Field Monitoring of Shrinkage Cracking Potential in a High

FIELD MONITORING OF SHRINKAGE CRACKING

POTENTIAL IN A HIGH-PERFORMANCE CONCRETE

BRIDGE DECK

TIM WALKOWICH

A Thesis submitted to the

Graduate School – New Brunswick

Rutgers, The State University of New Jersey

in partial fulfillment of the requirements

for the degree of

Master of Science

Graduate Program in Civil and Environmental Engineering

Written under the direction of

Dr. Hani Nassif

and approved by

New Brunswick, New Jersey

January 2011 ABSTRACT OF THE THESIS

Field Monitoring of Shrinkage Cracking Potential in a High-

Performance Concrete Bridge Deck

Thesis Director:

Dr. Hani H. Nassif

Over the past decade many state engineers throughout New Jersey have reported cracking on High Performance Concrete (HPC) bridge decks at early ages. The presence of cracking early in the life of a high performance deck offsets the benefits gained in using the material as the potential for corrosion begins at the onset of cracking. While many factors apply to bridge deck cracking, the shrinkage of the concrete’s mass is a primary concern. Because of shear studs and boundary conditions, among other causes that act in restraining the deck itself, it is important to understand the mechanics of concrete under restraint.

The AASHTO Passive Ring Test (PP 34-06) is seeing an increase in use in studies analyzing restrained shrinkage. The test simulates a concrete member of infinite length and allows researchers to study the effects of various parameters on restrained shrinkage. This thesis presents the results of a study that analyzed the ring test’s ability to simulate restrained shrinkage on HPC bridge decks. The investigation incorporated an instrumented, simply supported

ii composite bridge deck with laboratory samples taken on the day of the pour as well as a finite element analysis. The results suggest the AASHTO Passive Ring Test simulates the restrained shrinkage of simply supported HPC decks reasonably well. Fewer than 1% of all cracking present on the ring specimens saw complete penetration through the sample with 80-90% of all cracking considered to be micro cracking. While the presence of several cracks along the bridge deck itself showed no correlation with the shrinkage ring specimens, finite element analysis suggests these cracks are a result of adjacent live load. Also, the findings of this study highlight the importance of following design in the field as well as the effect of live load on staged construction of HPC bridge decks.

iii

ACKNOWLEDGMENTS

I would like to thank Dr. Hani H. Nassif for all the opportunities I have been fortunate enough to take as well as his support throughout my time at Rutgers. The knowledge and experience I have gained through my relationship with him has been and I am sure will continue to be invaluable.

I would also like to thank Dr. Husam S. Najm and Dr. Kaan Ozbay for being on my committee and providing their insight.

I would like to thank my father and mother, Anthony and Mary Ellen Walkowich for helping me to grow into the person I am today. Without their guidance I would not have achieved half of what I have at this point in my life. Special thanks to my sisters, Heather and

Jessica, without whom I would not have the laughter and memories that keep me going when times are rough.

I am forever in debt to Carl Fleurimond, Dan Su, Etkin Kara, Ufuk Ates and Gunup

Kwan. Their friendship and guidance were critical in my success at Rutgers and I wish them all nothing but the best in their future efforts.

Without the help and participation of both the NJ Turnpike Authority and the SHAW

Group, Inc., this thesis would not be possible. Thank you in particular to Adel, Scott, and Paul for allowing me on site whenever my research required.

Thank you to Mike, Chris, Alex, Parth, John, Peng and everyone and anyone that ever helped in the lab in any way big or small. The assistance you provided made my experience at

Rutgers possible.

Last, I would like to thank my closest friends Artie, Eric, Chris, Bryan, Matt and Jake.

Through the good times and the bad you have all stuck with me and I cannot put into words what that has meant to me.

TABLE OF CONTENTS

ABSTRACT OF THE THESIS ii

ACKNOWLEDGMENTS vi

CHAPTER I – INTRODUCTION 1

1.1. PROBLEM STATEMENT 1 1.2. RESEARCH OBJECTIVES AND SCOPE 2 1.3. THESIS ORGANIZATION 2

CHAPTER II – LITERATURE REVIEW 4

2.1. INTRODUCTION 4 2.2. TYPES OF SHRINKAGE 4 2.2.1. PLASTIC SHRINKAGE 5 2.2.2. THERMAL SHRINKAGE 5 2.2.3. AUTOGENOUS SHRINKAGE 6 2.2.4. DRYING SHRINKAGE 6 2.3. SHRINKAGE FACTORS 7 2.4. RESTRAINED SHRINKAGE RING TEST 8 2.4.1. RING TEST BACKGROUND 8 2.4.2. RING TEST SETUP 11 2.5. PREVIOUS WORK 12

CHAPTER III – EXPERIMENTAL SETUP 35

3.1. INTRODUCTION 35 3.2. MATERIAL PROPERTIES OF MIX 36 3.3. MIXING AND FRESH SAMPLING OF CONCRETE 37 3.3.1. SLUMP TEST 38 3.3.2. AIR CONTENT 39 3.3.3. SAMPLING OF SPECIMENS AND CONSOLIDATION 40 3.3.4. CURING 41 3.4. INSTRUMENT DETAILS AND FIELD IMPLEMENTATION 42 3.4.1. EMBEDDED VIBRATING WIRE STRAIN GAUGES 42 3.4.2. PORTABLE DATA LOGGER 45 3.4.3. ACCELEROMETERS 46 3.4.4. LASER DOPPLER VIBROMETER 46 3.4.5. STRUCTURAL TESTING SYSTEM 47 3.5. LABORATORY TESTING PROCEDURES 49 3.5.1. COMPRESSIVE STRENGTH OF CONCRETE SPECIMENS 49 3.5.2. SPLITTING TENSILE STRENGTH OF CONCRETE SPECIMENS 50 3.5.3. MODULUS OF ELASTICITY 51 3.5.4. FREE SHRINKAGE TEST 52 3.5.5. RESTRAINED SHRINKAGE TEST 53 3.5.5.1. ENVIRONMENTAL CHAMBER 54

CHAPTER IV – TEST RESULTS 56

4.1. INTRODUCTION 56 4.2. MECHANICAL PROPERTIES 56 4.2.1. COMPRESSIVE STRENGTH 56 4.2.2. SPLITTING TENSILE STRENGTH 58 4.2.3. ELASTIC MODULUS 59 4.2.4. FREE SHRINKAGE 60 4.3. LABORATORY TEST RESULTS 61 4.3.1. SHRINKAGE RINGS 61 4.4. FIELD TEST RESULTS 68 4.4.1. FIELD STRAINS 68 4.4.2. BRIDGE DECK CRACKING 76

CHAPTER V – FINITE ELEMENT MODELING 82

5.1. INTRODUCTION 82 5.1.1. MODEL ELEMENT TYPES 82 5.1.1.1. BEAM ELEMENT 83 5.1.1.2. SHELL ELEMENT 83 5.1.1.3. STEEL REINFORCEMENT 84 5.1.1.4. SHEAR STUDS 84 5.1.1.5. BOUNDARY CONDITIONS 84 5.1.1.6. CONSTRAINT AND RELEASE ELEMENTS 85 5.1.2. MATERIAL PROPERTIES 85 5.2. FINITE ELEMENT ANALYSIS RESULTS 87 5.2.1. BRIDGE DECK ANALYSIS 87 5.2.2. FINITE ELEMENT CONCLUSION 102

CHAPTER VI – SUMMARY AND CONCLUSIONS 103

6.1. SUMMARY AND CONCLUSIONS 103 6.2. FUTURE SCOPE OF WORK 104

REFERENCES 105

vii

LIST OF TABLES

Table 2.5-1. Phase I and II Mix Proportions (Whiting, et. al. 2000) 17 Table 2.5-2. Time to First Crack for FDD Mixes (Whiting, et. al. 2000) 20 Table 2.5-3. Time to First Crack for TDO Mixes (Whiting, et. al. 2000) 20 Table 2.5-4. Time to First Crack for Phase II Mixes (Whiting, et. al. 2000) 21 Table 2.5-5. North Eastern Ohio Field Survey Results (Delatte, et. al.) 23 Table 2.5-6. Absorptive Light Weight Aggregate Mix Proportions (Delatte, et. al.) 23 Table 2.5-7. Cracking Dates for Ring Specimens (Delatte, et. al.) 24 Table 2.5-8. Time to Cracking of Ring, RE and RBE Test Samples (Weiss, et. al.) 26 Table 3.2-1. NJ Turnpike Mix Design 37 Table 3.2-2. NJ Turnpike Admixture Content 37 Table 3.4-1. VWSG Orientations and Locations 44 Table 3.5-1. Summary of Laboratory Tests Performed 49 Table 4.2-1. Compressive Strength of Concrete Mix over Time 57 Table 4.2-2. Compressive Strength of Concrete Mix over Time 58 Table 4.2-3. Compressive Strength of Concrete Mix over Time 60 Table 4.2-4. Free Shrinkage of Concrete Mix over Time 61 Table 4.3-1. Crack Width Distribution Over All Samples 67 Table 4.4-1. Crack Map Details 77 Table 5.2-1. Strain Validation 88 Table 5.2-2. Concrete Properties for New Concrete Sections 89 Table 5.2-3. FE Model Strains Resulting from Case 1 Loading 91 Table 5.2-4. FE Model Strains Resulting from Case 2 Loading 93 Table 5.2-5. FE Model Strains Resulting from Case 3 Loading 95 Table 5.2-6. FE Model Strains from Case 4 Loading 96 Table 5.2-7. FE Model Strains from Case 4 Loading 97 Table 5.2-8. FE Model Strains from Case 6 Loading 99 Table 5.2-9. FE Model Strains from Case 7 Loading 100 Table 5.2-9. FE Model Strains from Case 7 Loading 101

