Testing to Failure of a 55-Year-Old Prestressed Concrete Bridge in Kiruna

Testing to Failure of a 55-year-old Prestressed Concrete Bridge in Kiruna

Bending, Shear and Punching of Girders and Slab. Fracture Properties of Materials. Test Results, Modelling and Assessment

Final Report BBT 2017-030

Jonny Nilimaa, Rasoul Nilforoush, Niklas Bagge, Lennart Elfgren

Testing to Failure of a 55-year-old Prestressed Concrete Bridge in Kiruna

Bending, Shear and Punching of Girders and Slab. Fracture Properties of Materials. Test Results, Modelling and Assessment

Final Report BBT 2017-030

Jonny Nilimaa Rasoul Nilforoush Niklas Bagge Lennart Elfgren

Luleå University of Technology Department of Civil, Environmental and Natural Resources Engineering Division of Structural and Fire Engineering Cover photo: The Mining Bridge in Kiruna during testing in June 2014. The photo is taken from the old City Hall (NE of the bridge) and shows the slag heap from the LKAB iron ore mine in the background.

ISSN 1402-1528 ISBN 978-91-7790-693-3 (print) ISBN 978-91-7790-694-0 (pdf)

Luleå 2020 www.ltu.se

PREFACE AND ACKNOWLEDGEMENTS This report presents results from tests to failure of a 55-year-old prestressed concrete bridge in Kiruna. It is the final report for project BBT 2017-030 (Trafikverket). It is a continuation of an earlier project, BBT 2013-029, which focused on the planning, testing and load-carrying capacity of the bridge, whereas this project focus on material properties, basic test results and experiences from assessment. The project has mainly been carried out at Luleå University of Technology, LTU, in cooperation with Trafikverket and partners from the Swedish Universities of the Built Environment (SBU). The project has been supervised by Håkan Thun, Trafikverket. Jonny Nilimaa has written most of the report. Rasoul Nilforoush has carried out and presented new fracture mechanics tests. Niklas Bagge and Lennart Elfgren have provided background material and assistance. During the original test campaign and initial analysis, contributions were given by Håkan Johansson, Georg Danielsson, Mats Petersson, Erik Andersson, Lars Åström, Roger Lindfors and Ulf Stenman, MCE/Complab, LTU (Testing); Thomas Blanksvärd, Björn Täljsten, Gabriel Sas, Peter Collin, Ola Enoksson, Ulf Ohlsson, Natalia Sabourova, Carina Hannu and Tarja Karlsson, Structural Engineering, LTU (Planning and Execution); Niklas Grip, Mathematics, LTU (Dynamic Measurements and Analysis); Patrik Larsson, Building construction, LTU (Management); Kurt Bergsten and Karl Erik Nilssson, Nordisk Spännarmering (Loading); Jan Lundgren and Tony Nordqvist, Luossavaara-Kiirunavaara AB, LKAB (Coordination); Anders Carolin, Håkan Thun and Björn Paulsson, Trafikverket (Guidance and supervision); Zheng Huang and Yongming Tu, Southeast University, Nanjing, China (Damage detection); Mario Plos, Reza Haghani Dogaheh, Mohammad Al-Emrani and Jiangpeng Shu, Chalmers University of Technology (Strengthening; Punching Analysis); Håkan Sundquist, the Royal Institute of Technology (LCC Analysis); Oskar Larsson-Ivanov, Lund University of Technology (Temperature Analysis); Irene and Sven Erik Elfgren, Mölndal (Photo). Total funding for the continuation project has been 1.2 MSEK, whereof 60% from Trafikverket and its program for research and innovation regarding works in the transport sector (Bransch- program för forskning och innovation avseende Byggnadsverk inom Transportsektorn, BBT), Project No 2017-030: “Experiences from testing to failure of a 55-year old prestressed concrete bridge in Kiruna” (Riktlinjer för tillståndsbedömning av betongbroar - Erfarenheter från provning till brott av en 55-årig spännbetongbro i Kiruna). Additional funding has been provided by Hjalmar Lundbohm Research Center, HLRC/LKAB, 0.2 MSEK (17%); the European Research Project In2Rail, 0.17 MSEK (14%); and Luleå University of Technology, LTU, 0.11 MSEK (9 %). The original project (BBT 2013-029) had a total funding of 3.8 MSEK, whereof 47.5 % from Trafikverket, 31.5% from LKAB/HRLC, 10% from SBUF and 11% from LTU. We are grateful for all the interesting results that have come out of the project and extend our sincere thanks to all who have contributed. Luleå in September 2020 Jonny Nilimaa Rasoul Nilforoush Niklas Bagge Lennart Elfgren

II ABSTRACT Results are presented from the testing to failure of a 55-year-old prestressed concrete bridge with five continuous spans and a total length of 121.5 m. The bridge was situated in Kiruna in northern Sweden. Results are given from load, deflection and strain measurements during bending-shear tests of the girders and form a punching test of the slab. The testing was carried out in June 2014. Extensive assessment and modelling of the bridge with finite element methods have taken place and is summarized. The strength of the bridge was much higher than what could be found with ordinary code methods. The advanced non-linear models were, after calibration, able to predict the behaviour in a good way.

In numerical assessments of concrete bridges, the value of the concrete tensile strength fct and the fracture energy GF plays an important role. However, mostly these properties are only estimated based on the concrete compressive strength using empirical formulae. In order to study methods to determine the concrete tensile strength and fracture energy for existing bridges, tests were carried out in 2019 on notched cylindrical concrete cores drilled out from the Kiruna Bridge. Different methods to determine the concrete fracture energy are discussed and recommendations are given for assessment procedures.

III

SVENSK SAMMANFATTNING Bakgrund: Stora samhällsekonomiska vinster kan göras genom förbättrad tillståndsbedömning av befintliga broar. Det kan ge ökad resurseffektivitet i ett mer hållbart byggande, kan utveckla byggprocesserna och kan bidra till kompetensutvecklingen i branschen Syfte och mål: Den övergripande målsättningen med projektet (BBT 2017-030) var att förbättra precisionen vid tillståndsbedömning av befintliga konstruktioner genom att sammanställa erfarenheter från provning av Gruvbron i Kiruna 2014 och att bestämma verkliga materialdata för draghållfasthet och brottenergi för betongen. Projektet fullföljer ett tidigare projekt med tyngdpunkt på genomförandet av fullskaleförsöket (BBT 2013-029). Upplägg, metod och genomförande: Sammanställning av resultat från provningen av Gruvvägsbron i Kiruna. Utformning av metod för provning av betongens draghållfasthet och brottenergi med efterföljande provning av betong från Gruvvägsbron. Sammanställning av teknisk slutrapport. Koppling till och relevans för branschprogrammet: Projektet besvarar frågor i branschprogrammet om (1) hållbart byggande, (2) uppföljning och utveckling av befintliga konstruktioner beträffande, säkerhet, bärighet, funktion och miljö. Projektet stöder utveckling av (3) byggprocesser och (4) kompetensutveckling genom att ny kunskap och insikt sprids. Projektorganisation och medverkande parter: Projektet genomfördes vid Luleå tekniska universitet (LTU). En projektgrupp av forskare leddes av en styr- och referensgrupp med representanter från LTU, KTH, Chalmers, Trafikverket och Sveriges Bygguniversitet. Resultat: Detaljerade resultat från brottprovningen sammanställs i denna rapport (kap. 6 - 7) liksom utformningen av en testmetod för direkta dragförsök på utborrade betongkärnor för att erhålla betongens draghållfasthet och brottenergi (kap. 8). Resultaten visar att både draghållfastheten och brottenergin kan uppgraderas genom provning av betongen i gamla broar. Detta förfarande utgör en potential för uppgradering av hållfastheten för befintliga broar och en möjlighet till att spara resurser och skona miljön genom ett bättre utnyttjande av kapaciteten i våra befintliga konstruktioner (kap. 8 - 9). Spridning av resultat: Resultaten har presenterats i två doktorsavhandlingar, i ett 30-tal tidskrifts- och konferensartiklar samt vid branschseminarier och i undervisning vid de tekniska högskolorna i Sverige.

TABLE OF CONTENTS PREFACE AND ACKNOWLEDGEMENTS ...... II ABSTRACT ...... III SVENSK SAMMANFATTNING ...... IV TABLE OF CONTENTS’ ...... V 1. INTRODUCTION ...... 1 1.1 Background ...... 1 1.2 State of Art ...... 1 1.3 Overview of contents ...... 2 2. THE KIRUNA BRIDGE ...... 3 2.1 General description ...... 3 2.2 Geometry ...... 3 2.3 Material ...... 5 3. TEST PROCEDURE ...... 6 3.1 General description ...... 6 3.2 Strengthening ...... 6 3.3 Preloading ...... 7 3.4 Bridge girder failure test ...... 9 3.5 Bridge slab failure test ...... 9 3.6 Assessment of post-tensioned cables ...... 9 3.7 Material tests ...... 10 4. INSTRUMENTATION ...... 13 4.1 General description ...... 13 4.2 Girder test ...... 13 4.3 Slab test ...... 18 4.4 Non-destructive test of prestressing ...... 19 5. GIRDER TEST RESULTS ...... 20 5.1 General. Application of Loads ...... 20 5.2 Longitudinal Strains ...... 23 5.3 Displacements ...... 26 5.4 Column strains ...... 29 5.5 Curvature ...... 32 5.6 Temperature ...... 34 5.7 Strain in vertical reinforcement ...... 35 5.8 Strain in Near Surface Mounted Reinforcement (NSMR) and laminates ...... 38 6. SLAB TEST RESULTS ...... 40

7. FRACTURE MECHANICS MATERIAL TESTS ...... 44 7.1 General ...... 44 7.2 Tensile tests ...... 45 7.3 Comparison with empirical models ...... 49 7.4 Discussion. Estimated and Measured Concrete Properties ...... 51 8. ASSESSMENT METHODS ...... 52 8.1 General ...... 52 8.2 Prestressing forces ...... 54 8.3 Flexural-Shear Failures of Girders ...... 54 8.4 Punching Failure of Slab ...... 59 8.5 Damage Detection and Life-Cycle Assessment ...... 60 8.6 Final Demolition of bridge ...... 61 8.7 Outlook ...... 62 9. CONCLUSIONS ...... 64 REFERENCES ...... 66

Appendix A – Drawings Appendix B – Fracture Energy - Summary of uniaxial tests Appendix C – Fracture Energy - Paper submitted to NCR Symposium, Sandefjord, 2020 Appendix D – Fracture Energy - Paper submitted to IABSE Symposium, Wroclaw, 2020 Appendix E – Punching - Paper submitted to Journal of Bridge Engineering Appendix F – Rough basic assessments Appendix G – Results from preloading

VI 1. INTRODUCTION

1.1 Background In order to meet current and future demands for sustainability and structural resistance, existing bridges might be in need for repair, upgrading or replacement. For instance, responses to a questionnaire by infrastructure managers in 12 European countries, acquired and analysed in the European Research FP7 Project MAINLINE, indicated a need for strengthening 1500 bridges, replacing 4500 bridges and replacing 3000 bridge decks in Europe during the coming decade, Mainline D1.1 (2013). A Swedish Government Proposal 2012 recommended an investment of SEK 522 billion (EUR 60 billion) from 2014 to 2025, to meet transport infrastructure requirements in Sweden, Reinfeldt & Elmsäter-Svärd (2012). With adjustment for inflation this represents a 20 % increase relative to the previous investment level, as detailed in an earlier Proposal, Reinfeldt & Torstensson (2008), indicating a need for substantial actions to maintain robust and sustainable infrastructure. Due to the major social, economic and environmental benefits of avoiding demolition and reconstructing of existing bridges, they should be repaired and strengthened rather than replaced in cases where this is cost-effectively feasible, Sustainable Bridges (2007), Power (2008), Mainline D1.2 (2013). Thus, advanced methods should be used for accurately assessing bridges’ condition, Bien et al. (2007), SB 6.2 (2007), and identifying the optimal operations to maintain, strengthen or replace them, from a perspective based on life-cycle cost minimisation, Jalayer et al. (2011), Ye et al. (2020). .

1.2 State of Art To obtain reliable assessments of existing bridges, which are crucial for rigorous life-cycle cost analysis, it is essential to address current uncertainties regarding key variables, such as structural and loading parameters and possible deterioration mechanisms, Frangopol. (2011). In the past decade monitoring concepts have been developed to update models for bridge assessment, reducing the uncertainties, based on empirical data, Frangopol et al. (2008). Moreover, proof loading has been suggested, Faber et al. (2000), and subsequently implemented for reinforced concrete structures in ACI Standard 437.2 (2014), as an approach to verify the reliability of relevant models and reduce uncertainties regarding the true condition of existing bridges. Thus, testing and monitoring of bridges at service-load levels is now an accepted and well-known approach for assessment, Lantsoght (2019), Elfgren et al. (2018, 2019). Detailed, large-scale laboratory tests of bridges and their materials have been reported by e.g. Amir (2014) and Nilimaa (2013, 2015). Destructive investigations have also been described of prestressed concrete by e.g. McClure & West. (1984), of post-tensioned concrete, by e.g. Schmidt et al. (2014), Oh et al. (2002) and Täljsten et al (2019), and of non-prestressed reinforced concrete bridges by e.g. Jorgensen & Larson (1976), Scanlon & Mikhailovsky (1987), Miller et al. (1994), Ross (2007), Zhang et al. (2011), and Puruula et al. (2013, 2015). However, such studies have generally focused on specific components or elements, for instance, the bridge slab tested by Schmidt et al. (2014). Few complete full-scale bridges have been tested to failure in order to improve understanding of their true structural behaviour, and rigorously calibrate methods and models, Bagge et al. (2018a). Hence, more comprehensive empirical information is required on the behaviour of concrete bridges, especially of prestressed and post- tensioned concrete, as they approach failure, and cost-effective methods to avoid risks of failure.

