

U.5.1 ADDED WATER MASS

Based on: (Kolkman & Jongeling, 1996)

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The added water mass is hardly influenced by the strength of the flow and is therefore calculated for an object in still water with the assumption of rotation-free flow and ignoring the viscosity. The added mass for high vibration frequencies of a horizontally vibrating gate on top of a threshold with

water at one side is presented in Figure U-33. The added mass is a function of a coefficient Cm, the density

of water ρ, and the height at the dry side of the gate h1.This schematisation does not completely match for weir Culemborg because a downstream water level is present for zero discharge. However this figure is used as a first approximation for the added mass for the horizontal vibrations.

The height of the upstream and downstream side of the gate are equal, so . The coefficient for added

mass is equal to 0.542 [-] for this ratio. The added mass per running metre is presented in Equation U-42 and the added mass for the total width of the gate is presented in Equation U-43. The added mass presented in Equation U-42 is used for the calculation of the natural frequencies of a bending beam and the added mass presented in Equation U-43 is used for the calculation of the natural frequencies of a one degree of freedom system (1 DOF system).

Figure U-33 Added mass for spillway gates (Kolkman & Jongeling, 1996)

[ ] [ ] [ ]

Equation U-42 Added mass per running metre

[ ] [ ] [ ] [ ]

Equation U-43 Added mass for the full span

U.5.2 VIBRATIONS OF A ONE DEGREE OF FREEDOM SYSTEM (1 DOF SYSTEM)

Based on: (Kolkman & Jongeling, 1996), (Irvine, 2006), (Spijkers, et al., sd).

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The ‘normal’ equation of motion of a normal system is presented in Equation U-44. Next to the added mass, added damping and added spring stiffness are present, which are presented in Equation U-45.

Equation U-44 Normal equation of motion for a mass-spring-damper system

Equation U-45 Equation of motion for a mass-spring-damper system in water

The normal mass and the added mass are combined in a total mass called the effective mass of a submerged body as presented in Equation U-46.

Equation U-46 Effective mass of a submerged body

In which: m = physical mass [kg]

mw = added mass [kg]

The natural frequency of the system is calculated with Equation U-47.

√

Equation U-47 Natural frequency of a 1 DOF system

In which:

fn = natural frequency [Hz] k = spring stiffness [N/m]

me = effective mass [kg]

The spring stiffness of the gate is calculated for a steel gate using Equation U-48 and for a FRP gate using Equation U-49.

[ ]

Equation U-48 Spring stiffness for a steel gate

[ ]

Equation U-49 Spring stiffness for a FRP gate

In which: q = distributed load [N/m]

lspan = span of the gate [m] E = tensile modulus [N/m2] I = moment of inertia [m4] G = shear modulus [N/m2] A = cross sectional area [m2]

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The parameters used for the calculation are presented in Table U-13. These parameters are introduced in the Mathcad sheet in order to calculate the natural frequencies. The results are presented in the following enumeration:  The natural frequency and the period of the FRP gate designed in U.1 are:

o  The natural frequency and the period of the steel gate for the cross section based on R.1.2 are:

Table U-13 Parameters for the calculation of the natural frequency for a 1DOF parameter Symbol Value Unit Distributed load (SLS) q 292,4 kN/m

Gate height hgate 8 m Span lspan 41 m Tensile modulus FRP EFRP 31 Gpa Tensile modulus steel Esteel 210 GPa Shear modulus G12 5,6 GPa 4 Moment of inertia of the FRP gate IFRP 3,92 m 4 Moment of inertia of the steel gate Isteel 0.539 m 2 Cross sectional area of teh FRP gate AFRP 2,26 m 2 Cross sectional area of the steel gate Asteel 0,204 m 6 Added mass for a 1 DOF mw 1,42*10 kg 3 density of FRP ρFRP 1766 kg/m 3 density of steel ρsteel 8000 kg/m

The results of the FRP gate and the steel are nearly equal. This is caused by the modulus of rigidities of the FRP and the steel gate and the large added mass:  The EI of the FRP gate and the steel gate are nearly equal which is presented in Equation U-50 which results in a nearly equal spring stiffness.

[ ] [ ]

Equation U-50 Ratio of rigidities of a steel and FRP gate

 The total effective masses of the FRP gate and the steel gate are nearly equal which is presented in Equation U-51. This is caused by the large added mass.

[ ] [ ] [ ]

Equation U-51 Ratio of added masses

U.5.3 VIBRATIONS OF A BENDING BEAM

Also the natural frequencies of a pinned-pinned beam are determined for higher modes to determine the natural frequencies of the first and higher modes. The first, second, and third mode of a simply supported bending beam are presented in Figure U-34. The first mode has one antinode at midspan. The second mode has two antinodes, and the third mode has three antinodes which are presented in Figure U-34.

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Figure U-34 Bending beam vibrations (Spijkers, et al., sd)

The natural angular frequency for a mode is calculated using Equation U-52

( ) √

Equation U-52 Natural angular frequency

In which: n = mode [-]

mw = added mass per running metre

The natural frequencies and the period are calculated with Mathcad for the first 5 modes using the parameters presented in Table U-13. The results for the vibrations of a beam and the results for a 1 DOF system are presented in Figure U-35 and Figure U-36. The results of the first mode approach the natural frequency of the 1 DOF and the periods for higher modes are smaller with respect to the first mode. So, the period of the first mode is the longest period for which the beam vibrates. The natural frequencies and the periods are the nearly equal for the steel and FRP bending beam due to the equal EI and the large added mass.

Figure U-35 Natural frequencies

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Figure U-36 Natural periods

U.5.4 VERIFICATION FOR OTHER HYDRAULIC GATES

It has to be verified whether the calculated natural periods matches the periods of existing hydraulic gates. Measured frequencies and measured periods determined by model tests of the Eastern Scheldt barrier, weir Hagestein, general sluices, and the Hartel barrier are listed in Table U-14. The calculated periods of Figure U-36 are smaller than 0.6 seconds and larger than 0.02 seconds for the first 5 modes which corresponds with the periods presented in Table U-14. So the calculated periods of the gate of weir Culemborg matches with the vibrations of existing gates.

Table U-14 Measured frequencies and periods of vibrating gates for model tests (based on: (Kolkman & Jongeling, 1996)) Measured frequencies Measured periods Weirs and barriers [Hz] [s] Eastern Scheldt barrier 6,5 0,15 Weir Hagestein 0,8 - 2,8 0,36 - 1,25 General open discharge sluice 2,5 - 9,5 0,1 - 0,4 Hartel barrier 0,5 -2,5 0,4 - 2

U.5.5 CONCLUSION

The vibrations of a FRP gate matches with the vibrations of a steel gate and corresponds for both materials with the vibration characteristics of existing hydraulic structures. However further investigation has to be performed for the vibrations of a FRP gate due to the large thickness and the flow characteristics at over and underneath the gate:

U.6 MATHCAD SHEETS

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References

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