viii

LIST OF FIGURES

Figure 2.4.1. AASHTO Ring Test Geometry 11 Figure 2.5.1. Shah, Ouyang, et. al. Element Mesh 14 Figure 2.5.2. Tensile Strength Prediction (A) and Experimental (B) (Shah, Ouyang, et. al.) 16 Figure 2.5.3. Shrinkage of HPC mixes with varying w/c and silica fume content (Whiting, et. al) 18 Figure 2.5.4. Shrinkage of HPC mixes with varying geometry and w/c (Whiting, et. al) 19 Figure 2.5.5. RE and RBE Test Method Setup (Weiss, et. al.) 25 Figure 2.5.6. Typical Cracking Observed on NJ Bridge Decks (Saadeghvaziri and Hadidi) 27 Figure 2.5.7. Bridge Survey Distribution throughout NJ (Saadeghvaziri and Hadidi) 27 Figure 2.5.8. Bridge Survey Form (Saadeghvaziri and Hadidi) 29 Figure 2.5.9. End Condition Effect on Deck Cracking (Saadeghvaziri and Hadidi) 30 Figure 3.1.1. Site Plan and Cross-Section 35 Figure 3.3.1. Measurement of Slump 39 Figure 3.3.2. Type – B Pressure Meter 40 Figure 3.3.3. Free Shrinkage and Cylinder Sample Molds 41 Figure 3.3.4. Restrained Shrinkage Ring Mold 41 Figure 3.3.5. Sample Covered with Wet Burlap 42 Figure 3.3.6. Sample Sealed in Polyethylene Sheet 42 Figure 3.4.1. Geokon, Inc. VWSG for Embedment 42 Figure 3.4.2. VWSG Installation 43 Figure 3.4.3. VWSG Locations 44 Figure 3.4.4. Portable CR1000 Data Logger 45 Figure 3.4.5. Kistler Single Axis Accelerometer 46 Figure 3.4.6. Laser Doppler Vibrometer (Polytec, Inc.) 47 Figure 3.4.6. Installed Strain Transducer 48 Figure 3.4.7. STS Transducer, Receiver Box and Collection Unit 48 Figure 3.5.1. Forney One Million Pound Machine 50 Figure 3.5.2. Tinius Olsen Compression Machine 50 Figure 3.5.3. Concrete Sample under Loading 50 Figure 3.5.4. Concrete Sample with Compressometer 51 Figure 3.5.5. Free Shrinkage Prisms 52 Figure 3.5.6. Length Comparator with Reference Bar (right) and Concrete Prism (left) 52 Figure 3.5.1. Environmental Chamber 54 Figure 4.2.1. Compressive Strength of Concrete Mix over Time 57 Figure 4.2.2. Splitting Tensile Strength of Concrete Mix over Time 58 Figure 4.2.3. Elastic Modulus of Concrete Mix over Time 59 Figure 4.2.4. Free Shrinkage of Concrete Mix over Time 60 Figure 4.3.1. Final Crack Mapping of Ring Specimen 1 (Side Profile) 62 Figure 4.3.2. Final Crack Mapping of Ring Specimen 1 (Top/Bottom Profile) 63 Figure 4.3.3. Final Crack Mapping of Ring Specimen 2 (Side Profile) 65 Figure 4.3.4. Final Crack Mapping of Ring Specimen 2 (Top/Bottom Profile) 66 Figure 4.4.1. Longitudinal Mid-span Strains at Top of Deck During Early Age of Concrete 68 Figure 4.4.2. Transverse Mid-span Strains at Top of Deck During Early Age of Concrete 69 Figure 4.4.3. Longitudinal Mid-span Strains at Bottom of Deck During Early Age of Concrete 69 Figure 4.4.4. Transverse Mid-span Strains at Bottom of Deck During Early Age of Concrete 70 Figure 4.4.5. Longitudinal Quarter-span Strains at Top of Deck During Early Age of Concrete 71 Figure 4.4.6. Transverse Quarter-span Strains at Top of Deck During Early Age of Concrete 71 Figure 4.4.7. Longitudinal Strains Along Top of Mid-span During Long-Term Period 72

Figure 4.4.8. Transverse Strains Along Top of Mid-span During Long-Term Period 73 Figure 4.4.9. Longitudinal Strains Along Bottom of Mid-span During Long-Term Period 73 Figure 4.4.10. Transverse Strains Along Bottom of Mid-span During Long-Term Period 74 Figure 4.4.11. Transverse Strains Along Top of Quarter-span During Long-Term Period 75 Figure 4.4.12. Transverse Strains Along Top of Quarter-span During Long-Term Period 75 Figure 4.4.13. Crack Mapping of Bridge Deck with Sensor Locations 77 Figure 4.4.14. Crack Microscope Used in Deck Crack Mapping 78 Figure 4.5.15. VWSG Temperatures During Initial 36 Hour Period 80 Figure 5.1.1. Four Node Shell Element Detailed with Integration Points 83 Figure 5.1.2. Typical stress-strain curves of structural steel (Salmon and Johnson, 1997) 86 Figure 5.2.1. ABAQUS FE Model Layout 89 Figure 5.2.2. Loading Cases Considered in FE Analysis 90 Figure 5.2.3. FE Diagram Index 91 Figure 5.2.4. Deck Strains Due to Case 1 Live Load 92 Figure 5.2.5. Deck Strains Due to Case 2 Live Load 93 Figure 5.2.6. Comparison of Stress Distribution to Crack Orientation 94 Figure 5.2.7. Deck Strains Due to Case 3 Live Load 95 Figure 5.2.8. Deck Strains Due to Case 4 Live Load 96 Figure 5.2.9. Deck Strains Due to Case 5 Live Load 98 Figure 5.2.10. Deck Strains Due to Case 6 Live Load 99 Figure 5.2.11. Deck Strains Due to Case 7 Live Load 100 Figure 5.2.11. Deck Strains Due to Case 7 Live Load 101

CHAPTER I

INTRODUCTION

1.1 PROBLEM STATEMENT

Possibly the greatest threat to the integrity of concrete is the unavoidable cracking that occurs. Excessive cracking results in weakening from freeze-thaw cycles, corrosion of reinforcing steel because of chloride infiltration and increased maintenance costs. The ability of water and chlorides to penetrate through cracks reduces the service life of the structure and can affect the ability to function as designed. Cracking can occur for several reasons including concrete shrinkage during curing, temperature changes, lack of satisfactory support, magnification of applied loads and restraint conditions.

Proposed in 1998 as a provisional standard the AASHTO Passive Ring Test simulates shrinkage cracking experienced by concrete under partially restrained conditions. The test consists of casting a concrete ring around an inner steel ring with foil strain gauges installed along the middle of the ring’s inner circumference. As shrinkage occurs compressive stresses form in the steel while tensile stresses generate in the concrete. If the strains produced in the concrete exceed that of the splitting tensile strain, cracking will occur. Due to the many variables responsible for deck cracking the tests design is for comparison of mixes and selection of a mix that will perform well

2 under restraint. As of 2006 AASHTO has been balloting for the acceptance under full standards.

Using mixes with low cracking potential is a great way to increase the service life of a bridge and reduce maintenance requirements. The ability to determine the performance of a mix prior to its use in the field would improve the integrity of partially restrained bridge decks. The goal of this thesis is to determine the validity of the laboratory ring tests method to simulate actual bridge deck restrained shrinkage.

1.2 RESEARCH OBJECTIVES AND SCOPE

The goal of this research is to test the ring tests prediction of restrained shrinkage of high performance concrete bridge decks. Collection of field samples took place during the deck pouring. All samples underwent a fourteen-day wet-cure. Determination of basic properties was possible through testing of the samples. Properties necessary for analysis include compressive and splitting tensile strength, modulus of elasticity, and free and restrained shrinkage.