1.3 Overview of contents In the study presented here a 55-year-old post-tensioned concrete bridge, see Figure 1-1, was thoroughly instrumented (with up to 141 sensors) and tested to failure. The aims were to calibrate and refine methods and models for assessing existing reinforced concrete bridges, and to assess the utility of methods using carbon fibre reinforced polymers (CFRPs) for upgrading reinforced concrete structures, Nilimaa (2015), Bagge (2014, 2020), Bagge et al. (2018b). Since there have been few full-scale tests on post-tensioned bridges, a focus was on assessment of the post-tensioned system. The complete test and measuring programme is described here, together with the results. Parts of the investigations have been presented earlier in e.g. the theses of Nilimaa (2015) and Bagge (2017), in the BBT Report 2013-029, see Elfgren et al (2015), and in many conference and journal papers. Here we now try to give an over-all summary of the results. The bridge is presented in chapter 2. The test procedure is detailed in chapter 3 and the instrumentation in chapter 4. Original graphs from sensors used in the girder and slab tests are given in chapter 5 and 6 respectively. New fracture mechanics tests are presented in chapter 7. Earlier presented work on assessment methods and results are summarized in chapter 8. Finally, conclusions are given in chapter 9.

Figure 1-1. Photograph of the Kiruna Bridge from the north-east, showing the slag heap from the LKAB iron ore mine in the background (2014-06-25).

2. THE KIRUNA BRIDGE

2.1 General description The Kiruna Bridge, located in Kiruna, Sweden, was a viaduct across the European route E10 and the railway yard close to the town’s central station, see Figure 2-1. It was constructed in 1959 as part of the road connecting the city centre and the mining area owned by Luossavaara- Kiirunavaara AB, LKAB . The bridge was designed by Meemo Trepp at “Civilingenjör Sven Hultquist konsulterande ingenjörsbyrå”, Stockholm, and was built in 1960 by Contractor AB, Skellefteå. The sub-level caving method for extracting the ore in the Kiruna mine causes subsidence. Thus, to ensure the continuing utility of the Kiruna Bridge, in 2006 LKAB initiated geodetic position measurements of the bridge supports. In 2008 Luleå University of Technology (LTU) started to monitor the bridge continuously, Enochsson et al. (2011). Due to ongoing subsidence, LKAB decided to permanently close the bridge in October 2013 for demolition in September 2014, providing an opportunity for LTU to test it to failure in May-August 2014.

Figure 2-1. The Kiruna Bridge from south during winter before testing. The bridge crosses the Iron Ore Railway Line from Luleå to Narvik to the left and the road E10 to the right.

2.2 Geometry The bridge was a 121.5 m continuous post-tensioned concrete girder bridge with five spans: 18.00, 20.50, 29.35, 27.15 and 26.50 m long, Figure 2-2. According to construction drawings both the longitudinal girders and bridge slab in the western part (84.2 m) were supposed to be curved with a radius of 500 m. However, inspection of the actual geometry showed that the slab’s girders consisted of straight segments with discontinuities at the supports. Moreover, there were 5.0 % and 2.5 % inclinations in the longitudinal and transverse directions, respectively. Longitudinal movements of the bridge were allowed at the eastern abutment by three rolling bearings (support 6 in Figure 2-2), while the western abutment was hinged (support 1). Devices were installed at the bases of the intermediate supports 2-5, each consisting of three columns, in 2010 to enable vertical adjustment of the supports to counter possible uneven settlement of the basements, Bagge et al. (2015c).

The superstructure consisted of three parallel, 1923 mm high, longitudinal girders connected with a slab on top, Figure 2-3. Including the edge beams the cross-section was 15.60 m wide, and the free distance between the girders was 5.00 m. In the spans, the girders were 410 mm wide, gradually increasing to 650 mm, 4.00 m from the intermediate supports and widened to 550 mm at anchorage locations of the post-tensioned cables, two fifths of the span lengths west of support 3 and three tenths of the span length east of support 4. The bridge slab was 300 mm thick at the girder-slab intersection and 220 mm 1.00 m beside to the girders. The Kiruna Bridge was post-tensioned in two stages with the BBRV system. In the first stage, six cables per girder were post-tensioned in each end of the central segment. In the second stage, four and six cables per girder were post-tensioned from the free end of the western and eastern segments, respectively. Each cable was composed of 32 wires with a 6 mm diameter.

Figure 2-2. Geometry of the Kiruna Bridge and location of the load application in the test programme. There is a hinge in section 1 and a roller bearing in section 6

1500 12000 1500

Southern Central Northern girder girder girder Figure 2-3. Cross-section of the Kiruna Bridge.

The girders were each reinforced with three 16 mm diameter bars at the bottom, and 10 mm diameter bars at the sides with either 150 mm spacing for the central girder or 200 mm for the

others. The vertical reinforcement also consisted of 10 mm diameter steel bars with 150 mm spacing. The concrete cover was 30 mm thick, except for the 16 mm diameter reinforcement bars, for which the horizontal concrete cover was 32 mm thick. Original Drawings are given in Appendix A. The bridge was originally designed according to Provisional Regulations of the Royal Civil Engineering Board, KVVS (1955). Before the tests, the pavement on the slab was removed from the road crossing the bridge.

2.3 Material According to construction drawings the concrete quality in the substructure and the superstructure was K 300 (fcc ≥ 30 MPa) and K 400 (fcc ≥ 40 MPa), respectively, while the reinforcing steel quality was generally Ks 40 (fy = 400 MPa), except in the bridge slab (Ks 60, fy = 600 MPa). The steel quality for the post-tensioned reinforcing BBRV reinforcing system was denoted St 145/170 (fy/fu = 1450/1700). The bridge was constructed in accordance with the National Steel Regulation, SOU (1938), and the National Concrete Regulations, SOU (1949).

The concrete compressive strength (fcm,is and fck,is) and the modulus of elasticity (Ecm,is and Eck,is) were in the first stage of material evaluation determined from 25 standardised tests on concrete cylinders with 100 mm diameter and 200 mm height, according to the European standards SS-EN 12390-3 (2009); SS-EN 12390-13 (2013) and SS-EN 12504-1 (2009).

3. TEST PROCEDURE

3.1 General description An experimental programme was designed to assess the behaviour and load-carrying capacity of the bridge using both non-destructive and destructive test procedures. For safety reasons, related to continuing use of the European route E10 during the tests, the experimental programme was developed for loading in span 2-3, with associated monitoring in spans 1-4, Figure 2-2. The experimental programme can be summarised by the following, chronological steps: 1. Non-destructive determination of residual post-tensioned forces in cables in span 2-3 (May 27-28, 2014). 2. Preloading Test Schedule 1, of un-strengthened bridge girders, including destructive determination of residual post-tensioned forces in cables in span 2-3 (June 15-16, 2014). 3. Preloading Test Schedule 2, of strengthened bridge girders (June 25, 2014). 4. Failure test of the bridge girders (June 26, 2014). 5. Failure test of the bridge slab (June 27, 2014). 6. Complementary non-destructive determination of residual post-tensioned forces in cables in midspans 1-4 (June 27 and August 25, 2014). 7. Material tests of concrete, reinforcing steel and post-tensioned steel. 8. Condition assessment of post-tensioned cables.

Steps 1-6 were carried out at the Kiruna Bridge, with the test dates in parenthesis. Steps 7-8 were took place in the Complab laboratory at LTU after demolition of the bridge.

3.2 Strengthening The experimental programme included tests of two separate systems for strengthening concrete structures using carbon fibre reinforcing polymers, which were attached to the lower sides of the central and southern girders in span 2-3 (see Figure 3-1 and Figure 4-5). However, the northern girder remained un-strengthened. A system of three near-surface mounted (NSM) 10x10 mm2 Carbon Fibre Reinforced Polymer (CFRP) rods was installed in the concrete cover of the central girder, SB 6.2 (2007), SB 6.3 (2007). The bar lengths were limited to 10.0 m, due to transportation constraints, thus several overlaps (1.0 m) were required to apply the strengthening over the entire span length. A set of full-length CFRP rods was installed centrically in span 2-3, with sets of 5.80 and 5.74 m CFRP rods on the western and eastern sides, respectively To strengthen the southern girder, a system of three 1.4x80 mm2 prestressed CFRP laminates was applied to the blasted concrete surface, Al-Emrani et al. (2013), Kliger et al. (2014). The lengths of the middle and outer laminates were 14.17 and 18.91 m, respectively, in order to provide space for the anchorage device at each end. Each laminate was tensioned to 100 kN at the eastern end, controlled with a load cell, as the force was applied using a manually operated hydraulic jack. The force was gradually transferred to the concrete by the anchorage device. In this manner no force is expected to be transferred at the end, while it is fully transferred after

6 1.20 m. The anchorage devices were attached to the bridge until disassembly after Preloading Test Schedule 2. This was the first reported full-scale test to failure of the method using prestressed laminates with these innovative anchor devices. For the CFRP rods and laminates, denoted StoFRP Bar IM 10 C and, StoFRP Plate IM 80 C respectively, the mean modulus of elasticity and tensile strength were 210 GPa (200 GPa) and 3300 MPa (2900 MPa), respectively, with mean values specified in parenthesis. Epoxi StoPox SK41, a commercially available and CE-approved thixotropic epoxy adhesive, was used to bond both strengthening systems. 4 hydraulic jacks with cables anchored in bedrock N 4 3 2 1

Beam 2 Beam 1

3 prestressed 3 NSM CFRP CFRP laminates rods (10x10 mm2) (1.4x80 mm2)

Figure 3-1. Arrangement for loading the bridge girders in midspan 2-3.

3.3 Preloading In span 2-3 two welded steel beams (outer dimensions 700x1180x5660 and 700x1180x6940 mm3) were arranged horizontally to apply loads in the midspans of each girder, see Figure 1-1 and Figure 3-1. They consisted of a double web (thickness 15 mm) with flanges (thickness 30 mm) and were of the steel grade S355J0. The beams were supported by steel load distribution plates (steel grade S275JR), with areas of 700x700 mm2 and total thicknesses ranging from 20 to 265 mm, due to the inclination of the bridge slab. A horizontal concrete surface was also cast locally under the plates. The bridge was loaded using four hydraulic jacks with cables, threaded through drilled holes in the bridge slab, anchored over a length of 14.60 m in the bedrock, as illustrated in Figure 3-1. The distances from the centre of the jacks and cables to the centre of the support of the transverse steel beam were 885 mm. The capacity of the jacks was approximately 7.0 MN, with a 150 mm stroke length. The piston cross-section area was 1282 cm2 for jacks 1, 3 and 4, and 1284 cm2 for jack 2. Each cable consisted of 31 wires with diameters of 15.7 mm (5/8 inch).. The bridge was preloaded by applying two schedules of incrementally increasing loads using the four jacks to both strengthened and un-strengthened girders in midspan 2-3, as listed in Table 3- 1 and illustrated in Figure 3-1. The schedules were designed to reach the cracking load of the girders, as predicted by preliminary nonlinear finite element analysis. Before the force- controlled loading to a specified level, given by the actual load case, the bridge was unloaded.

To ensure no drift in the measurements and stable loading, peak pressure was maintained for load cases 7, 9, 13 and 29. Load cases 15-18 in Schedule 1 were designed to determine the remaining forces in the post-tensioned cables (see Section 3.6). Some results from the preloading ests are given in Appendix G. Table 3-1. Load cases for preloading the un-strengthened and strengthened bridge girders in midspan 2-3. Jack 1 Jack 2 Jack 3 Jack 4 Load case kN kN kN kN 11,2 500 250 250 500 21,2 500 500 - - 31,2 - - 500 500 41,2 1000 1000 - - 51,2 - - 1000 1000 61,2 1500 1500 - - 71,2 1500 1500 - - 81,2 - - 1500 1500 91,2 - - 1500 1500 101,2 500 250 250 500 111,2 1000 500 500 1000 121,2 1500 750 750 1500 131,2 1500 750 750 1500 141,2 2000 1000 1000 2000 151,2 2000 1000 1000 2000 161,2 2000 1000 1000 2000 171 2000 1000 1000 2000 181 2000 1000 1000 2000 191 500 500 - - 201 - - 500 500 211 1000 1000 - - 221 - - 1000 1000 231 1500 1500 - - 241 - - 1500 1500 251 500 250 250 500 261,2 1000 500 500 1000 271,2 1500 750 750 1500 282 2000 1000 1000 2000 292 2000 1000 1000 2000 1 Load case for preloading the un-strengthened girder 2 Load case for preloading the strengthened girder

3.4 Bridge girder failure test Preloading was followed by a test to failure of the strengthened girders, according to the setup described in the previous section. Each girder was equally loaded to 12.0 MN in total (the approximate load-carrying capacity predicted by preliminary nonlinear finite element analysis): 4.0 MN delivered by the outer jacks and 2.0 MN by the inner jacks. The pressure in jack 4 was subsequently increased to failure of the southern girder and then the pressure in jacks 2 and 3 was increased to failure of the central girder, while the settings of the other jacks remained unchanged, so they provided approximately constant loads. The jacks’ grip positions were changed as necessary to accommodate deflections exceeding the stroke length.