1.3 THESIS ORGANIZATION

This thesis is structured into five chapters as follows:

Chapter I covers the introduction, research objectives and scope as well as the organization of the thesis.

Chapter II covers a general background as well as a literature review of the various types of shrinkage, the factors that affect the shrinkage of a concrete structure, the

3 performance of HPC as it pertains to free shrinkage and restrained shrinkage as well as similar work performed by others.

Chapter III covers the experimental setup including all mechanical and field testing, field implementation and the material properties of the concrete mix analyzed.

Chapter IV covers the finite element modeling employed in this study.

Chapter V covers all test results.

Chapter VI covers the summary and conclusions of the thesis as well as goals for future research.

CHAPTER II

LITERATURE REVIEW

2.1 INTRODUCTION

Concrete will always experience some change in volume. If located in a region with no restraint, this change will not result in any cracking. The concrete mass will merely change in volume. Foundations, subgrades, reinforcement and connecting members all act as partial restraint in field conditions. To examine the effects of volume change it is important to understand the reasons for this change to occur.

Because most concrete elements are significantly longer in one dimension than the other two, volume change is described linearly when referring to concrete. Typically, volume reduction occurs because of changes in moisture and temperature. This change is termed shrinkage. The four types of shrinkage of concern to engineers are plastic, thermal, autogenous and drying shrinkage.

2.2 TYPES OF SHRINKAGE

Shrinkage begins immediately after the pouring if fresh concrete. Volume change can continue for years after curing. Cement properties, type and gradation of aggregates, rate of drying and other reasons affect the type and rate of shrinkage. The following sections detail the types of shrinkage.

2.2.1 Plastic Shrinkage

Plastic shrinkage of concrete occurs at the surface of fresh concrete. Shortly after pouring concrete, excess water makes way to the surface. Evaporation of surface water faster than the rate of bleeding results in plastic shrinkage, and therefore a reduction in volume of the fresh concrete. HPC displays a low rate of bleeding in comparison to normal concrete. This low rate of bleeding requires caution to avoid plastic shrinkage during pouring.

Protection from evaporation removes plastic shrinkage as a concern. Fogging, windbreaks, shading, plastic sheet covers, or wet burlap help in protecting against plastic shrinkage.

2.2.2 Thermal Shrinkage

Thermal shrinkage is a result of the generation of heat of hydration. Heat is a by- product of the chemical process by which the cement paste hydrates. This can cause an expansion of the concrete during initial curing. As curing continues and this temperature begins to drop the concrete experiences a decrease in volume and undergoes thermal contraction. This can result in cracking if the concrete cannot dissipate heat effectively.

In slender structures such as bridge decks, heat dissipation occurs rather rapidly.

Therefore any thermal shrinkage is negligible.

2.2.3 Autogenous Shrinkage

Autogenous shrinkage is a change in volume because of cement hydration.

Curing requires water for cement hydration to occur. Without an external water supply

6 the cement will use pore water found through the concretes voids. Continued consumption of pore water results in self-desiccation.

Concrete with a low water to cement ratio, such as HPC, is susceptible to autogenous shrinkage. Curing with an external water supply for a minimum of seven days will eliminate autogenous shrinkage. Specialized shrinkage reducing admixtures are an acceptable method as well.

2.2.4 Drying Shrinkage

Drying shrinkage is the result of volumetric changes in hardened concrete because of evaporation. This occurs gradually and can continue for years after the pouring.

Should there be no restraining conditions present the mass will shrink without any increase in stresses or cracking. Most field conditions, however, include some form of partial restraint (shear connectors, foundations, etc.) which result in shrinkage cracking.

Larger masses exhibit lower rates of drying shrinkage because of their lower surface area to volume ratios. Bridge decks on the other hand, whose dimensions yield high surface areas, are more susceptible to volume change.

The most controllable factor in minimizing drying shrinkage is the water to cement ratio of a concrete mix. Using the smallest possible water content will result in the lowest shrinkage. Similarly, aggregate properties play a role in concrete shrinkage.

Harder aggregates such as quartz, granite, feldspar, limestone and dolomite resist the effects of drying shrinkage because of their high resistance to compression (ACI

Committee 224). Various forms of wet curing will hold off drying shrinkage until curing has ended.

2.3 SHRINKAGE FACTORS

The most important controllable variable in shrinkage is the water content. The water per cubic yard of concrete will contribute more to shrinkage than any other aspect of a mix. Higher the water content yields more evaporation during drying. This means greater overall levels of shrinkage. Keeping the water content as low as possible reduces shrinkage experienced by concrete. The water required in a mix is affected by several factors including, but not limited to, aggregate volume, aggregate size and gradation, cement content and fineness, chemical admixtures, and curing history.

One way to regulate the water needed is to use more aggregate. Aggregates typically display higher resistance to compression than the surrounding cement paste.

Added coarse aggregate resists the shrinking paste better and decreases the overall shrinkage experienced. The more aggregate used (within reasonable limits) the higher the concretes resistance to shrinkage. Similarly, the size and gradation of aggregate affects how much shrinkage will result. Using the largest practical size reduces the total surface area of aggregate that requires coating with cement. This results in lower water and cement demands in the mix effectively reducing shrinkage. Aggregate gradation is equally important in reducing shrinkage. A well graded concrete will have smaller and finer aggregate filling in voids leaving less room for cement. This overall decrease in void space translates to a lower cement and water content requirement of the mix.

2.4 RESTRAINED SHRINKAGE RING TEST

There are several methods available for testing the restrained shrinkage performance of concrete mixes. These methods include the flat panel test, linear restrained test, and restrained shrinkage ring test. The most popular is the ring test as it is simplistic and economical.

2.4.1 Ring Test Background

The restrained shrinkage ring test consists of casting a concrete ring around a steel ring of smaller diameter. As shrinkage occurs stresses form throughout the system.

Within the steel these stresses are compressive and four foil strain gauges (FSG), installed along the inner circumference of the ring, measure the resulting strain. The concrete experiences tensile stresses that counterbalance the compressive stresses in the steel. Should these tensile stresses exceed the allowable stresses within the concrete cracking will occur. Installing VWSG along the top of the rings allows for quantification of concrete stresses even if cracking does not occur.

The first testing using restrained rings were performed by R.W. Carlson and T.J.

Reading (1988) in the study of shrinkage cracking in concrete building walls. At the time there were no standard methods for testing restrained shrinkage available. In the test concrete was cast around polished steel rings and dried 25, 50 and 75% relative humidity to discover the effects on restrained shrinkage. The rings were smaller than modern

3 methods allowing the aggregate possible for testing to be no more than /8” (9 mm) in size. The steel rings used were coated with paraffin wax to allow slippage of concrete on the ring if cracking occurred. Two sides of the ring were sealed to allow drying only

9 from the outer surface. The investigation found the concrete exposed to lower humidity, and therefore more rigorous drying, withstood higher stresses before cracking. Cracking also occurred more quickly in these samples. The study found that aggregate type plays a large role in crack resistance and the ring test provided useful information and warranted further study as a possible standard test method.

Grzybowski and Shah (1990) modified the previous test in their analysis of shrinkage cracking of fiber reinforced concrete. Incorporating two different types of fibers, steel and polypropylene, the study focused on the resistance to shrinkage cracking as a result of composite reinforcement. The findings stated introducing a small amount of steel fibers (0.25% by volume) reduced average crack widths by as much as 20% and maximum crack widths by 50%. Polypropylene fibers, on the other hand, were determined to be far less effective at reducing crack widths.

In 2002, Weiss and Shah studied the role of shrinkage reducing admixtures as well as specimen geometry and their relationship to shrinkage cracking. The study used two different sizes of restraining rings. The tall group consisted of 9 rings with a height of 150 mm while the short group incorporated 9 rings with a height of 30 mm. Each group was further classified according to wall thicknesses of 25, 75, and 150 mm (the study consisted of three samples of each per group). Analysis of the samples showed cracking occurring in the 25 mm short rings at an average of 8 days as opposed to 11 days in the 50 mm ring. No cracking occurred on the 150 mm thick ring. Within the tall group only two of the three 25 mm thick rings showed signs of cracking while all other specimens remained uncracked. These results led the researchers to conclude the early age cracking behavior of concrete is geometrically dependent.

The work of Grzybowski and Shah was expanded on by Ahmed and Mihasi

(2009) by applying the study of restrained shrinkage to High Performance Fiber

Reinforced Cementitious Composites (HPFRCC). The test samples were prepared by casting a 40 mm thick concrete ring around a 150 mm steel ring. The drying environment consisted of a relative humidity of 60% and a temperature of 68 ⁰ F. A premix mortar specimen was cast for comparison to the fiber reinforced specimen. The study found that while the reinforced sample yielded many cracks (the premix mortar showed only 6 cracks), the crack widths were significantly smaller in the reinforced mortar. The higher number of cracks with smaller crack widths was credited to the strain hardening and toughness properties of the HPFRCC.

2.4.2 Ring Test Setup

The AASHTO Ring Test was developed for testing the effects of concrete mix parameters on shrinkage under restrained conditions. Altering mix ingredients with the ring test provides researchers with the opportunity to understand the effects each component has on the restrained shrinkage of a particular concrete or cementitious composite. The advantage of this test lies in its simplicity of construction and versatility.

The test allows customization to study nearly any variable of a concrete mix design.

The ring setup consists of a concrete ring cast around an inner steel ring. The

1 apparatus requires a steel ring with a wall thickness of 12.7 mm ± 0.4 mm (0.5 in ± /64 in), an outer diameter of 305 mm (12 in) and a height of 152 mm (6 in). The concrete mold should allow for the hardened concrete ring to have an outer diameter of 457 mm

(18 in). These dimensions will result in the desired 3” concrete wall thickness. The

AASHTO geometric requirements can be seen in Figure 2.4.1.

Figure 2.4.1. AASHTO Ring Test Geometry

The AASHTO ring test uses a smaller steel ring than similar test methods (ASTM

Ring Test). This allows for the use of larger aggregate sizes in testing. The increased concrete wall thickness causes cracking to occur at later ages than in similar test methods.

Occasionally no cracking occurs at all. Installing FSGs along the inner circumference of the steel ring aids in determining the cracking age of a mix. These gauges measure the compressive strain in the steel throughout the test. Abrupt changes in strain are credited to cracking in the concrete. This allows for precise determination of cracking age.

2.5 Previous Work

In an attempt to develop a method to predict the shrinkage cracking of concrete

Shah, Ouyang, et. al (1998), used standard ring tests to assess shrinkage ring cracking.

Up to that time ring testing had mainly been used for evaluation of concrete cracking as

12 opposed to crack prediction. The goal of the study was to produce a model that could be provide a guide for the design of concrete pavements and slabs.

The model itself was founded on the principles of fracture mechanics. The assumption that cracking in concrete is related more to fracture energy than tensile strength was backed by fiber reinforcement. The research team sited introducing fiber reinforcement delays cracking without significantly affecting the allowable tensile stress of the concrete. Fiber reinforcement does however alter the fracture energy of a concrete specimen. This reasoning led the team to focus on fracture mechanics in their study as opposed to the more traditional allowable tensile stress method.

The fracture resistance curve approach (R-curve) was used in the generation of the model. Fracture resistance theory is described as follows: By applying a load to a structure with an initial crack length of a 0, strain energy, U, is created. The rate of strain energy release with any arbitrary crack length, a, is known as the strain energy release rate, G. As load is applied some of the energy is also consumed as the crack tip propagates. This propagation energy is defined as W. The rate of change of W with respect to crack length, a, is known as the fracture resistance, R. A crack propagates unsteadily when the following two conditions are true:

[Eq. 4.1]

Equation 4.1 is considered the energy requirement for unstable crack growth where G is considered the crack driving force and R is considered the crack resistance force.

Linear Elastic Fracture Mechanics (LEFM) defines G as:

[Eq. 4.2] where: = Stress Intensity Factor = Modulus of Elasticity

R, as defined by a previously derived R-curve, is calculated as:

[Eq. 4.3]

where: [Eq. 4.4] 1

and: [Eq. 4.5] ,

where: Constants Based on Material Fracture Parameters ,

The stress intensity factor, , was computed using the boundary element method. By generating the element mesh in Figure 2.5.1 it was possible for the researches to compute the required parameters for analysis.

Figure 2.5.1. Shah, Ouyang, et. al. Element Mesh

The stress intensity factor was found to be:

. [Eq. 4.6] √ /

where: = Ratio of External Ring Radius to Internal Ring Radius (r e/r i) = Geometric Function of Ring /

and: [Eq. 4.7]

where: Stress at Cracking

The final parameters required for the generation of the R-curve for the rings in the study,

α and β, were found to be related to two fracture parameters. The critical stress intensity factor, , and the critical crack tip opening displacement, . Applying further principles of fracture mechanics allowed the research team to define the parameters as follows:

[Eq. 4.8] 1.122

where: = Critical Cracking Stress

and:

. [Eq. 4.9] E 1 0.50 0.434 0.154 1

By calculating the ratio of the squares of Eq. 4.9 to Eq. 4.8, α was determined:

. . . [Eq. 4.10]. 1 1

At which point β could be solved for as well:

[Eq. 4.11]

The α and β parameters were used to calculate the respective R- and G-curves for the ring specimens under study. By substituting the resulting curves into Eq. 4.1 and 4.2 the team was able to predict the tensile stresses the concrete would exhibit over time. The theoretical values were found to match closely with experimentally obtained values from laboratory testing (Figure 2.5.2).

Figure 2.5.2. Tensile Strength Prediction (A) and Experimental (B) (Shah, Ouyang, et. al.)

Similarly, the cracking age was predicted by equating the difference between the measured free shrinkage and estimated creep and the maximum tensile strain. Predicted values were again found to be reasonably close to experimental.

In 2000, Whiting, et al. aimed to determine the optimum silica fume content and w /c ratio for bridge deck applications. In order to do so the research program included

Types I and II Portland cement as well as Type K shrinkage compensating cement. The

Type K cement was incorporated in the study to evaluate its effectiveness at reducing the cracking tendency of selected mixes.

Two cases were analyzed – full depth decks (FDDs) and relatively thin deck overlays (TDOs). The mixes for each case followed AASHTO proportion requirements

kg 3 w in order to best simulate field tendencies (370 /m cementitious content and /c of 0.35 to

kg 3 w 0.45 for FDDs and 415 /m and /c of 0.3 to 0.4 for TDOs). The study was further

broken down into two phases in order to analyze the effect of silica fume on a concretes

susceptibility to cracking. The proportions for Phases I and II can be seen in Table 2.5-1.

Table 2.5-1. Phase I and II Mix Proportions (Whiting, et. al. 2000)

kg 3 Material Quantities ( /m ) Phase I - FDD % Silica Fume 0.0 1.8 6.0 10.2 12.0 w/c 0.40 0.36 0.43 0.35 0.40 0.45 0.36 0.43 0.40 Cement 369 364 360 343 342 341 331 328 321 Silica Fume 0 7 7 22 22 21 37 37 44 Fine Aggregate 733 751 716 743 722 699 742 708 720 Coarse Aggregate 1071 1097 1046 1085 1055 1020 1084 1034 1052 Water 148 133 158 128 146 163 132 157 146 Phase I - TDO % Silica Fume 0.0 1.8 6.0 10.2 12.0 w/c 0.35 0.31 0.38 0.30 0.35 0.40 0.31 0.38 0.35 Cement 414 407 408 386 387 392 370 373 363 Silica Fume 0 8 8 24 25 25 42 42 49 Fine Aggregate 852 874 839 869 843 826 871 832 840 Coarse Aggregate 865 887 852 882 856 839 876 845 853 Water 145 129 158 123 144 167 128 158 144 Phase II - FDD % Silica Fume 0.0 6.0 9.0 w/c 0.35 0.40 0.45 0.35 0.40 0.45 0.35 0.40 0.45 Cement 369 369 367 347 344 346 336 336 335 Silica Fume 0 0 0 22 22 22 33 33 33 Fine Aggregate 735 716 693 732 706 692 733 712 699 Coarse Aggregate 1119 1090 1055 1115 1075 1054 1116 1083 1065 Water 148 148 165 129 146 166 129 148 159

All specimens were consolidating by rodding and wet cured for 24 hours under

wet burlap before being stored in an environmental chamber. Two sample types were

cast – drying shrinkage prisms (AASHTO T 160) and shrinkage rings (NCHRP Project

12-37) to evaluate cracking tendency. The prisms were measured at 4, 7, 14 and 28 days

18 as well as at 8, 16, 44 and 64 weeks while the rings were checked throughout the duration of the test cycle to monitor cracking.

At the completion of the test cycle the researchers verified the relationship between the w/c of a HPC mix to the amount of shrinkage it experiences (Fig. 2.5.1).

Figure 2.5.3. Shrinkage of HPC mixes with varying w/c and silica fume content (Whiting, et. al)

It was determined that the full depth mixes generally experienced less shrinkage than the thin deck overlays. This was attributed to the lower paste content of the FDD mixes as well as the longer curing time (FDD mixes were cured for 7 days, TDO mixes were cured for 3 days). This trend can be seen in Figure 2.5.4.