3.5 Bridge slab failure test The bridge slab in midspan 2-3 was tested to failure using an arrangement similar to load model 2 (LM 2) described in Eurocode 1, SS-EN-1991-2 (2003), with its centre located 880 mm from the outer side of the northern girder (Figure 3-2). By rotating steel beam 1 (Figures 3-1 and 3-2), hydraulic jack 1 was reused to apply load on the slab, through two 350x600x100 mm3 steel plates spaced 2.00 m apart. A horizontal concrete surface was also cast locally under the plates. Due to the widening of the bridge girders at the anchorages of the post-tensioned cables, the distances from the centres of the western and eastern load distribution plates to the inner sides of the girders were 470 and 330 mm, respectively. As in the previous tests, the loading was force controlled.

3.6 Assessment of post-tensioned cables The residual force in the post-tensioned cables was non-destructively determined by monitoring strains at the lower surface of each girder resulting from gradually cutting the concrete with a saw on both sides of a strain sensor, Kukay et al. (2020), placed one-tenth of the span length west of midspan 2-3, before the bridge and slab failure tests. After the failure tests, the procedure was also applied to each girder in midspan 1-2, the northern girder in midspan 2-3 and the central and southern girders in midspan 3-4. In order to keep the reinforcing steel intact, the arrangements of sensors and saw cutting lines were designed to avoid cutting either the stirrups or longitudinal reinforcing steel. The cutting proceeded to an approximate depth of 35 mm, or the actual depth of the longitudinal reinforcing steel. All the non-destructive tests were carried out without applying external loads. As part of Preloading Test Schedule 1, the cracking moment test according to Osborn et al. (2012) was applied to calibrate the non-destructive test method. During load cases 1-14 cracks formed, and instruments described in the next section were used to monitor the behaviour of selected cracks and adjacent areas between load cases 14 and 15. Thus, the remaining force in the post-tensioned cables can be determined from data acquired from load cases 15-18, based on the sequence of reopening of the cracks, Bagge et al. (2017).

1 hydraulic jack with cables

rans anchored in bedrock direction

Bridge t verse verse direction 1

Beam 1 350 600 700 700 600

Figure 3-2. Arrangement for loading the bridge slab in span 2-3.

3.7 Material tests To determine characteristics of the bridge’s materials, tests of the concrete, reinforcing steel and post-tensioned steel were also carried out. Thus, before the tests described here at least six concrete cylinders were drilled out from the superstructure in both midspans 1-2 and 3-4, and each of the columns at support 4. In addition, during demolition of the bridge several 10, 16 and 25 mm diameter steel bars, and a specimen of the post-tensioned cables, were obtained for uniaxial tensile tests., Bagge (2017). Concrete Characteristic values as specified in the Swedish design and assessment codes (fck and Eck) are given for the quality classes for the substructure and the superstructure as specified in original construction drawings in Table 3-2. In accordance with the Swedish assessment code TDOK 2013:0267 (2017), version 4, it is permitted to assume improved concrete properties (fck,upgr and Eck,upgr) when assessing bridges which have been designed based on regulations between 1947 and 1960. Section 1.3.2.1.7 indicates that for up to K35 you may increase with three concrete classes (for K30 from 21,5 to 32 MPa) and for K40 you may increase with two classes (from 28,5 to 35,5 MPa)

The mean values of the tested material properties of the concrete (fcm,is and Ecm,is) are given together with coefficients of variation in parentheses. However, tests of concrete taken from the columns, girders and slab indicated similar compressive strengths regardless of the structural part, with an overall average value of 62.2 MPa and a notably high coefficient of variation (16 %). Thus, material tests showed that the code upgrade increase is appropriate and, in this particular case, the favourable time-dependent effects on the compressive strength were even larger than given by the recommendations. The reason for the increase is partly that the cement was more roughly grinded than today and had not reached its full strength after 28 days, Thun (2006), Thun et al. (2006).

Table 3-2. Results from the first stage of material tests on concrete, Bagge (2017). Specified values for quality classes: original (ck) and upgraded (ck,upgr); test mean values (cm,is). Compressive strength Modulus of elasticity

Quality Structural No of fck fck,upgr fcm,is fck,is Eck Eck,upgr Ecm,is class element tests MPa MPa MPa* MPa GPa GPa GPa

K300 Column 7 21.5 32.0 61.8 56.8 30.0 33.0 32.4 (11%) (6.8%) K400 Girder 11 28.5 35.5 56.4 47.2 32.0 34.0 31.2 (19%) (10%) K400 Slab 7 28.5 35.5 71.6 66.6 32.0 34.0 33.3 (3.2%) (5.6%) All 25 62.2 47.2 32.1 (16%) (8.3%) *Coefficient of variation in parentheses

Reinforcement steel Uniaxial tensile standard tests were carried out to determine the stress-strain response of the reinforcing steel, characterised by the yield strength or the 0.2 % proof strength for the prestressing steel (fym,is and fyk,is), the tensile strength (ftm,is and ftk,is) and the strain at the peak stress (εum,is and εuk). The measured values for each steel quality and bar diameter (Ø) have been summarised in Table 3-3, together with the properties fyk, ftk and uk specified in the Swedish assessment code TDOK 2013:0267 (2016). The tests followed the procedure given by the European standards SS-EN ISO 6892-1, SS- EN ISO 15630-1 and SS-EN ISO 15630-3. In some of the tests, conditions and results deviated from the requirements and those were excluded from the determination of the characteristic values. In general, the tested steel properties were consistent with the values specified in the assessment code and, for the yield- and tensile strength, a relatively small spread in the test results was observed. The tests can also be compared to the test of the reinforcement in a 50-year-old prestressed bridge in Örnsköldsvik. There it was found that for corroded Freyssinet 7 mm prestressing strands the strain at maximum stress was ca 4% for corroded strands while it was more ductile with an elongation of ca 7 % for non-corroded bars, Appendix B.3 in SB7-3 (2006), Sederholm (2005).

Table 3-3. Results from the material tests on reinforcement steel. Specified characteristic values for quality classes (k), test mean values (m,is) and new characteristic values based on test values (k,is). Yield strength Tensile strength Strain at peak stress Quality Ø No. of (b) (a)(c) (a) class mm tests fyk fym,is fyk,is ftk ftm,is ftk,is εuk εum,is εuk,is MPa MPa* MPa MPa MPa* MPa* % %* %* 1606 1734 4.74 St145/170 6 5 1450 1530 1700 1682 3.5 4.08 (1.4%) (6.9%) (5.1%) 484 702 13.5 Ks40 10 10 410 412 600 629 16 10.1 (5.8%) (2.9%) (12%) 439 705 12.9 Ks40 16 3 410 403 600 654 16 11.1 (2.4%) (6.3%) (5.1%) 389 629 13.7 Ks40 25 5 390 332 600 543 16 10.3 (4.3%) (3.3%) (9.2%) 679 1000 9.9 Ks60 10 8 620 585 750 920 16 8.70 (5.0%) (1.7%) (5.5%) 584 831 11.5 Ks60 16 6 620 545 750 779 16 10.0 (2.1%) (2.4%) (5.4%) a) coefficient of variance in brackets, b) fp0.2k for quality class St145/170, c) fp0.2m,is for quality class St145/170

4. INSTRUMENTATION

4.1 General description To evaluate the bridge’s behaviour a comprehensive measuring programme was designed. This section summarises the instrumentation used to measure changes in monitored variables during the bridge girder and slab tests and the non-destructive tests with no external load. In addition, measurements during strengthening were carried out according to the description in previous section. Before initiating any experimental investigation, existing cracks in the entire span 2-3 and the half-spans 1-2 and 3-4 adjacent to the loaded span were mapped. The focus was on cracks in the girders, the crossing beams and the slab at the loading point. In order to follow the formation of cracks, the mapping was repeated after each test sequence. The cracks were mapped manually, and their widths were not measured, apart from several cracks specified in the measuring programme, Bagge et al. (2014). In addition to the monitoring during bridge loading, long-term measurements were carried out during the nights before Preloading Test Schedules 1-2 and the failure test of the girders. The durations of the monitoring on these occasions were 22398, 21613 and 45558 s, respectively, and the same instrumentation was used as in the followed bridge loading, excluding manual measurements. Moreover, the bridge was examined when the anchorage device for the prestressed CFRP laminates was disassembled. Most measurements of the bridge were generally acquired at a sampling frequency of 5 Hz, except for the long-term measurements (1 Hz).

4.2 Girder test A battery of instruments was installed before the tests of the longitudinal bridge girders to obtain as comprehensive measurements as possible, within budgetary constraints, of the resulting forces, displacements, curvatures, strains and temperatures. These measurements were complemented by monitoring using several video and still cameras. Data were acquired from all the instrumentation described in this section during the full programme of tests of the bridge girders unless otherwise stated. Force The applied load on the structure was measured by monitoring the oil pressure in each hydraulic jack (1-4), illustrated in Figure 3-1, using UNIK 5000 sensors (GE Measuring and Control; A5075-TB-A1-CA-H1-PA), which have a measuring range between 0 and 600 bar. Displacement Displacements of the bridge were measured using the following instruments. Draw-wire displacement sensors (MICRO-EPSILON; WDS-500(1000)-P60-CR-P) were installed to measure deflections at positions 1-10 and 13-15 (Figure 4-1): in midspan 2-3 on the lower sides of the girders, and lower sides of the crossing beams 500 mm from the outer columns (positions 4-5 and 9-10). All these sensors had a measuring length of 500 mm except those used at positions 6-8 (1000 mm). Twisted lines connected each sensor to a reference point on the ground or the basement.

DB16-17 DB13-15 DB1-3

DB10-12 DB4-6 DB7-9

1 2 3 4 5 6

1 4 6 9 13 2 7 14 16 10 17 3 5 8 11 15

Figure 4-1. Positions of bridge displacement sensors. At positions 11 and 12 both the longitudinal and transverse displacement were monitored using Noptel PSM-200 sensors. The reference point for the horizontal displacement of the bridge slab at the centre line of support 3 was 150 mm perpendicularly away from the southern side of the basement. A transmitter was installed on the basement and a receiver on the lower side of the bridge slab, oriented vertically to the transmitter. The Noptel PSM-200 sensors were only active during the failure test of the bridge girders. In addition, the displacement of the basements’ upper side at support 2-3 was manually measured during the girder failure test, 500 mm against the centre of the bridge in transverse direction, in relation to positions 4-5 and 9-10, and the reference point was an unaffected point beside the bridge. For safety reasons, the incremental monitoring proceeded until a certain load was reached, 9.0 MN in total. At positions 16-17 (Figure 4-1) longitudinal displacements of the upper part of the rolling bearings, i.e. the lower side of the girders, was measured using linear displacement sensors (Micro-Measurements; HS 100) with a 102 mm measurement range, and positions in the abutment as reference points. In Preloading Test Schedule 1, load cases 15-27, the width of one crack in the centre of the lower side of each girder (110, 910 and 1380 mm east of midspan for the northern, central and southern girders, respectively) was measured, using crack opening displacement sensors (EPSILON; 3541-010-150-ST) with the measuring range of 10 mm. Data were also acquired from the sensor on the girder strengthened with laminates during Preloading Test Schedule 2 and the bridge girder failure test. Curvature The curvature at support 2, support 3 and midspan 2-3 was measured over distances of 4.82, 5.08 and 5.00 m, respectively, using rigs composed of steel beams, supported at the ends, and five linear displacement sensors with 800 mm spacing based. At the supports the rigs were located on the bridge slab, while the midspan rig was located under the girder. Due to the discontinuities at the supports, i.e. changes in directions of the girder, and straightness of the curvature rigs, the instrumentation was installed along the line of the central girder in span 2-3. The sensors were HS 100, HS 50 and HS 25 instruments (Micro-Measurements) with measurement ranges of 102, 51.5 and 26 mm, respectively, set at the positions increasingly distant from the centre of the rigs.