Figure 2.5.4. Shrinkage of HPC mixes with varying geometry and w/c (Whiting, et. al)

Of interest in this study was the discovery made about the Type K shrinkage compensating cement. Observations yielded that those mixes using Type K cement showed significantly greater shrinkage at all test dates. While this goes against the conventional wisdom of why Type K cement would be used the initial expansion of the mix induced compressive stresses within the concrete. These stresses must be overcome by shrinkage for any tensile stresses to form. While the shrinkage of Type K cement was found to be higher the compressive stresses from expansion during initial set offset any tensile stresses that may have formed cracking.

The test group samples (no silica fume) were found to crack around day 35.

Adding silica fume to these mixes resulted in cracking of 1.8 to 4% later. It was also

w discovered that those mixes with the highest /c and a silica fume content of 6% were found to crack sooner than the test group. This was to be expected because of the higher

w /c ratio. No cracking was noted in the Type K mixes at any point throughout the duration of the study. The observations of the FDD mix shrinkage can be seen in Table

2.5-2.

Table 2.5-2. Time to First Crack for FDD Mixes (Whiting, et. al. 2000)

Mixture w/c, % silica fume Specimen 1 Specimen 2 Average 0.40, 0% silica fume 32 37 35 0.36, 1.8% silica fume 42 No Crack 42* 0.43, 1.8% silica fume 42 42 42 0.35, 6.0% silica fume 51 No Crack 51* 0.40, 6.0% silica fume No Crack No Crack No Crack 0.45, 6.0% silica fume 18 23 21 0.36, 10.2% silica fume No Crack No Crack No Crack 0.43, 10.2% silica fume No Crack No Crack No Crack 0.40, 12% silica fume No Crack No Crack No Crack *Specimen 1 Data Only

All TDO mixes were observed to crack during the test duration. In contrast with the

FDD mixes the presence of silica fume in thin deck overlays was found to result in cracking sooner than experienced by the test group samples. As mentioned before this was attributed to the higher cement content of the TDO mixes which resulted in higher rates of shrinkage. The observations of the TDO mixes can be seen in Table 2.5-3.

Table 2.5-3. Time to First Crack for TDO Mixes (Whiting, et. al. 2000)

Mixture w/c, % silica fume Specimen 1 Specimen 2 Average 0.35, 0% silica fume 25 20 23 0.30, 6.0% silica fume 8 11 10 0.40, 6.0% silica fume 13 14 14 0.35, 12% silica fume 7 11 9

Phase I results for FDD mixes were determined to be limited in scope. Therefore

w Phase II was developed. Phase II incorporated the same variables found in Phase I ( /c,

w % silica fume and curing time). Three /c (0.35, 0.40 and 0.45), three silica fume contents (0, 6 and 9%) and two wet curing lengths (1 and 7 days) were used in analysis.

The 1-day wet cure was used to simulate the worst case. This case was included to mimic when curing conditions were not adhered to (such as allowing the moist curing blankets to dry). The 7-day wet cure simulated adherence to common construction standards. The results of the Phase II cracking tendency analysis can be seen in Table

2.5-4.

Table 2.5-4. Time to First Crack for Phase II Mixes (Whiting, et. al. 2000)

One Day Wet Cure Seven Day Wet Cure

w Silica Fume Content w Silica Fume Content /c /c 0% 6% 9% 0% 6% 9% 33 19 17 33 25 30 36 14 18 35 20 46 0.45 27 17 11 0.45 38 19 28 27 13 16 40 20 28 22 23 20 33 16 31 24 17 30 36 49 51 26 12 17 24 18 14 0.40 22 23 11 0.40 34 25 34 24 18 22 18 21 55 31 10 14 50 56 52 35 13 11 53 29 67 43 12 23 29 57 32 0.35 22 14 18 0.35 32 17 53 24 28 28 28 37 66 35 17 21 32 64 30

The presence of silica fume was determined to be damaging to the cracking resistance of concrete if only one day of wet curing is allowed. Given the specified seven day wet cure, concrete incorporated with silica fume was found to yield higher cracking resistance

w with lower /c. This observation, as mentioned previously, agrees with accepted theory.

Whiting, et. al. noted that silica fume content has a negligible effect on the long- term shrinkage experienced by bridge decks. Of concern is the effect of silica on early age shrinkage. Based on their findings the researchers concluded shrinkage during early

w ages shows increased sensitivity to /c as the silica fume content is increased.

To study the effects of internal curing Delatte, et. al. (2006) conducted a study into the use of absorptive lightweight aggregate (LWA) or superabsorbent polymers

(SAP) as a means of reducing cracking of HPC bridge decks. Without external hydration

23 in the form of steaming or wet curing the low water-to-cement ratio of HPC mixes

(typically 0.42 or less) will result in self-desiccation because of the lack of water for complete hydration. One possible remedy to this problem is the idea of internal curing.

Internal curing is defined as a proposed increase in strength gain accompanied by a reduction in early age shrinkage. Soaking aggregates before use in mixing it is possible to introduce extra water into the pore spaces that aids in hydrating the cement internally.

The initial concern over the effectiveness came from a field survey of six bridges found in North Eastern Ohio. The results of this survey are detailed in Table 2.5-5 and on review it becomes clear there may be a connection between the use of absorbent aggregate and crack resistance.

Table 2.5-5. North Eastern Ohio Field Survey Results (Delatte, et. al.)

Bridge Decks with Transverse Cracks Project Location Coarse Aggregate Absorption 281('99) WB I-480 over Rockside 0.41 197('00) EB Fairmount over I-271 0.41 528('00) Dover Center, Cahoon and Canterberry 0.41 107('01) Wagar, Northview and W159 th 0.41 Uncracked Bridge Decks Project Location Coarse Aggregate Absorption 480('99) NB and SB I-271 over Tinkers Creek 1.39 157('01) Highland Road over I-271 1.52

To discover a trend between the use of absorptive LWA and crack reduction the team created 21 mixes. Following the seven mix proportions shown in Table 2.5-6, each was mixed using low, medium and highly absorbent LWA.

Table 2.5-6. Absorptive Light Weight Aggregate Mix Proportions (Delatte, et. al.)

CA # w Mixture 57 CA # 8 FA LWA Cement SCM /c HP3 - 1490 1355 - 480 180* 0.40 HP4 - 1490 1370 - 440 220** 0.40 HP4 Blend 1 1335 360 1245 - 400 200*** 0.42 HP3 Blend 2 1184 297 1350 - 480 180* 0.40 HP3 LWA - 1490 1012 227 480 180* 0.40 HP$ LWA - 1490 1041 227 440 220** 0.40 HP4 Blend LWA 1335 360 946 206 400 200*** 0.42 *HP3 – 150 lbs. of fly ash and 30 lbs. of silica fume p.c.y. **HP4 – 190 lbs. of ground granulated blast furnace slag and 30 lbs. of silica fume p.c.y. ***HP4 Blend 1 and Blend LWA – 170 lbs. of ground granulated blast furnace slag and 30 pounds of silica fume p.c.y.

All mixes were cast into standard ring specimens and monitored throughout the test

duration to detail any cracking that occurred.

A summary of the results is provided in Table 2.5-7.

Table 2.5-7. Cracking Dates for Ring Specimens (Delatte, et. al.)

Mixture Minimum Time Maximum Time Average Standard Deviation HP Low Absorption 23 37 32.8 6.9 HP Medium Absorption 23 90 47.1 29.5 HP High Absorption 11 90 26.7 30.0 HP Low Absorption with LWA 19 81 46.3 29.9 HP Blend 33 59 55.6 9.1

The mixtures with absorbent LWA were found to have a higher strength than those mixes

without especially at early ages. The average increase in compressive strength at 7 and

28 days was found to be 6.4%. Regarding cracking resistance, however, the HP blends

(no LWA) displayed the longest average days to cracking. The highly absorptive

aggregate mixes were found to be the least resistant to crack formation while the low and

medium aggregates were similar with the medium being slightly more resistant to crack

25 formation. These results do not reproduce the field data consistently enough to warrant any declaration of the effectiveness of absorbent LWA in crack reduction. However, the use of LWA is growing and a new experimental setup is underway that plans to incorporate core samples from the field as well.