14 Strain Strain gauges supplied by Kyowa were installed on the longitudinal and vertical reinforcing steel bars, CFRP rods and laminates, and the concrete surfaces of both the columns at supports 2-3 and next to some major cracks during Preloading Test Schedule 1, load cases 15-27. All of these gauges had 120 ohm resistance, and those installed on the longitudinal reinforcing steel, stirrups or CFRPs and concrete had measuring lengths of 10, 5 and 60 mm, respectively (KFG- 10-120-C1-11L1M3R, KFG-5-120-C1-11L1M3R, KC-60-120-A1-11L1M3R). In total, 35 strain gauges were systematically arranged on the longitudinal reinforcing steel bars: at sections A-K in Figure 4-2 and cross-section positions 1-12 in Figure 4-3, 1879 mm from the centre lines of the supports on each side (B and J), and 1433 and 2226 mm from each side of midspan 2-3. The locations of the sections were at angles of 45° to the centre line of the supports and 60° and 45°, respectively, to the load distribution plates. On the reinforcement bars they were installed in the corners of the closed stirrups (Figure 4-3) except at positions 2, 5, 8 and 11, where they were located 1248 mm from the lower side of the girders. All the strain gauges, apart from those in the bridge slab, were placed on the side of the girders. The locations of the sensors for each section and cross-section position, are specified in Table 4-1. Due to the greater width of the girder in sections G-H and the corresponding increase in concrete cover, strain gauges 25-28 were not used in the final measuring programme. Care was taken to avoid damaging the girder in any way that could potentially affect the quality of the strengthening.

H C G F B K E A J D I

A D B I C E J F K G H Figure 4-2. Positions of strain gauges on longitudinal reinforcing steel. N 12 9 6 3

11 8 5 2

10 7 4 1 Figure 4-3. Cross-section positions of strain gauges on longitudinal reinforcing steel.

Table 4-1.Positions of strain measurements on longitudinal reinforcing steel. No. Position1 No. Position1 No. Position1 1 A6 14 E7 27 H7 2 A12 15 E4 28 H4 3 B1 16 F1 29 I6 4 B2 17 F2 30 I12 5 B3 18 F3 31 J1 6 B6 19 F4 32 J2 7 B10 20 F5 33 J3 8 B11 21 F6 34 J6 9 B12 22 F7 35 J10 10 C6 23 F8 36 J11 11 C12 24 F9 37 J12 12 D7 25 G7 38 K6 13 D4 26 G4 39 K12 1 Section A-K in Figure 4-2 and cross-section position 1-12 in Figure 4-3. As illustrated in Figure 4-4, strain gauges were also installed in three lines on the vertical reinforcing steel on the northern side of the southern girder in span 2-3, at 900 mm spacing starting from the edge of the loading plate. Thus, strain gauges 6-9 were located 1250 mm from the central point of the load application. Vertical distances from the bottom side of the girders to the sensors were 148, 548, 948 and 1348 mm, respectively. In addition, an ARAMIS system in 5M configuration was used to optically record deformations of the surface on the southern girder on the opposite side to the instrumentation of the vertical reinforcing steel, and accompanying software was utilised to analyse the strains. The optical monitoring was based on a grid, centred 2.0 m west of midspan 2-3, from the bottom of the girder. Thus, strain gauges 3-4 according to Figure 4-4, on the opposite side of the girder, were located within the monitored area, which theoretically covered 1050x880 mm, see Figure 5-33. LKAB Midspan 2-3 Vertical Girder-slab reinforcement intersection

8 5

7 4 2

6 3 1

Longitudinal reinforcement Figure 4-4. Positions of strain gauges on vertical reinforcing steel.

In total 14 strain gauges were installed on the Near Surface Mounted Carbon Fibre Reinforced Polymer (NSM CFRP) rods and 10 on the prestressed CFRP laminates (Figure 4-5): gauges 1-8 recorded the strain at the western edge of the NSM strengthening; 9, 10 and 15 were located in the sections equipped with strain gauges on longitudinal reinforcing steel; 11 and 16-18 at midspan 2-3; 14-16 at major concrete cracks and 19-24 next to the anchorage of the laminates. The methods are described in more detail in Al-Emrani et al. (2013), Kliger et al. (2014), Nilimaa (2015), SB 6.2 & 6.3 (2007), Puurula et al. (2013, 2015) and Täljsten et al. (2018). To obtain the reaction forces in the columns adjacent to the load application in midspan 2-3, i.e. supports 2 and 3, the concrete strains were measured by installing a sensor 800 mm above the bottom in the centre of each side of each column. Before the bridge tests, the methodology of using strain gauges to determine the reaction forces was validated using load cells, while preloading the column with hydraulic jacks and utilising the column’s vertical adjustment device.

4500

100 100 100 100

Concrete crack 2 6 910 8 N 1 3 4 5 7 12 50 50

13 14 4500

Central girder: 9 793 1433 910 3 3

5800 1000 5000 10 11 5000 1000 5740

Southern girder: 16 2370 7085 15 17 7085 24 2370

1433 18 4500 2500

4500 500 300 100 50 23

19 20 21 22

Figure 4-5. Geometry of bridge strengthening systems in span 2-3 and positions of the strain gauges on the NSM CFRP rods and prestressed CFRP laminates.

On each side of the cracks instrumented by crack opening displacement (COD) sensors as described above, the concrete strains were measured. Like the COD sensors, the strain gauges were located in the centre of the lower side of the girders. These sensors were only active in Preloading Test Schedule 1, load cases 15-27. Temperature During the experiments temperatures were measured at several locations in midspan 2-3 (Figure 4-6), using type T (04 N/N-24-TT) temperature wires inserted into holes to specified depths in relation to the concrete surface: 30 mm for positions 1, 3 and 6; 60 mm for positions 2 and 7; 50 mm for positions 4 and 8; and 80 mm for position 5.

1000 2500 2500 N

8 4+5

1000

6+7 3 1+2

1000

Figure 4-6. Positions of temperature wires in the concrete at midspan 2-3.

4.3 Slab test Relevant instrumentation that was still intact after the girder failure tests, and complementary instrumentation, was used to monitor the behaviour of the bridge during the following bridge slab failure test. The sensors still active during this test were: - the oil pressure sensor for hydraulic jack 1, as shown in Figure 3-1 and Figure 3-2, - draw-wire sensors 4-7 and 14-15, as shown in Figure 4-1; - strain gauges 1-39 as specified in Table 4-1, excluding gauges 12, 17 and 24-28, which were not used for various reasons; - strain gauges 1-9 as shown in Figure 4-4, except gauge 8, which was out of order; - strain gauges 1-24 installed on the columns at supports 2 and 3; - temperature wires 1-8, as shown in Figure 4-6. Displacement The above instrumentation was complemented with four draw-wire sensors, with similar specifications to the sensors utilised in previous tests. Two were located on the lower surface of the slab, at the centre of the load applications, to measure deflections, and two on the lower side of the northern longitudinal girder, in both cases 2.00 m on either side from midspan 2-3. Curvature To monitor curvature in the slab test, the rigs used in the girder tests at supports 2 and 3 were installed on the top surface of the slab, parallel to the steel beam used for load application, 500 and 1000 mm southern to the centre of the loading plates. The midpoint of this instrumentation coincided with midspan 2-3.

4.4 Non-destructive test of prestressing Three strain gauges of the same type as previously specified for monitoring the concrete were used in the non-destructive tests to determine the residual forces in the post-tension cables, located in a line in the centre of the lower sides of each girder in span 2-3. In order to provide enough space to avoid damaging the sensors while cutting the concrete, the centre-centre distance was 120 mm, since the total length of the strain gauges was 74 mm with a 60 mm measuring length.

5. GIRDER TEST RESULTS

5.1 General. Application of Loads In the experimental programme for the girder tests, the bridge was instrumented with sensors at up to 141 positions in total, excluding the surface measurement using ARAMIS, and 93 sensors were used in the bridge slab failure test. General observations regarding the test procedures and the observed load-carrying capacity of the bridge are presented in this section. The loads applied in the preloading schedules and loading the bridge to failure, according to the recorded pressures in the hydraulic jacks, are illustrated in Figure 5-1 to 5-3, which show that the preloading followed the schedules listed in Table 3-1, with minor deviations due to difficulties in manually controlling the oil pressure. In Preloading Test Schedule 1 (Figure 5-1) the complementary instrumentation used to determine the remaining forces in post-tensioned cables was installed after approximately 7400 s. The time spent installing it (about 5.5 hours including associated operations) is not shown in the graph, but no corrections have been applied to the force-time courses shown in Figure 5-2. In the following sections, the values of the sensors are generally referred to the time scale indicated in Figure 5-3, as this was how the testing was organized. Examples are also given of evaluated results referring to the deflection, see also chapter 8 on assessment. In order to manually follow the basement settlements of the bridge safely, the loading was carried out stepwise up to a certain level, see Figure 5-3. Another reason for the irregularity in the loading procedure was the limited stroke length of the hydraulic jacks, which required the grip position to be changed several times to accommodate larger deflections of the bridge.

Figure 5-1. Observed loadings during Preloading Test Schedule 1, un-strengthened girder. The outer jacks 1 and 4 have about double the loads as the central jacks 2 and 3.

Figure 5-2. Observed loadings during Preloading Test Schedule 2, strengthened girders.

Jack 4 Jack 1

Jack 2 Jack 3

Figure 5-3. Observed loadings during the bridge girders failure test. The southern girder failed at 6304 s and the central at 7724 s. The loads in the jacks No 1, 2, 3 and 4 were 3935, 1961, 1964 and 5446 kN respectively for the southern girder and 3446, 3047, 3075 and 3228 kN respectively for the central girder

After applying a total load of 12.0 MN (4.0 MN for each girder), the pressure in jack 4 was increased to reach failure of the southern girder, while the pressure in the other jacks remained nearly constant. The loads in the jacks No 1, 2, 3 and 4 were 3935, 1961, 1964 and 5446 kN respectively at the failure of the southern girder after 6304 s. However, in the subsequent loading of the central girder to failure, the pressure in jacks 1 and 4 slightly decreased as the central girder was loaded using jacks 2-3, in responses related to the deformations of the bridge. The loads in the jacks were 3446, 3047, 3075 and 3228 kN and the failure in the central girder took place after 7724 s.

Deflections of the bridge are illustrated by the load-displacements curves in Figure 5-4, showing the relationship between the total load and midspan deflection of the three girders. The figure also presents the behaviour according to standard finite element analysis with linear uncracked and cracked behaviour, Bagge (2017). The large variation in modelled stiffnesses shows the importance to use more advanced non-linear analysis for a better modelling of the behaviour. The final bending-shear failure of the south and central girders are also illustrated in Figure 5-4. The peak load at loading the southern girder to failure was 13.4 MN (5.5 MN in jack 1) and it was 12.7 MN for the central girder (6.1 MN in total in jacks 2-3).

Figure 5-4. Load-deflection relationship for the test of the bridge’s girders and photo of south and central girders after shear-bending failure, Bagge (2007).

As indicated in Figure 5-4, the loading of the bridge resulted in extensive concrete cracking at the midspan region of the girders, yielding of longitudinal reinforcement and stirrups and, consequently, large deformations. The south girder finally failed in a sudden manner under a total external load of 13.4 MN at 6304 s. Video monitoring of the test showed a complex failure mechanism. Failure initiation was indicated in the bridge deck slab and in the upper part of the girder where diagonal cracking was formed. Rupture of the stirrups crossing the cracks took place, first on the west side and then the east side (see Figure 5-29 to 5-32). Thus, the failure mechanism incorporated, beside bending, both out-of-plane shear in the slab, where the loading plate punched through, and in-plane shear in the girder. The subsequent failure loading of the central girder at ca 7724 s resulted in a similar failure mechanism at a total external load of 12.7 MN. In this test, the strengthening using NSM CFRP rods was functioning throughout the test. The test also showed a remarkable robustness of the bridge structure, with an ability to carry loads that were considerably higher than those that the structure had been designed for, despite the south girder having already failed (but not collapsed).

5.2 Longitudinal Strains Figures 5.5 to 5-7 show the longitudinal reinforcement strains for the upper reinforcement, close to the bridge surface, related to the time during the girder failure test, see Figure 5-3. The yield strain for Ks40 reinforcement is of the order y = y/Es ≈ 400/200000 = 0,2 % = 2000 m/m. In Figure 5-5, illustrating the north girder, SL5 and SL33 correspond to the strains over support 2 and 3, respectively, while SL18 was located at midspan. In Figure 5-6, illustrating the central girder, SL6 and SL34 correspond to the strains over support 2 and 3, respectively, while SL21 was located at midspan. Unfortunately, there were some problems with the signal from SL21.

Figure 5-5. Observed strains in the longitudinal upper reinforcement in the north girder during the girder failure test. For time scale see Figure 5-3. The southern girder failed at 6304 s and the central at 7724 s

Figure 5-6. Observed strains in the longitudinal upper reinforcement in the central girder during the girder failure test.

In Figure 5-7, illustrating the south girder, SL9 and SL37 correspond to the strains over support 2 and 3, respectively, while SL24 was located at midspan.

Figure 5-7. Observed strains in the longitudinal upper reinforcement in the south girder during the girder failure test.

Figures 5-8 to 5-10 show the longitudinal reinforcement strains for the bottom reinforcement, close to the girder bottom, during the girder failure test. In Figure 5-8, illustrating the north girder, SL3 and SL31 correspond to the strains at support 2 and 3, respectively, while SL16 was located at midspan.

Figure 5-8. Observed strains in the longitudinal bottom reinforcement in the north girder during the girder failure test.

In Figure 5-9, illustrating the central girder, SL13 and SL15 were located 2226 and 1433 mm west of the midspan and SL19 was located at midspan.

Figure 5-9. Observed strains in the longitudinal bottom reinforcement in the center girder during the girder failure test.

Figure 5-10, illustrates the strains in the south girder. Gauges SL7 and SL35 give the compressive strain at support 2 and 3, respectively, and SL22 the midspan tensile strain reaching yield early, while SL12 and SL14 give the stain between the midspan and the support, see Table 4-1.

Figure 5-10. Observed strains in the longitudinal bottom reinforcement in the south girder during the girder failure test. Gauges SL7 and SL35 give the strain at support 2 and 3 respectively, SL22 at midspan, and SL12 and SL14 in between.

5.3 Displacements Figures 5-11 to 5-16 show the displacements relating to time during the girder failure test, see Figure 5-3. In Figure 5-11, illustrating the vertical displacements along the north girder, DB1, 7 and 13 show the midspan deflections of span 1, 2 and 3, respectively. As a reminder, the loading was carried out in the midspan of span 2. Sensors DB 4 and 10 corresponds to the vertical displacement at support 2 and 3, respectively. As seen in the results, the midspan of span 2 was bending downwards, while the midspan of span 1 and 3 was bending upwards. In Figure 5-12, illustrating the vertical displacements along the central girder, DB2, 8 and 14 show the midspan deflections of span 1, 2 and 3, respectively.

Figure 5-11. Observed displacement of the north girder during the girder failure test. For time scale see Figure 5-3. The southern girder failed at 6304 s and the central at 7724 s.

Figure 5-12. Observed displacement of the central girder during the girder failure test.

In Figure 5-13, illustrating the vertical displacements along the south girder, DB3, 9 and 15 show the midspan deflections of span 1, 2 and 3, respectively. Unfortunately, the signals for DB9 was lost after about 2800 seconds as the draw wire was harmed. In Figure 5-14, illustrating the horizontal displacements at the eastern end support (roller bearing), DB16 and 17 show the longitudinal movements of the north and south girder, respectively. The results show that the bridge rotated slightly as the north side moved more than the south side.

Figure 5-13. Observed displacement of the south girder during the girder failure test.

Figure 5-14. Observed horizontal displacement at the east support during the girder failure test.

Figure 5-15 shows the horizontal displacements along of the south girder over support 3 (compared to the ground beneath the bridge). NopV was the horizontal movement in the longitudinal direction of the bridge (east-west) and NopH was the horizontal movement in the transverse direction (north-south).

Figure 5-15. Observed horizontal displacements at support 3 during the girder failure test.

Figure 5-16 shows the load displacement curves at the midspan of span 2. DB 7, 8 and 9 were located at span under the north, central and south girders, respectively. The draw wire sensor under the south girder was compromised as the wire was cut at a total load of approximately 10.5 MN.

Figure 5-16. Observed load-displacement curves for the three girders in the mid of span 2.

5.4 Column strains Six columns were monitored during the girder failure test. The results are shown in Figure 5-17 to 5-22 related to the time, see Figure 5-3. The results show a typical pattern that the side facing towards the midspan (SC2, 6, 10, 16, 20 and 24) has the highest compression and the opposite side has the lowest compression (SC4, 8, 12, 14, 18, and 22). For the columns along support 2 (west side), the strain was higher on the northern side of the column (SC1, 5 and 9) and lower on the south side (SC3, 7 and 11). The columns on support line 3 showed an opposite behaviour with higher strains on the southern side (SC 15, 19 and 23) than on the northern side (SC13, 17 and 21). In other words, there was some bending in the columns.

Figure 5-17. Observed concrete strains in column 2 under the north girder during the girder failure test. For time scale see Figure 5-3. The southern girder failed at 6304 s and the central at 7724 s. The yield strain for Ks 40 is of the order 2000 m/m

Figure 5-18. Observed concrete strains in column 2 under the central girder during the girder failure test.

Figure 5-19. Observed concrete strains in column 2 under the south girder during the girder failure test.

Figure 5-20. Observed concrete strains in column 3 under the north girder during the girder failure test.

Figure 5-21. Observed concrete strains in column 3 under the centeal girder during the girder failure test.

Figure 5-22. Observed concrete strains in column 3 under the south girder during the girder failure test.

5.5 Curvature Figure 5-23 to 5.25 show the displacement results for the curvature rigs, one on top of the slab over support 2, one on top of the slab over support 3 and one under the girder at midspan related to time, see Figure 5-3. Each curvature rig was equipped with five displacement transducers with a spacing of 800 mm. Figure 5.26 show the calculated curvature at support 2 (K1), midspan (K2) and support 3 (K3). The curvature was similar over support 2 and 3, with a slightly higher magnitude over support 2. Figure 5.27 show the load-curvature curve for the girder failure test and the maximum curvature at failure of the central girder was about 0.014 m-1 at midspan and 0.002 m-1 over the supports corresponding to radii of curvature of ca 70 and 50 m respectively.

Figure 5-23. Observed displacements for the curvature rig over support 2 of the central girder during the girder failure test. For time scale, see Figure 5-3. The southern girder failed at 6304 s and the central at 7724 s

Figure 5-24. Observed displacements for the curvature rig at midspan of the central girder during the girder failure test.

Figure 5.25. Observed displacements for the curvature rig over support 3 of the central girder during the girder failure test.

Figure 5.26. Observed curvatures at support 2 (K1), midspan (K2) and support 3 (K3) during the girder failure test. The curvature  = 0.014 1/m corresponds to a radius R = 1/ = 71,4 m

Figure 5.27. Observed load-curvature curves from the girder failure test, at support 2, K1 at support 2, K 2 at midspan and K3 at support 3. The final curvature at midspan  = 0.014 1/m corresponds to a radius R = 1/ = 71,4 m

5.6 Temperature Figure 5-28 shows the internal concrete temperature at midspan during the girder failure test. The test was carried out by the end of June, but the temperatures were very low, between 6 and 10°C.

Figure 5-28. Observed temperatures at midspan during the girder failure test. For time scale, see Figure 5-3. The influence of the temperature was studied in a MSc thesis at Lund University, Ekman - Fabricius (2016) based on methods presented by Oskar Larsson (2012). They found rather good correspondence between measrued and calculated values.

5.7 Strain in vertical reinforcement Figure 5-29 shows a picture of the south girder at failure with location of some of the vertical strain gauges (SG), compare with Figure 4-4. Several reinforcement bars have ruptures (marked with red rings). The strain gauges SG 3 – 5 (Figure 5-31) and SG 7 – 8 (Figure 5-32), installed west of the load application on the south girder, indicated yielding at load levels close to the tested peak load. The strains here generally reduced at failure since the measured area was unloaded as the adjacent part of the stirrups ruptured. Gauge SG 4 was located where one of the first diagonal cracks formed and indicated that steel yielding had already occurred during the preloading schedules (Phase 2). Thus, high strains were obtained at relatively low load levels; however, redistributions of stresses took place and this crack did not coincide with the final failure surface.

Figure 5-29 shows the location of some of the vertical strain gauges (SG) in the south girder at failure. Compare Figure 4-4. From Bagge (2017). Figure 5-30 to 5-32 show the strain in three vertical reinforcement bars in the south girder, see Figure 4-4, during the girder failure test related to the time, see Figure 5-3. The strain in the vertical bars are strongly depending of the position of cracks. Whenever a crack opened up, very close to the position of the strain gauge, there was a sudden peak for this particular gauge, whereas, as long as there were no cracks nearby, the strains remained close to zero. SV1, 3 and 6 were installed on different bars at a level 148 mm above the bottom surface of the girder. SV2, 4 and 7 were installed on different bars 548 mm above the bottom surface, SV5 and 8 were installed 948 mm over the bottom and SV 9 was installed 1348 mm over the bottom, see Figure 4-4. The first sign of a crack was detected by SV4.

Figure 5-30. Observed strains in vertical reinforcement bar 1 during the girder failure test. The southern girder failed at 6304 s and the central at 7724 s. See Figure 5-3 for time scale.

Figure 5-31. Observed strains in vertical reinforcement bar 2 during the girder failure test. The yield strain for Ks 40 is of the order 2000 m/m

Figure 5-32. Observed strains in vertical reinforcement bar 3 during the girder failure test.

The measurements using a camera for digital image correlation (DIC) were disturbed by vibrations in the rig carrying the camera (placed close to the southern girder) and by varying light conditions, see Figure 5-33, and unfortunately no useful results could be evaluated regarding the deformations in the southern girder. In earlier measurements on a bridge tested to failure in Örnsköldsvik, good results were obtained, SB-7.3 (2007), Sas (2011), Sas et al. (2012). However, tests with DIC on a metal truss bridge at Åby river were also negatively influenced by vibrations and light, Häggström (2016), Häggström et al. (2017a, b) and Elhag (2014). On the other hand, tests in the laboratory on beams and on slabs with openings, have given excellent results, see e.g. Blanksvärd (2007), Mahal (2015), Popescu (2017) and Sabau (2018).

Figure 5-33. Left: Movable rig for position of camera for digital image correlation, DIC. Right: DIC measurement area, Bagge (2017).

5.8 Strain in Near Surface Mounted Reinforcement (NSMR) and laminates Figure 5-34 to 5-35 show the strain in the NSM-strengthening during the girder failure test related to the time, see Figure 5-3. SF1-7 were installed on two NSM-bars in the region were the bars lapped each other. The region for the lapping was chosen to be located close to the theoretical section of zero bending.

Figure 5-34. Observed strains in NSMR during the girder failure test. For time scale, see Figure 5-3. The southern girder failed at 6304 s and the central at 7724 s

Figure 5-35. Observed strains in NSMR during the girder failure test.

Figure 5-36 to 5-37 show the strain in the laminate strengthening during the girder failure test. SF16-18 were installed on the three individual laminate strips at midspan. Unfortunately, these gauges only measured small strains due to the insufficient bond between the concrete and the carbon fibre. The laminates detached from the concrete after about 2200 s due to insufficient bond between the carbon fibre laminates and the concrete substrate caused by the (small) vertical curvature of the girder.

Figure 5-36. Observed strains in laminates during the girder failure test. For time scale, see Figure 5-3.

Figure 5-37. Observed strains in laminates during the girder failure test.

The results are presented in more detail in Nilimaa (2015) and Nilimaa et al. (2015 a, b).

6. SLAB TEST RESULTS The slab was subjected to two loads, east and west, with a center-to-center distance of 2.0 m. The load plates were 350 mm × 600 mm, located respectively 470 mm and 330 mm from the inner edge of the northern girder, according to Load Model 2 in Eurocode 1 SS-EN-1991-2 (2003), as shown in Figure 6-1. Draw wire sensors were installed under the slab to capture the deflection during the test. The positions of the draw wire sensors are marked with D1-D6 in Figure 6-1. The slab was loaded at a rate of approximately 80 kN/min with force control and the test setup is shown in Figure 6-2.

Figure 6-1. The arrangement of loading for the slab test (dimensions in mm).

The load – time and deflection – time responses of the test are shown in Figure 6-3. Figure 6- 3(a) shows that when the bridge was loaded to 2.6 MN, the stroke length was exceeded. At that time, the load was locked at 2.6 MN while a new grip was taken by the hydraulic cylinder. During the regrip there was a small loss of load as seen by the small reduction in deflection in Figure 6-3(b). The loading cable was reset and regripped (shown as a small shift in the load– displacement curve), Thereafter, the test continued to a failure load of 3.32 MN. Figure 6-3(b) shows the deflection history for all six draw wire sensors. An abrupt drop is seen at the point of failure for all sensors except for the one beneath loading plate D3 (directly under the plate where the failure occurred).

Figure 6-2. The loading system: (a) hydraulic jack and loading beam; (b) detail of loading beam, West loading plate and loading cable to the bed rock; (c) connection of the draw wire sensors to the ground and loading cable to the bed rock.

Figure 6-3. (a) Load–time response and (b) Displacement–time response.

Cracking, yield and failure are commonly observed indicators of the ultimate limit state of RC structures as was the case with the girders of this bridge. However, no evident cracks were observed before the sudden failure of the slab due to shear failure. The distance between the bottom surface of the slab and the inspector standing on the ground, was too long to detect any possible small hair-line cracks. The development of cracks was investigated in a nonlinear FE analysis and presented in Shu et al. (2020) and Appendix E..

Figure 6-4 shows load – displacement response captured by the draw wire sensors. A brittle shear failure occurred without any warning signs, at an ultimate load Pu = 3.32 MN. Figure 6- 4a(a) shows that the load–displacement curves for points close to the north girder are very similar and that there is much less deformation at the point on the central girder (D2) because the loading position is far away from the central girder. If the load–displacement curves of the slab for the externally applied loads related to the midspan deflection of the north girder are removed, there is much less deflection at the east loading point than at the west loading point. This is because the east point is much closer to the edge of the north girder.

Figure 6-4 (a). Load-displacement curves of the girders and slab

(b). Load–displacement curves of the slab for midspan deflection of the north girder.