Trying to develop new test methods for assessing the shrinkage cracking potential of concrete mixes, Weiss, et. al (1998) performed a study of a new method for shrinkage crack analysis designed to consider bridge decks of finite length as well as subbase friction. Incorporating normal concrete and HPC mixes, the study used ASTM Type I

Portland cement, graded river sand, and 9mm pea gravel aggregate (ASTM C-33) for all mixes. An evaluation of a non-expansive shrinkage reducing admixture was incorporated into the study as well. All specimens were wet cured for 24 hours before demolding and storage in an environmental chamber (40% R.H., 30°C). Compression and free shrinkage measurements were determined using standard testing methods. Measurement of restrained shrinkage were taken according to three setups – the standard ring test, the restrained end test (RE test) and the restrained base and end test (RBE test). The ring test was included to simulate a bridge deck of infinite length. The RE test was developed to analyze slabs of finite length and the RBE test was developed to consider the effects of subbase friction on bridge deck shrinkage. The apparatus developed for the RE and RBE test setups can be seen in Figure 2.5.5.

Figure 2.5.5. RE and RBE Test Method Setup (Weiss, et. al.)

Results of the time to first three test specimens can be seen in Table 2.5-8. As can be seen the high strength specimens cracked earlier than the normal concrete samples.

This was attributed to the higher initial strain, initial set and gain in stiffness of high strength concrete in comparison to normal concrete.

Table 2.5-8. Time to Cracking of Ring, RE and RBE Test Samples (Weiss, et. al.)

Specimen Experimental Model Experimental Model Experimental Model Geometry Average (0%) (0%) Average (1%) (1%) Average (2%) (2%) Normal Strength Concrete (NSC) Rings 10 8.6 11.7 19.5 47.5 38 RE 7 7.4 11.7 16.7 - 32.7 RBE 11.3 9.1 26.3 19.8 - 40.5 High Strength Concrete (HSC) Rings 3.2 2.7 4.8 3.5 11.7 12.4 RE 3.5 2.2 3.5 2.9 6 9.4 RBE 4.3 2.5 4 3.3 9 11.3

Adding shrinkage reducing admixtures (SRA) to the mixes was found to reduce the free shrinkage of all tested samples significantly (nearly 45% reduction for a 2% addition of

SRA). The major advantage of SRA was its ability to delay the time to cracking which resulted in smaller crack widths which resulted in more durable concrete. The researchers also saw that theoretical predictions of the day to cracking (R-Curve Method,

Shah, 1998) agreed reasonably with experimental results.

In an attempt to understand the increased bridge deck cracking that has been discovered throughout New Jersey, Saadeghvaziri and Hadidi (2002) organized a field survey of 24 bridges throughout the state. An example of the typical cracking seen in many NJ bridges can be seen in Figure 2.5.6. Cracking such as that shown in the figure can lead to corrosion of reinforcement and decreased structural integrity.

Figure 2.5.6. Typical Cracking Observed on NJ Bridge Decks (Saadeghvaziri and Hadidi)

The only requirements for consideration were the bridge must be five years old or younger and must span more than 75 feet (single or multiple spans). Most of the bridges analyzed were found to have cracked and a database was formed based on the survey results. During the evaluation crack spacing, width and location were recorded and all available design and construction documents were reviewed. The bridges surveyed can be seen in Figure 2.5.7.

Figure 2.5.7. Bridge Survey Distribution throughout NJ (Saadeghvaziri and Hadidi)

The database was created for statistical analysis purposes and incorporated five categories – general information, design information, material properties and mix design information, construction information and crack information. General information included items such as the bridge location as well as the year of construction. Design information included specifications such as dimensions, number of spans, traffic directions and girder type. Material properties and mix design information detailed components of the mix such as water content, cement content, admixtures and slump results. This information was taken from NJDOT Inspection and Testing Datasheets.

Construction information taken from NJDOT Inspection and Testing Datasheets provided the air temperature and concrete temperature of the bridge decks during pouring.

Cracking information was recorded during field surveying and included crack type, approximate location, crack spacing and crack width.

The field study was created to identify causes of transverse cracking in bridge decks. During the survey four bridges were instrumented to discover the effects of ambient and hydration temperatures as well as the role of shrinkage on deck strains and stresses. Along with NJDOT Inspection and Testing Datasheets the information collected from the bridge survey forms (Figure 2.5.8) provided the researchers with enough information to conduct a statistical as well as finite element analysis.

Figure 2.5.8. Bridge Survey Form (Saadeghvaziri and Hadidi)

Completion of the statistical and finite element analysis allowed the researchers to conclude the end conditions of a deck play a large part in transverse cracking (Figure

2.5.9). It was determined that increased fixity in a bridge leads to increased deck cracking.

Figure 2.5.9. End Condition Effect on Deck Cracking (Saadeghvaziri and Hadidi)

The researchers also found that increases in deck thickness typically reduces crack formation. This agreed with similar research (Krauss and Rogalla, 1996). Finding that bridge decks with higher compressive strengths show higher rates of cracking, it was recommended that an upper limit be placed on the strength of a designed mix before pouring. As detailed in Figure 2.5.9, design boundary condition should be consistent with actual field conditions. Last, the research team suggests the use of restrained ring

32 testing before any implementation of concrete in the field to evaluate its cracking potential. It was noted, however, that while several agencies currently used restrained ring testing, little research has been performed on the quantification of results from such testing and more is needed before ring testing can be universally accepted.

A similar study was performed in 1999 by French, Eppers, and others to examine the causes of early age cracking in Minnesota bridge decks. The study was divided into two phases. Phase one incorporated the field investigation of 72 bridges to find the correlation between observed cracking and available design, material and construction related data. Phase two included a shrinkage and parametric study to test variables not easily analyzed during field investigation. In the shrinkage study two field mixes were tested in a laboratory to study the shrinkage of the specimens over time. The parametric study was designed to explore the influence of various factors including end restraint, girder stiffness, cross frame and splice location as well as differential temperature between the girders and deck. The study used finite element analysis to control and adjust relevant variables. Three bridges were modeled and compared with field observations for the parametric study.

The field investigation surveyed 72 bridges found throughout the state of

Minnesota – 34 composite simply supported prestressed girder bridges, 34 composite continuous steel plate girder bridges and 4 continuous steel rolled wide flange girder bridges. The selected bridges represented the variety of bridges found throughout the state as the population included both young and old bridges as well as local and interstate highway bridges.

The field survey found that simply supported prestressed girder bridges were in good condition as opposed to continuous steel girder bridges. This was attributed to the reduced end restrained and beneficial creep and shrinkage characteristics of prestressed bridge superstructures. The dominant design parameter found to have an effect on transverse cracking was the restraint coefficient, β. The restraint coefficient shows the degree of deck restraint provided by steel girders and is defined as the ratio of the cross sectional area of the girder to the effective area if the concrete slab. Typically, β values of 0.05 or less represent limited restraint and therefore a lower possibility of deck cracking. β values of higher than 0.12 represent high restraint and as a result have a higher chance of transverse cracking (Ducret, et al). Evaluation of construction documents also found that larger bar spacing (within design limits) of transverse top reinforcement correlated with lower cracking potential of those bridge decks surveyed.

The dominant factors were determined to be cement content, aggregate type and quantity and the air content of the concrete used in construction. Increased cement content was found to result in higher levels of cracking along the deck. Also, those mixes with typical to moderately high cement contents were not dramatically affected by

w changes in water content or /c ratio. Decks with higher quantities of aggregate typically required lower amounts of cement paste and were found to display lower levels of cracking than decks with lower aggregate quantities. As shrinkage occurs mainly in cement paste, the decrease in cement due to increased aggregate quantities results in concrete that experiences less shrinkage and therefore less cracking. It was noted, however, the use of aggregates with a high modulus could potentially increase the shrinkage of a mix and aggregates utilized in bridge decks should have lower moduli

(within design limits) to combat shrinkage cracking. The limited data obtained on the air content of design mixes showed that increased air content decreased observed cracking.

While this data is limited (all decks fell within 5-6% air content) this observation agreed with the results of others research.

Determination of the construction related parameters and their effect on deck cracking was determined by reviewing the high and low ambient temperatures experienced during the pouring of 18 bridge decks. The study found a slight trend between the ambient temperature and reduced cracking on bridge decks. Those decks poured on days that had a high temperature of 65 ºF to 75 ºF and a low of 45 ºF to 50 ºF showed lower levels of cracking than those bridges poured on warmer or cooler days.

8 Of the 34 prestressed bridges surveyed 25 were rated as /10 or higher. The major cracking factors in prestressed bridge decks were found to be deck overlays and reconstructions. When decks are replaced on older bridges the prestressed superstructure has experienced most of the creep and shrinkage it will in its lifetime and as a result acts more like the rigid superstructure of steel girder bridges. Of the 38 steel bridges studied

8 only 12 rated as /10 or higher which was attributed to the higher restraint present in steel bridges.