The cracks in the bottom and top of the bridge deck slab after the testing are shown in Figure 6- 5. The cracks indicate that final failure was a combination of one-way shear and punching shear and the failure crack developed parallel to the girder and propagated around the west load. This formed a shear crack that reached the top surface at the edge of the loading plates on the side closest to the girder, but further towards the mid-span of the slab, as indicated by the dashed line in Figure 6-5(a). On the bottom surface, a failure crack developed until a U-shaped failure surface was formed around the west loading plate, Figure 6-5(b). Delamination of the concrete

42 surface was also observed at the bottom of the slab close to the failure surface, with flexural deformation of the reinforcement.

Figure 6-5. Bridge deck slab loaded to failure: (a) failure of the slab viewed above from the south, and (b) failure of the slab viewed from underneath.

A summary of earlier Swedish research on punching is given in Sundquist (2005). Additional results from the slab tests and their analysis are presented in Bagge (2017), Shu et al. (2018, 2020), Hallgren et al. (2017) and in MSc theses by Eriksson-Karlsson (2016) and Kobler (2016).

7. FRACTURE MECHANICS MATERIAL TESTS

7.1 General In numerical modelling of concrete bridges, the value of the concrete tensile properties and the fracture energy GF play an important role. However, mostly the fracture energy is only estimated based on the concrete compressive strength using empirical formulae. In this chapter test results are presented from the Kiruna bridge and they are compared to estimated values and results from tests on other old bridges in northern Sweden. Fracture Mechanics for metals was introduced in the early 20th century by A. A. Griffith, see e.g. Rossmanith (1991). For reinforced concrete, a committee on Fracture Mechanics was started in 1979, RILEM TC 50-FMC, Wittmann (1983). It discussed and proposed methods for analysis and test methods. It was in 1986 followed by two committees, one for work on test methods (TC 89-FMT), Shah-Carpinteri (1991), and one for applications (TC 90-FMA), Elfgren (1989), Elfgren-Shah (1991) and Elfgren et al. (2001). Anchorage and bond were identified as important questions and a Round Robin investigation was arranged on tension behavior of reinforcement bars embedded in concrete, Elfgren - Noghabai (2002). Important contributions on test methods and size effects were given by Hillerborg (1985), Bazant and Planas (1997) and Wittmann et al. (1990). Test methods for fracture energy have further been studied by e.g. Daerga (1992), Gopalaratnam - Shah (1985), Guinea et al. (1992), Noghabai (1998) and RILEM TC-50-FMC (1985). The principle for uniaxial tests is outlined in Figure 7- 1 and for three-point bending tests in Figure 7-2. Test methods for fibre reinforced concrete have been studied by e.g. RILEM TC 162-TDF (2001), Barr et al. (2003). The American Concrete Institute has also organized a committee 446 on Fracture Mechanics, see ACI 446.1R-91 (1999) and ACI 446.3R-9 (1997).

Figure 7-1. Tensile test of a concrete prism with an area A. Top left: Test with load control gives a brittle failure. Top right: Deformation control makes it possible to register the descending branch and the fracture energy GF under the load-deformation graph. Bottom left: Successive formation of crack. From Elfgren (1989).

Figure 7-2. Three-point bending test for determination of fracture energy GF = (W0 +mg0)/Alig W = External work under F-mid curve up to 0 mg = Load from beam and fixtures between supports 0 = Displacement for F = 0 Alig = Original uncracked ligament area or the projected area of the fracture surface. From Noghabai (1998), RILEM TC50 FMC (1985).

When assessing the deformations and load-carrying capacity of existing structures by finite element methods, the fracture energy of the concrete is an important material property. Examples of tests and analysis of existing structures are given in e.g. the EU Integrated Project Sustainable Bridges, Bien et al. (2007), SB-LRA (2007), with a full-scale test of a reinforced concrete bridge in Örnsköldsvik in northern Sweden in 2006, SB 7.3 (2008), Puurula (2012), Puurula et al. (2013, 2015). Fatigue of concrete in tension has been studied with fracture mechanics by e.g. Thun et al. (2011), and anchor bolts by e.g. Nilforoush et al. (2017).

7.2 Tensile tests In order to check the testing procedure, 16 new concrete cylinders were cast and tested, see Appendix B. They had the same diameter as the concrete cores from the Kiruna bridge, approximately 100 mm. Before uniaxial tensile testing, the specimen were prepared by cutting the cores to identical lengths so that they were free of pores, faults, and cracks. The tests of the samples from the Kiruna bridge are decribed below. Summeries are also given in the conference papers in Appendices C and D.

Special attention was paid to properly finish and treat the upper and lower faces of the cores. The upper and lower faces of the cores were ground and leveled so that they become parallel to each other and also perpendicular to the core’s longitudinal axes. Proper treatment of the upper and lower faces of the cores is very important to prevent generation of any bending and/or torsion actions in the core during testing. After sample preparation, a lathed notch of 3 mm thickness with a depth of 15 mm was cut at the mid-height of the concrete cores as shown in Figure 7-3.

To fix the test specimen in the test rig and apply the uniaxial tensile load uniformly on concrete cylinders, two flat steel plates are glued to the top and bottom bases of the cylinders. Before gluing, the top and bottom bases of the testing cylinders, as well as the surface of steel plates, were cleaned with isopropanol and acetone. The steel plates were glued to concrete cylinders using a two-component epoxy adhesive known commercially as “HBM X 60 epoxy” which provided a strong adhesion. This adhesive is characterized as strain-free adhesive as it has quite larger strength and stiffness than concrete. One steel plate was glued to the top base of the cylinder outside the test rig. When the adhesive has been hardened, the test sample was fastened

Figure 7-3. Direct tensile tests. Left: Loading arrangement. Top right: Dimensions of cylinders and notch. Bottom right: COD gauges and positions. to the top grip of the test rig through 6 bolts. Then the bottom steel plate was fastened to the bottom grip of the test rig and sufficient epoxy adhesive was put on the centre of the bottom steel plate. Finally, the concrete cylinder was glued to the bottom steel plate inside the test rig by applying a small compressive load on the concrete cylinder to facilitate the hardening process. Figure 7-3 and 7-4 show photos of the loading arrangement and a close-up view of the concrete cylinder mounted inside the test rig.

Figure 7-4. Left: Test set up and loading arrangement. Right: Concrete cylinder mounted inside the test rig.

To obtain the tension softening branch of material, the uniaxial loading tests were mainly performed under displacement-controlled by incrementally increasing the average of crack mouth opening displacements at a constant rate as the feedback signal. The servo-hydraulic test

46 machine was prepared by setting the accurate values of the displacement- and load-bias, the displacement limits of the test and also the load stages (displacement interval, control mode, and rate). First, the load was manually applied on the concrete cylinder from zero to 0.5 kN where the mean signal of Crack-Opening-Displacement (COD) gauges starts to change. The loading was then switched from load control mode to control through the mean signal from the COD- gauges. The displacement rate of the input signal was 0.001 mm/min up to a crack opening displacement of 0.05 mm and thereafter increased to 0.01 mm/min to speed up the tests.

The uniaxial tensile strength of concrete was calculated by dividing the maximum load obtained at test to the notch cross-section area. The fracture energy of concrete was also calculated as the area under the stress-crack width curves. The maximum crack width was measured from stress- crack width curves when the tensile stress in the softening branch reaches zero.

Table 7-1 presents a summary of the geometry of the drilled core samples prepared for the uniaxial tensile loading. In total 27 samples were prepared: 8 samples from girders, 10 samples from slabs, and 5 samples from columns (for the positions where the cores were taken see Table 7-1). In this table, for the position of the drilled cores N stands for the north side, S stands for the south side, C stands for the center. Some test results are given in Figure 7-5.

Figure 7-5. - Test results. Examples of load-deflections curves from girders (top left), slabs (top right) and columns (bottom left) and failure pattern from girder no 22 (bottom right).

Table 7-1. Summary of the drilled core samples prepared for the uniaxial tensile loading.

H d0 D Anotch Core NO. Core from [mm] [mm] [mm] [mm2] 1 Girder 100.44 103.45 70.00 3848.45 2 Girder 100.44 103.45 70.01 3849.55 3 Slab 89.64 103.83 69.99 3847.35 4 Slab 89.63 103.49 70.00 3848.45 5 Slab 90.05 103.64 70.00 3848.45 6 Slab 89.93 103.92 70.00 3848.45 7 Slab 89.99 103.67 69.99 3847.35 8 Slab 89.76 103.53 70.00 3848.45 9 Girder 89.69 104.08 70.00 3848.45 10 Girder 89.92 104.08 69.99 3847.35 11 Girder 89.85 103.42 70.00 3848.45 12 Slab 89.71 103.69 70.01 3849.55 13 Slab 89.81 103.7 70.00 3848.45 14 Column 89.85 103.6 70.00 3848.45 15 Column 89.9 103.54 70.00 3848.45 16 Column 89.51 103.44 70.09 3858.35 17 Column 89.85 103.44 69.90 3837.46 18 Girder 89.87 103.78 70.12 3861.66 19 Girder 89.84 103.92 70.10 3859.45 20 Slab 89.84 104.04 70.10 3859.45 21 Slab 89.86 103.69 70.10 3859.45 22 Girder 89.85 103.79 70.08 3857.25 23 Girder 89.83 103.79 70.13 3862.76 24 Girder 90.02 103.84 70.12 3861.66 25 Girder 89.86 104.1 70.10 3859.45 26 Girder 89.92 104.07 70.10 3859.45 27 Slab 89.89 103.67 70.10 3859.45 H is the height of concrete core sample, d0 is the initial diameter of concrete core outside the notch section, D is the mean diameter of the concrete core at the notch section, Anotch is the cross-section area of the concrete core at the notch section

The results of uniaxial tests on the drilled cores taken from the Kiruna bridge are presented in Table 7-2. This table also shows the mean values along with the standard deviation (SD) and the coefficient of variation (COV) of the measured uniaxial tensile strengths (fct) and concrete fracture energies (Gf). The uniaxial tensile strength of concrete was calculated by dividing the maximum load obtained at test to the notch cross-section area. The fracture energy of concrete was calculated as the area under the stress-crack width curves. The maximum crack width was measured from stress-crack width curves when the tensile stress in the softening branch reaches zero. As can be seen in Table 7-2, the mean tensile strength and its coefficient of variation for the concrete in columns, slabs and girders are respectively 2.75 MPa (20%), 3.70 MPa (16%), and 3.34 MPa (16%). In addition, the mean fracture energy and its coefficient of variation for the concrete in columns, slabs and girders are respectively 155.26 Nm/m2 (19%), 152.07 Nm/m2 (23%), and 166.38 Nm/m2 (43%).

Table 7-2. Result of uniaxial tensile tests on drilled cores from Kiruna Bridge.

Fmax fct Gf ωMax δ Core No. Core from max [kN] [N/mm2] [Nm/m2] [mm] [mm] 1 Girder - - - - - 2 Girder 15.73 4.09 103.70 0.254 0.261 9 Girder 14.17 3.68 329.01 0.737 0.749 10 Girder 14.85 3.86 160.67 0.3 0.308 11 Girder 10.14 2.64 153.54 0.364 0.378 18 Girder 11.87 3.09 186.36 0.38 0.388 19 Girder - - - - - 22 Girder 11.94 3.10 156.96 0.385 0.391 23 Girder - - - - - 24 Girder - - - - - 25 Girder 10.73 2.78 130.78 0.201 0.219 26 Girder 13.49 3.50 110.01 0.151 0.16 Mean 12.87 3.34 166.38 0.35 0.36 SD 2.00 0.52 71.22 0.18 0.18 CoV [%] 15.57 15.60 42.81 51.78 50.31 3 Slab 15.31 3.98 127.30 0.198 0.209 4 Slab 14.71 3.82 222.65 0.250 0.259 5 Slab 14.01 3.64 114.52 0.188 0.195 6 Slab 14.92 3.88 114.00 0.168 0.176 7 Slab 10.19 2.65 167.04 0.326 0.336 8 Slab - - - - - 12 Slab 14.59 3.79 126.48 0.431 0.441 13 Slab 12.09 3.14 149.47 0.232 0.239 20 Slab 18.05 4.68 143.83 0.163 0.171 21 Slab 16.19 4.20 171.87 0.213 0.221 27 Slab 12.35 3.20 183.57 0.281 0.289 Mean 14.24 3.70 152.07 0.24 0.25 SD 2.24 0.58 34.63 0.08 0.08 CoV [%] 15.70 15.63 22.77 33.84 32.93 14 Column 10.53 2.74 178.85 0.287 0.293 15 Column 8.85 2.30 156.44 0.232 0.237 16 Column 13.62 3.53 173.09 0.285 0.296 17 Column 9.29 2.42 112.65 0.173 0.181 Mean 10.57 2.75 155.26 0.24 0.25 SD 2.15 0.55 29.95 0.05 0.05 CoV [%] 20.37 20.16 19.29 22.07 21.61

7.3 Comparison with empirical models

In the CEB-FIP Model Code 1990 (1993) a formula for GF was given as 0.7 푓푐 퐺퐹 = ∝퐹 ( ) 푓푐0

Here fc is the concrete compression strength, fc0 =10 MPa and F = 20, 30 or 50 N/m depending on the maximum grain size dmax = 8, 16 or 32 mm respectively. Values for low compression strengths got quite

low, see e.g. Apleberger et al. (1999). In the fib Model Code 2010 (2013) the formula has been modified to: 0.18 GF = 73·fc The values from the tested concretes are compared to this formula in Table 7-3. In the table are also added some values from other old bridges in Sweden: Örnsköldsvik SB 7.3 (2008), Thun (2006), as well as some values on new concrete with an age up to a couple of months, Noghabai (1998), Gabrielsson (1989) and Thun (2006).