The shrinkage study incorporated PBEAM , a finite element program that uses a fiber, or layer, approach. The critical variable in deck cracking due to shrinkage was determined to be the differential shrinkage between the deck and the girders. This movement and the restraint present in bridge decks was found to lead to transverse cracking at critical locations throughout the deck. While the shrinkage experienced by the deck was determined to have little effect on transverse cracking the initial rate of

35 shrinkage did have a large impact on the cracking potential. Reducing initial rates of shrinkage was determined to be a key way of preventing early age cracking in bridge decks.

The greatest effect on the extent of cracking being the end restraint conditions was verified within the finite model. The most extensive cracking occurred on the most restrained bridges. Locations of cross frames and splices were varied throughout the analysis and it was inferred the location of these components have little effect on the cracking a bridge deck experiences.

CHAPTER III

EXPERIMENTAL SETUP

3.1 INTRODUCTION

The experimental setup consists of the testing of an HPC mix utilized by the

Garden State Parkway in the Interchange 67 overpass (Figure 3.1.1) as well as laboratory samples taken during the pour.

Figure 3.1.1. Site Plan and Cross-Section

The bridge is located in Barnegat Township, NJ, and spanned the southbound New Jersey

Turnpike at exit 67.

Field strains were collected with embedded vibrating wire strain gages (VWSG).

Readings were taken every five minutes for the first five days and every hour after day five. Strain readings were recorded with the use of a portable data logger stationed in the field and powered by a rechargeable battery.

Acceleration of the bridge deck was observed with single axis accelerometers.

Readings were taken just prior to pouring of the deck and immediately afterwards. This was done to record the deck acceleration during the two traffic phases of construction.

Acceleration data was recorded with the use of the eDaq data acquisition system.

Displacement and velocity readings were collected with a Laser Doppler

Vibrometer which allowed for the simultaneous collection of both values. Strain readings in the steel were obtained by installing strain transducers at predetermined locations throughout the steel superstructure of the overpass.

3.2 MATERIAL PROPERTIES OF MIX

The NJ Turnpike mix included silica fume and fly ash as a percentage of the total

w cementitious content. The mix yielded a /c ratio of 0.40 and contained 658 lb. of cementitious material. The mix design and all relevant details can be seen in Table 3.2-1.

Table 3.2-1. NJ Turnpike Mix Design

NJ Turnpike Mixes for All Projects - HPC Component Specific Gravity Provider Location Quantity (lbs.) Water 1.00 260 Cement 3.15 Essroc Type 1 - Nazareth, Pa. 501 Cement (Slag) - 132 Cement (Fly Ash) 2.15 Mineral Resources - Jersey City, NJ 25 Cement (Flume) 2.35 Nocchem - Haupauge, NY 1850 Coarse Aggregate 2.88 Trap Rock Industries - Pennington 1184 Sand 2.63 Clayton Sand - Lacey Total 3952

In addition to fly ash and silica fume, the mix included a variety of admixtures to aid in shrinkage resistance and to delay the set time in order to accommodate the large amount of concrete that was required. The admixtures used and their quantities can be seen in Table 3.2-2.

Table 3.2-2. NJ Turnpike Admixture Content

Admixtures Admixture Type Quantity (lb./yd 3) Setcon 6A - Air Entrainment 9.9 Chemstrong A - Water Reducer 19.7 Chemstrong R - Water Reducer/Retarder 6.6 Chemstrong - SP - High Range Water Reducer 79 Total 115.2

3.3 MIXING AND FRESH SAMPLING OF CONCRETE

Mixing was performed by Clayton Concrete Plant located in Toms River, New

Jersey. Fresh concrete testing including slump (ASTM C 143-05a) and air content

(ASTM C-231) were performed by the contractor on site to guarantee adherence to

NJDOT specifications. Samples were consolidated using the rodding technique and cured for 14 days in accordance with NJDOT field specifications. The mixing, sampling, and laboratory testing of the specimens are detailed in the following sections.

3.3.1 Slump Test (ASTM C 143-05a)

The slump of a concrete mix is found using a slump cone. The cone is filled in three layers with each being roughly one-third the volume of the cone. Each layer is rodded twenty-five times through its depth to ensure adequate consolidation. Caution should be taken when rodding to avoid penetration into lower regions that have already been consolidated. The final layer fills the cone until overflow occurs to allow the complete volume be filled before rodding. At the culmination of consolidation any excess concrete is struck off the cone by rolling the tamping rod across the top in a fluid motion.

Once filled and consolidated the cone is removed vertically, allowing the specimen to deform, or “slump”. When lifting the slump cone any motion that could result in lateral or torsional forces should be avoided. The cone is then inverted and placed next to the specimen and the tamping rod help flush against the rim of the cone.

The slump is measured as the distance between the extended rod and the original center of the top surface of the concrete sample. An example of slump measurement can be seen in Figure 3.3.1.

Figure 3.3.1. Measurement of Slump

3.3.2 Air Content (ASTM C – 231)

Air content of a mix is determined by using a Type-B pressure meter. After wetting the inside concrete is added to the pressure meter in three layers, similar to filling the cone in the slump test. Each layer is roughly one-third the volume of the bowl and each layer is rodded twenty-five times throughout its depth. Additionally, to ensure excess air is removed the bowl is tapped on its side with a rubber mallet after consolidation of each layer. After completion of the final layer any excess concrete is struck off in a manner similar to that of the slump test. The top of the pressure meter is checked and cleaned if necessary to make certain a pressure tight seal is possible.

After securing the lid onto the bowl the valve between the chamber and the bowl is be closed and the petcocks opened. Water is then injected into one petcock with the use of a rubber syringe until it is seen escaping from the second petcock. After closing both petcocks air is pumped into the sealed meter until the gauge hand comes to rest on the initial line. Added time is taken before completion of the test to ensure the gauge

41 hand does not move from the initial position as a result of cooling of the air within the specimen. The final air content of the sample is found by releasing the air valve and reading the percentage of air off the meters gauge. The Type-B pressure meter can be seen in Figure 3.3.2.

Figure 3.3.2. Type – B Pressure Meter

3.3.3 Sampling of Specimens and Consolidation

Forty-five cylinders (4” x 8”) were taken for standard ASTM tests. Two restrained shrinkage rings were cast for restrained shrinkage testing under AASHTO standards. Three free shrinkage prisms were also taken to determine the free shrinkage of the mix and for correlation with the restrained shrinkage results. All samples were consolidated by rodding. Figures 3.3.3 and 3.3.4 show free shrinkage prism and cylinder molds and restrained shrinkage molds used in casting field specimens, respectively.

Figure 3.3.3. Free Shrinkage and Figure 3.3.4. Restrained Shrinkage Cylinder Sample Molds Ring Mold

3.3.4 Curing

All collected samples were subjected to a fourteen day wet cure according to the

NJDOT field specifications. Samples were cured for the first twenty-four hours in the field to allow for the initial curing conditions to be identical. All specimens were demolded after twenty four hours and covered with wet burlap and polyethylene sheets.

The samples were placed in the curing chamber for the remainder of the fourteen day curing period. The curing chamber was held at a constant relative humidity of 50±4% and a constant temperature of 74 ⁰F. Examples of the wet cure method utilized can be seen in Figures 3.3.5 and 3.3.6.

Figure 3.3.5. Sample Covered with Figure 3.3.6. Sample Sealed in Wet Burlap Polyethylene Sheet

3.4 INSTRUMENT DETAILS and FIELD IMPLEMENTATION

3.4.1 Embedded Vibrating Wire Strain Gauges

To find the strain experienced in the HPC bridge deck fifteen vibrating wire strain gauges (VWSG) designed for embedment in concrete were used. The sensors (fig. 3.4.1) were provided by Geokon, Inc. and are capable of recording the strain as well as the temperature of the surrounding concrete.

Figure 3.4.1. Geokon, Inc. VWSG for Embedment

The basis behind the VWSG is the vibrating wire principle. A steel wire in the gauge is tensioned between the two end blocks of the gauge which is set in the bridge deck. Any

44 deformations result in a change in overall gauge length which can then be used in calculating deck strain at a particular location. Tension readings are acquired by plucking the wire and measuring the resonating frequency with the use of an electromagnetic coil attached directly to the data logger. The gauges also include a thermistor that collects temperature readings as well.

Installation of the gauges was performed by securing the sensors to the reinforcing cage with plastic ties (fig. 3.4.2). Styrofoam spacers were used to insure proper orientation and a secure connection to the cage.

Figure 3.4.2. VWSG Installation

VSWG locations were selected to provide strain data for the mid and quarter span of the deck and can be seen in Figure 3.4.3. Three sensors were installed at each mid-span location. One gauge was installed along the bottom of the cage and two along the top

(transverse and longitudinal directions). Two sensors were installed at each quarter span location (top of cage only). A detail of gauge locations can be seen in Table 3.4-1.