Table 7-3. Tested and modelled values of Fracture Energy, GF [N/m] for old and new concrete. Modeled values refer to fib Model Code 2010 (2013). Values in parenthesis refer to covariances

References to fc ft GF,Test GF,MC GF,Test /GF,MC Old Concrete MPa MPa N/m N/m - This report Column 61.8 (11%) 2.75 (29%) 155.3 (19%) 153.4 1.01 Girder 56.4 (19%) 3.34 (16%) 166 (43%) 150.9 1.10 Slab 71.6 (3%) 3.7 (16%) 152.1 (23%) 157.5 0.97 SB Övik (2008) 68.5 (8%) 2.2 (23%) 154 (53%) 156.2 0.99 Thun (2006) Old 72.6 (6%) 3 (12%) 159.9 (32%) 157.9 1.01

References to fc ft GF,Test GF,MC GF,Test/GF,MC New Concrete MPa MPa N/m N/m - Appendix in H100/10 66.3 (3%) 5,55 (4%) 140.6 153.4 0.91 this report H100/15 130.7 (21%) 0.84 H200/10 128.3 (19) 0.83 H200/15 137 (20%) 0.88 Thun (2006) New 72.2 (3%) 3 (7%) 116.5 (15%) 157.7 0.74 Gabrielsson C15A 130 (2%) 4.4 (5%) 207 (34%) 175.3 1.18 (1999) C21A 185 (2%) 6.07 (16%) 237 (19%) 186.8 1.27 Noghabai 4 114.2 (2%) 5.1 (5%) 217 (6%) 171.3 1.27 (1991) 7 121.6 (3%) 5.01(14%) 216 (28%) 173.2 1.25 15 113.4 (3%) 4.15 (16%) 159 (25%) 171.2 0.93 20 98.7 (2%) 3.77 (5%) 127 (20%) 168.8 0.76 27 65.1 (4%) 4.48 (13%) 170 (9%) 154.8 1.10

The ratio between the tested values to the values computed based on the compressive strength get the values 1.1, 0.97 and 1.01 for the three tested concretes from the Kiruna bridge. Also, for the two other old bridges the ratio is close to 1. This is rather good, so for the tested bridges, the simple formula in the Model Code gives a good prediction of the fracture energy GF. For the new concrete the variation is larger and for high strength concretes it seems mostly to be conservative. An exponential formula for the stress-deformation -w relationship has been proposed by Hillerborg (1989) It gives quite a good description of the deformation processes in Figure 7-3:

−3 f  =+fw10.5 t t  G f Guidelines for modelling are also given in Dutch recommendations, Rijkswaterstaat (2017). The influence of the value of the fracture energy GF in assessments is discussed in Bagge et al. (2017). They found that the boundary conditions and the value of the fracture energy are the most important parameters when evaluating the capacity of a structure.

7.4 Discussion. Estimated and Measured Concrete Properties

For concretes of ages 30 – 50 years, the tested samples give results which quite well follow the simple 0.18 formula for fracture energy GF = 73·fc in the fib Model Code 2010 (2013). However, the variation in results is considerable with covariances from 19 to 53%. For new concretes of ages up to a few months, the variations relative to the formula are larger, especially for high strength concretes with compression strengths fc > 80 MPa. Further evaluations and tests are necessary to give firm recommendations. All results are based on uniaxial tests. Tests of beams in three-point bending are more common, as this is a simple test procedure when you can cast the beams at the same time as the casting of the structure. However, it is much more complicated to saw out beams from an existing structure, so here drilled out cores are preferable. The estimated and measured values of concrete uniaxial tensile and compressive strength, elastic modulus, and fracture energy for the concrete from the Kiruna Bridge are summarized in Table 7-4. Table 7-4 – Measured and estimated values of concrete properties

Estimated values Measured values

Structural Concrete fcm,es Ecm,es fctm,es Gf,es fcm Ecm fctm Gf element class [MPa] [GPa] [MPa] [N/m] [MPa] [GPa] [MPa] [N/m] Column K300 30 30.6 2.4 134.6 61.8 32.4 2.7 155.3 Girder K400 40 33.3 3.0 141.8 56.4 31.2 3.3 166.4 Slab K400 30 30.6 2.4 134.6 71.6 33.3 3.7 152.1

8. ASSESSMENT METHODS

8.1 General Structural analysis has a crucial role in bridge assessments. To support rational improvements of such analyses, a multi-level strategy has been developed, Bagge (2017, 2020), Bagge et al. (2019b). By gradually increasing the efforts, the structural response and the load-carrying capacity can be more accurately estimated. This methodology can preferably be carried out in three steps: (1) Initial, (2) Intermediate and (3) Enhanced, see Figure 8-1. The method is an extension of a strategy developed for bridges in Schneider (1994), SB-ICA(2007), SB-LRA (2007), Plos et al. (2007), UIC (2009), Mainline (2014), fib Model Code 2010 (2013), ISO 2394 (2015), Paulsson et al (2016) and a procedure for deck slabs by Plos et al. (2016). Various degrees of degradation and/or fatigue must also be taken into consideration, see e.g. Elfgren (2015).

Figure 8-1. Assessment procedure in three phases: (1) Initial, (2) Intermediate and (3) Enhanced. Based on Sustainable Bridges (2007), Mainline (2014), UIC (2009), ISO 2394 (2015), and Paulsson et al (2016). Original concept from Schneider (1994). In order to enable a better representation of the structural behaviour in the intermediate Phase 2 of the assessment, three sublevels have been defined: A - Linear elastic analysis B - Linear elastic analysis with limited redistribution C - Plastic analysis These methods utilize well-established procedures. For instance, methods for linear elastic analysis with limited redistribution are provided by design codes as e.g. the fib Model Code 2010 (2013) and for plastic analysis by Hillerborg (1975, 1996) and Nielsen & Hoang (2016).

In Phase 3 of the assessment, Finite Element (FE) analysis may be carried out with gradually increased accuracy - Nonlinear analysis reflecting flexural failure behaviour - Nonlinear analysis reflecting flexural and shear failure behaviour Nonlinear analysis reflecting flexural, shear and bond failures (including reinforcement slip). An example of this is the analysis of the bond failure of the Carbon Fibre Reinforced Polymer (CFRP) bars used to strengthen the Örnsköldsvik Bridge, see SB 7.3(2008), Puurula (2012), Puurula et al (2013, 2015, 2016). An attempt to include the influence of corrosion and cracks on anchorage is given by Blomfors (2020). You may also increase the accuracy by going from partial safety coefficients to global safety coefficients and finally to fully probabilistic methods, Bagge (2017, 2020). In all steps the costs for the assessment procedure must be weighed against the profits of a longer use of an existing bridge or the costs to upgrade/strengthen/repair it or to destruct and exchange it to a new one. Theoretical studies in this area are being done by e.g. Leander et al. (2018) and Bertola et al. (2020). They use the term Value of Information (VoI) for the difference between the expected cost (risk) of strengthening/repair/renewal and the expected cost for assessment/monitoring. This Value of Information can be substantial for old structures or for structures where you want to increase the capacity, while it can be small (or negative) for extensive assessment/monitoring of new structures. In practise, it is often difficult to accurately estimate these costs as it postulates both a deep knowledge of the function of the structure as well as extensive experiences of cost estimations for special works. Nevertheless, essential sums can be saved for the society by striving for this. An example of unnecessary costs for society is the failure of the 50-year-old Polcevera (Morandi) viaduct in Genoa on August 14, 2018, with 43 casualties and a two-year traffic stop. Concrete encased cables broke due to corrosion and fatigue caused by cracks in the concrete supposed to protect them. The commercial company responsible for maintenance was aware of the corrosion problems but underestimated the risk and did not take warnings from observations and monitoring seriously enough, Calvi et al. (2018), Domaneschi et al. (2020) and Morgese et al. (2020). Overviews of Structural Health Monitoring Systems (SHMS) are given in e.g. SB-MON (2007), I2R D4.2 (2017), Täljsten et al (2019, 2020) and COST TU 1406 (2019). Examples of interesting cost-effective results by reduced, well-planned monitoring (“Pocket-Monitoring”) are given by e.g. Brühwiler (2018) and Sawicki & Brühwiler. (2020). Methods and results from load testing used to assess structures are presented in e.g. Häggström (2016), Häggström et al. (2017a, b), Bagge et al. (2018a). Landsoght (2019) and Elfgren et al. (2018, 2019). Upgrading, strengthening and repair of bridges are treated in e.g. SB-STR (2008), Mainline D1.1 & D1.2 (2013), Zilch et al. (2014), Nilimaa (2015), Nilimaa et al. (2015c), Täljsten et al, (2016), BASt (2016) and C4R (2017). Below some results from the assessment of the Kiruna bridge are summarized and compared to the test results followed by an outline of what ought to be done in the future.

8.2 Prestressing forces For the prestress losses, a comparison between theoretically assessed and measured results are given in Table 8-1, Bagge (2017), Bagge et al (2017). Estimated losses generally ranged between some 5 to 70 %. The measured ones varied even more. However, the values in parentheses for section A, referring to an updated position of the prestressing strands (125 mm higher than in the drawings), give more reasonable values. Thus, the differences between theoretically estimated and measured prestressing losses vary from 0 % to 35%. More studies are obviously needed to give better guidance. Table 8.1 Theoretically and experimentally determined prestress losses, Bagge (2017), Bagge et al (2017). Section A refers to midspan 1, section B to midspan 2 and section C to midspan 3. N refers to the North girder, C to the central and S to the south girder. Values in parentheses refer to updated positions of prestressing strands.

8.3 Flexural-Shear Failures of Girders For the tested girders, a load-deflection graph is given in Figure 8-2. It also shows the range for assessed failures according to Eurocode 2, SS-EN 1992-2 (2003) and the fib Model Code 2010 (2013), Bagge (2017, 2020), see further Figure 8-3, Bagge (2017). The lowest code assessments are very conservative giving one fourth of the real shear capacity (3.4/13.4 = 0.25), while more advanced methods still only give up to 8.8/13.4 = 66 % (shear) of the test result using original material properties and up to 9.1/13.4 = 68 % (moment) using tested material properties. As a comparison, a rough basic hand calculation has also been done, see Appendix F. The calculated bending failure load for the girders varied between PM = 7.4 and 10,3 MN depending on assumptions and can be compared to the tested value of Ptest =13.4 MN. For shear the calculated value varied between PVs = 11.6 and 12.6 MN depending on the assumptions. The modelling of the bridge has also been studied in MSc theses by Björklund & Nilsson (2014) and Sawicki (2015).

Figure 8-2. Load-deflection curves and preliminary assessed capacity according to Eurocode 2 and fib Model Code 2010. Based on Bagge (2017, 2020).

Figure 8-3. Load-carrying capacity regarding moment and shear using different local code resistance models for structural assessment, Bagge (2017).

The first finite element (FE) models showed good agreement with the test for loads up to approximately 9 MN. The models were, however, based on characteristic material properties, with no bond models for the laminates and with assumed levels of post-tensioning force. Before the failure test, the bridge girders were already loaded to concrete cracking adjacent to the load application. This process was not considered in the FE analyses so a difference in stiffness occurred. Moreover, the FE analysis was based on equally loaded beams. This confirms that even with a limited amount of information a good prediction of the behaviour of a bridge can be obtained. Wirth more advanced and calibrated non-linear FE methods very good correlation was obtained, see below. An assessment in the Ultimate Limit State (ULS) for loads from classification vehicles according to TDOK 2013:0267 (2017) has also been done, Bagge (2017, 2020). Figure 8-4 shows a critical vehicle and Figure 8-5 gives resistances and actions.

Figure 8-4, Configuration (l) in Group B for classification vehicles for assessment of highway bridges, TDOK 2013:0267 (2017). This Group B has a total load of 6,6B from 14 axles with axle loads varying from 0,33B to 0,55B. Group A consists of one single axle load of A kN. Bagge (2017, 2020).

Regarding the flexural capacities of the girders, the regions adjacent to the supports were shown to have a critical sagging moment adjacent to the supports (see Figure 8-5a). This is associated with relatively low flexural resistances in sections where the prestressing reinforcement is located close to the center of gravity. Moreover, curtailment of the non-prestressed reinforcement at the base of the girders contributes to the low flexural resistance. The curtailment can be a lack in the reinforcement design and is probably the cause of cracks observed to be formed at the intermediate supports. The north girder is critical with the most adverse section being located west of Support 4 (A = 243 kN and B = 183 kN with tested material properties). For each girder, the shear capacity consistently limited the permitted axle load. The section identified being critical was located at Support 5 (see Figure 8-5b) with A = 166 kN and B = 94 kN, using tested material properties. Following the truss analogy, these values are based on verification of shear force not closer than a distance equal to the effective depth from the edge of the support, and thus higher shear forces are to be allowed at the support (see support region in Figure 8-5b). For the bridge deck slab, the flexural capacity was limiting. The critical sections were identified in the girder connection of the intermediate slab (A = 318 kN and B = 533 kN) and the girder connection of the cantilever slab (A = 329 kN and B = 419 kN). However, the deck slab was shown to have an essential overcapacity in terms of shear and punching.

Figure 8-5. Resistances based on in situ tested material properties and action effects for traffic loads, associated with the classification vehicle of Groups A and B, for the south girder: (a) flexure moment with critical section adjacent to Support 4, and (b) shear force with critical section adjacent to Support 5. Bagge (2020)

Figure 8-6 gives load-deformation graphs calculated with non-linear finite element methods for the cantilever slab at Support 5, which was the most critical for a shear failure. In Figure 8-6a the values correspond to the classification of a Group A vehicle composed of a single axle with the load A and in Figure 8-6b the values correspond to a Group B vehicle with the bogie load B for the most adverse of the classification vehicles (l), composed of multiple axles (Figure 8-4). The methods at the initial assessment are highly conservative regarding the shear capacity of the girders. This can be explained by the range of simplifications used in both the structural analysis and the models to determine the local resistances. In the structural analysis, a linear behaviour was assumed, although considerable redistribution of forces can be expected. Regarding the shear resistance model, for instance, less than 70% of the tensile strength of the stirrups was used since the failure was defined at the start of yielding. Moreover, the contribution of the flange of the T-shaped girder (i.e., the bridge deck slab) and the positive impact of loading close to supports were disregarded. Assuming a similar outcome from enhanced analyses of the other girders for the critical sections and load cases, the shear capacity can be concluded as not being critical for the bridge superstructure. However, the most adverse load cases for the flexural capacities of the girders have not been included in the enhanced assessment above and have to be further investigated. Eventually, the slab would also need to be further analysed, depending on permitted load level due to flexure.

Figure 8-6. Load–deflection behaviour of the cantilever slab at Support 5 (most critical for shear failure) given by nonlinear finite element (FE) analysis using average, characteristic and design material properties: (a) classification vehicle Group A with an initial assessed axle load of A = 166 kN and (b) classification vehicle Group B with an initial assessed bogie load of B = 94 kN (corresponding to a total load of 6,6B = 620,4 kN and an average axle load of B/2 =47kN), compare Figure 8-3. Bagge (2017, 2020).

A comparison of the initial and enhanced assessments thus indicated a large potential to increase the permitted axle loads by improving the structural analysis. In Figure 8-7, the ratio of the shear capacity of the initial and enhanced levels of analyses has been shown for different safety concepts, Bagge (2020). The methods for the initial assessment are highly conservative with regard to the shear capacity of the girders. This can be explained by the range of simplifications used in both the structural analysis and the models to determine the local resistances.

Figure 8-7. Ratio of the permitted axle loads predicted at the enhanced (Renh.) and initial (Rinit. = 94 kN) levels of structural analysis. Methods used: PSF = Partial Safety Factor, GRSF = Global Resistance Safety Factor, ECOV = Estimation of Coefficient of Variation, Imp ECOV = Improved ECOV, Probabilistic = Full Probabilistic Method. Bagge (2020).

A comparison of initial (B = 94 kN) and enhanced full probabilistic (B = 14.1∙94 =1325 kN) assessments indicates the large potential to increase the permitted axle loads by improving the structural analysis.

8.4 Punching Failure of Slab Assessments of the punching shear failure of the slab have been carried out by Bagge (2017) and Shu et al (2018, 2020), see Figure 8-8. A rough hand calculation in appendix F gave PV ≈ 2.84 MN as compared to test result Ptest = 3.32 MN. There is a considerable difference between the assessment values derived with standard design codes and the initial rough estimate and the advanced assessments. One reason for the low code values is the no-warning, brittle, failure, which motivates conservative methods

Figure 8-8. Load-deflection graph for punching test of slab with assessment with codes- The results are further discussed in Appendix E and Shu et al. (2020). The modelling of the slab has also been studied by Hallgren et al. (2017).

8.5 Damage Detection and Life-Cycle Assessment Corrosion of post-tensioned tendons in bridges is a problem in the assessment and maintenance of prestressed concrete bridges. In connection with the testing of the Kiruna Bridge this phenomenon was studied and almost no sign of corrosion could be found, see Figure 8-9.

Figure 8-9. Photograph from inspection of post-tensioned tendons.

The tests can be compared to an earlier study of a 50-year old prestressed bridge in Örnsköldsvik. There, tests on six prestressing strands showed that many of the strands had started to corrode. Their ultimate stress was not affected but the corroded strands had a more brittle failure and only half the failure strain of not corroded strands. The ordinary reinforcement was free from corrosion, SB7.3 (2008). Methods to detect possible damages in the bridge by using deformation and acceleration measurements have been studied in Enochsson et al. (2011), Grip (2013), Huang et al. (2016)

and Grip et al (2017). This has also resulted in papers on acceleration monitoring and assessment by Grip & Sabourova (2011), Forsberg et al. (2013), Sabourova et al (2012, 2019a, b, 2020), Bagge et al. (2015c) and Duvnjak et al. (2018, 2019). In Bagge, Nilimaa, Sundquist et al. (2016) some reflections are given on the life cycle of the Kiruna bridge. Since the bridge was privately owned, there is no written information available about inspections and repair actions performed during the bridge life span. The observations made during the full-scale test made it likely that no repair actions have been performed during the 55 years in service. There was almost no corrosion found on the post-tension strands. A Life Cycle Cost Analysis (LCCA) performed using the methodology presented in ETSI (2013) and using input data from Safi (2013), would have shown that repairs should have been needed and thus giving costs in an LCCA. The time intervals between maintenance and repair actions presented in Safi (2013) are based on historical data for Swedish bridges evaluated from the database BaTMan (Bridge and Tunnel Management, maintained by Trafikverket). The bridge was designed according to the Swedish concrete standards from 1934 and 1949, SOU (1949), giving fewer, conservative rules for concrete cover and no special precautions due to the risk for frost action in the concrete, compared to the standards of to-day.

How come that the bridge survived better than the average Swedish concrete bridges? There are at least three explanations all due to the local climate at the bridge site far up in northern Sweden: - No de-icing salt is used due to the cold climate; - The number of frost cycles is much lover in this part of Sweden, Krus (1996); - The average humidity is lower at the bridge site than in central Sweden.

So, what is to be learnt? A high quality LCCA should take the local conditions into account; in this case giving lower LCC. It might also be relevant, in this climate, to apply less conservative rules for environmental degradation than stated in the standards, thus saving money and reducing global emissions, see Du (2015).

8.6 Final Demolition of bridge

The easiest way to demolish a bridge is usually to blast it into pieces. If environmental conditions prevent this, you must use drilling, sawing and crushing of the concrete and cutting of the steel with e.g. oxygen or water jet. If there is enough space you may use a launching truss and successively demount the bridge on to it, see e.g. Franz & Ansorge (2018), Schacht et al. (2018) and Wagner (2017)

When the Kiruna bridge finally was to be demolished, all fixed supports were first weakened by drilling and crushing of the concrete. Then the bridge spans were pushed down in sections, see Figure 8-10. This was quite a spectacular show.

Figure 8-10. Final demolition of the Kiruna bridge after weakening of joints

8.7 Outlook Some lessons learnt from a state-of-art study of load testing to failure are presented in Bagge et al. (2018). About 28% of full-scale tests on 30 bridges ended with a failure mode different to that predicted. In some cases, this was related to inaccuracies in the methods for determining the load-carrying capacity but, in the majority of cases, it was caused by a lack of insight into aspects shown to be critical, particularly associated with the shear and punching capacities and the boundary conditions. Consequently, there is a need of further studies in order to provide reliable codes and guidelines on how to accurately assess the capacity of bridges. Many tested structures had a considerable “hidden” capacity which can be disregarded during ordinary assessment processes and which is accounted for neither in standards nor in design guidelines. One reason is the high safety factors that are used both for loads and materials in the construction phase and which may not be necessary in an assessment process where geometry, materials and load may be better known, Paulsson et al. (2016), fib TG 3.2 (2020). Probabilistic methods can be applied successfully to improve the study of reliability and safety of existing structures. More experience and acceptance of reduced reliability factors for different existing structures are here needed, see e.g. Bagge (2020). Fatigue is an important factor when the load on a structure is increased. The rate of damage when the stresses are increased rises with a logarithmic factor that can be three to five times the stress increase. This means than an increase of stress range may cause a proportionally much larger reduction in the number of allowable load cycles. Methods to determine the remaining fatigue capacity would be very valuable both for steel and concrete bridges. The concrete fatigue capacity in shear is not as critical as many codes envision (Elfgren 2015). Shear stresses are in the design phase often converted to tensile and compressive stresses and the tensile stresses are mostly carried with reinforcement or eliminated by prestressing. Furthermore, concrete in compression seldom gives any fatigue problems. Recent examples of fatigue monitoring and modelling in a concrete arch bridge in Sweden is presented by Wang et al (2019) and in a prestressed box-girder bridge in Germany by Hansen (2020). Bridge managers need to navigate cautiously between taking too large risks, which may cause catastrophic failures, or playing it too safe by demolishing and exchanging fully functional

bridges just because they do not fulfil all codes for new structures. A careful, wise, assessment may here save several millions of Euros by prolonging the life length of a bridge. The results acquired during the investigations of the Kiruna Bridge suggest that the following aspects warrant further attention: - Condition assessment of post-tensioned steel tendons including influence of fatigue and corrosion - Shear force resistance of full-size bridge girders and shear force and punching resistance of bridge slabs - Temperature effects on deformations and strains and influence of varying climate conditions - Robustness, ductility, strengthening methods and possible recirculation of parts and materials after end of use for different types of bridges - Calibration of and recommendations for finite element model updating and reliability-based analysis

Society may learn and save money from the experiences from “full-scale” failure tests. They can act as a complement to the experiences from unwanted and unexpected failures due to increased loads, scour, corrosion and other form of deterioration. It is therefore recommended that additional tests are to be carried out in order to further improve the understanding of existing bridges. The tests should as far as possible be based on realistic load cases, in order to optimize the outcome. Different bridge types can be tested to check their real capacity and give a background for establishing and calibrating numerical models of them.

9. CONCLUSIONS Closure of the Kiruna Bridge provided a rare opportunity for LTU and collaborating colleagues to monitor a post-tensioned concrete bridge during tests to failure using a wide array of instruments. The primary aim was to acquire relevant data for calibration and development of methods for assessing prestressed and post-tensioned concrete structures. The test programme included evaluations of the super-structure’s behaviour, two CFRP strengthening systems and conditions of the post-tensioned tendons. Thus, the results provide abundant information for refining existing methods and models, which was a key aim for this research project In the tests of the Kiruna bridge it was found that: • The empirical formulae for concrete fracture energy, GF, according to the fib Model 0.18 Code 2010, GF = 73 fcm , gave a good estimate for the tested concrete with GF ≈ 150 2 Nm/m . The mean values of the tensile strength fct were 2.7, 3.3 and 3.7 MPa for the columns, girders and slabs respectively and the mean values of the compressive strength fcc were 61.8, 56.4 and 71.6 MPa for the columns (K300), girders (K400) and slabs (K400) respectively. • When the remaining prestressing forces were calculated by standard methods and estimated from measurements, the differences of the losses according to the two methods varied from 0 % to 35% of the original prestressing forces. • The strengthening procedure worked well with Near Surface Mounted (NSM) Carbon Fibre Reinforced Polymer (CFRP) bars. For the prestressed laminates, debonding occurred at 65% of the failure load due to a curved bridge surface. If precautions had been taken to eliminate the curvature, a higher capacity had probably been the result. • The girders failed in bending - shear for a load of 13,4 MN, while code methods gave loads between 3,4 MN and 9,1 MN. The slab failed in a punching failure for a load of 3,3 MN while code methods only gave ≈ 1,5 MN. When the bridge was assessed for standard code vehicles, the load-carrying capacity could be increased up to 14 times going from standard code methods to more advanced probabilistic methods. • Generally, the bridge had a high hidden load-carrying capacity which could not be predicted with standard code methods. Assessment methods using nonlinear finite element models (FEM) gave very accurate results after calibration of actual material properties and boundary conditions.

The report presents the experiences from a tested bridge in Sweden. The bridge had a considerably higher capacity than the original design calculations indicated. The tests to failure showed differences, e.g. in material properties, load distribution and support conditions, which resulted in a considerable extra “hidden” capacity of the bridge. Society may learn and save money from the experiences from “full-scale” failure tests. They can act as a complement to the experiences from unwanted and unexpected failures due to increased loads, scour, corrosion, fatigue and other forms of deterioration. It is therefore recommended that additional tests are to be carried out in order to further improve the understanding of existing bridges. The tests should as far as possible be based on realistic load cases, in order to optimize the outcome. As the tests are costly it is important that planning, preparations and analysis are done in a careful way. Different bridge types can be tested to check their real capacity to give a base for establishing and calibrating numerical models to be used for reliable assessment of existing bridges in order to improve quality control, a cost-efficient bridge management and a sustainable usage of the existing bridge stock.

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