Figure 3.4.3. VWSG Locations

Table 3.4-1. VWSG Orientations and Locations

Sensor No. Girder No. Location Orientation 1 10 Mid-span Top Transverse 2 10 Mid-span Top Longitudinal 3 10 Mid-span Bottom Longitudinal 4 8 Mid-span Bottom Longitudinal 5 8 Mid-span Top Transverse 6 8 Mid-span Top Longitudinal 7 6 Mid-span Top Longitudinal 8 6 Mid-span Top Transverse 9 6 Mid-span Bottom Transverse 10 10 Quarter span Top Longitudinal 11 10 Quarter span Top Transverse 12 8 Quarter span Top Longitudinal 13 8 Quarter span Top Transverse 14 6 Quarter span Top Longitudinal 15 6 Quarter span Top Transverse

3.4.2 Portable Data Logger

Data collection in the field was performed by a Campbell Scientific CR1000 data logger (Figure 3.4.4). With three multiplexors the unit collected data from a maximum of

48 simultaneous channels. Powered by a rechargeable battery the data logger could collect data continuously for fourteen days before needing a replacement battery.

Figure 3.4.4. Portable CR1000 Data Logger

Able to store up to 4 megabytes of information internally, the collected data was downloaded by a laptop computer every two weeks while on location to replace the unit’s battery.

The unit recorded the strains and temperatures in the bridge deck every five minutes for the first week. The interval was changed to hourly collections from day 7 onward to save memory and reduce the overall amount of data for analysis.

3.4.3 Accelerometers

Bridge deck acceleration due to bordering traffic was monitored with single axis accelerometers provided by Kistler (fig. 3.4.5). The instruments were clamped to the bottom flanges of the bridge girders and readings were taken both before pouring and immediately afterwards.

Figure 3.4.5. Kistler Single Axis Accelerometer

This was done to analyze acceleration conditions from the two traffic patterns experienced during and after the pour. Before pouring both of the adjacent lanes remained open to traffic. Only the far lane remained open from the start of the pour until nine hours had passed. By recording values during both phases it was possible to determine the normal acceleration as well as altered traffic acceleration of the bridge deck.

3.4.4 Laser Doppler Vibrometer

Stage one displacement and velocity were recorded with a Laser Doppler

Vibrometer (Polytec, Inc.) (Fig. 3.4.6). The unit operates by bouncing light off installed targets and measuring shifts in the light spectrum as the target moves due to deflection.

Figure 3.4.6. Laser Doppler Vibrometer (Polytec, Inc.)

Reflective targets were placed near the mid-span of girders two and four and readings were taken during several tests both before and after the test to discover the normal deflection and velocity of the members during both traffic patterns.

3.4.5 Structural Testing System

Collection of girder strain as a result of live load was possible through Bridge

Diagnostics, Inc. Structural Testing System I (STS I). Designed for use on highway and railroad bridges the system allows for quick installation of transducers with little surface preparation. The robustness of the transducers allows the sensors to be clamped to the bridge girders with standard C-clamps (fig. 3.4.6).

Figure 3.4.6. Installed Strain Transducer

Data acquisition is possible through a network of STS I receiver boxes installed throughout the structure. Receiving information from up to four sensors, each box is wired in series before being connected to the main collection unit (Fig. 3.4.7).

Figure 3.4.7. STS Transducer, Receiver Box and Collection Unit

3.5 LABORATORY TESTING PROCEDURES

A wide variety of laboratory tests were conducted on the collected samples to

obtain the mechanical properties of the mix. A summary of all tests conducted can be

seen in Table 3.5-1. The specific test methods and procedures are discussed in greater

detail in the following sections.

Table 3.5-1. Summary of Laboratory Tests Performed

No. of Applicable Test Samples Standard Curing Conditions Sample Age at Testing (Days) Slump 1 ASTM C143 None 0, Fresh Fresh Air 1 ASTM C231 None 0, Fresh Content Free Shrinkage 3 ASTM C157 14 Day Wet Cure 1 - 91 Restrained 2 AASHTO PP34 14 Day Wet Cure 1 - Cracking Age (Max. 91 Days) Shrinkage Compressive 15 ASTM C39 14 Day Wet Cure 3, 7, 14, 28, 56 and 91 Strength Splitting Tensile 15 ASTM C496 14 Day Wet Cure 3, 7, 14, 28, 56 and 92 Strength Modulus of 15 ASTM C469 14 Day Wet Cure 3, 7, 14, 28, 56 and 93 Elasticity

3.5.1 Compressive Strength of Concrete Specimens (ASTM C-39-05)

For each testing day two (4” x 8”) specimens were subjected to compression

testing. All samples were tested using a Forney One Million Pound Machine (fig. 3.5.1).

Prior to testing each sample was capped with high strength sulfur to provide uniform

application of compressive force. The loading rate was held constant throughout the

duration of the procedure and the maximum strength for each cylinder was recorded.

Figure 3.5.1. Forney One Million Pound Machine

3.5.2 Splitting Tensile Strength OF Concrete Specimens (C – 496 –04e1)

Similar to the compression testing two (4” x 8”) cylinders were tested in splitting tension for each testing date. Splitting tension was applied by the Tinius Compression

Machine rather than the Forney machine because of the former’s longer head extension.

Also, to minimize human error, a 250-kip digital load cell is applied to the apparatus.

The splitting tensile strength test can be seen in Figures 3.5.2 and 3.5.3.

Figure 3.5.2. Tinius Olsen Compression Figure 3.5.3. Concrete Sample Machine Under Loading

3.5.3 Modulus of Elasticity

The elastic modulus was found by testing two samples each test day in compression with the use of a compressometer (Fig. 3.5.4). Each specimen capped with high strength sulfur and is secured to the compressometer and preloaded to ensure proper installation.

Figure 3.5.4. Concrete Sample with Compressometer

Once the preloading was satisfactory the samples were loaded a minimum of two times at a constant rate of 30-40 psi per second. Loading was increased until 40% of the compressive strength of the concrete (as calculated from compression testing) was reached. Throughout the test load and deformation were recorded regularly. By plotting the stress versus the strain experienced by the sample it was possible to determine the resulting slope, better known as the elastic modulus.

3.5.4 Free Shrinkage Test (ASTM C-157)

The free shrinkage test required preparation of three (3” x 3” x 10”) concrete prisms (Fig. 3.5.5). Each specimen was cast within a steel mold fitted with gauge studs at the extreme ends of the sample.

Figure 3.5.5. Free Shrinkage Prisms

At the end of curing each sample was measured and compared to the a reference bar. All measurements involved in this test were performed with a length comparator (Fig. 3.5.6).

Figure 3.5.6. Length Comparator with Reference Bar (right) and Concrete Prism (left)

Samples were rotated when positioned in the comparator and the minimum measurements were recorded. Length change throughout the testing period was calculated using Equation 3.1.

[ Eq. 3.1 ] 100 where: = Difference in length between specimen and reference bar at day x = Difference in length between specimen and reference bar at day 1 = Total length of specimen (10”)

3.5.5 Restrained Shrinkage Ring Test

Measurement of restrained shrinkage was conducted following the provisions of the AASHTO PP34 test method. Molds were prepared and the rings were cast in the field on the day of the pour. The setup consisted of an inner steel ring around which concrete is cast. A combination of foil strain gauges (FSG) and vibrating wire strain gauges (VWSG) allow for quantification of strain. As the specimen shrinks compressive stresses are produced within the steel ring. These compressive stresses are offset by tensile stresses that occur in the surrounding concrete due to shrinkage. Should these tensile stresses at any point exceed the tensile allowable stress of the concrete cracking will occur. Cracks are monitored daily with the use of a crack microscope and a crack map of the ring is created for analytical purposes.

The AASHTO procedure also requires installing four FSG and the center of the steel ring to monitor stress increases. Sudden changes in steel stress because of stress relief in the concrete due to cracking are an indicator of the precise date of cracking.

Further, six VWSG are also installed across the top of the rings by embedding threaded bolts during casting of each ring. The VWSGs were arranged to note the stresses present in the concrete throughout the test period. Readings exceeding cracking strain signaled the location cracking was expected and aided in the crack surveying of the rings. An added advantage to the VWSG setup is the ability to quantify concrete stresses in those regions that do not experience cracking. This allows for better understanding of the results.

3.5.5.1 Environmental Chamber

Shrinkage of concrete is affected by the temperature and relative humidity of its surrounding environment. As a result it is important to control these causes and therefore ensure the integrity of any collected data. Control of these elements was possible with an environmental chamber (Fig. 3.5.7). Consisting of aluminum insulated walls, the 24’ x

16’ by 8’ room was managed by an external digital control unit.

Figure 3.5.1. Environmental Chamber

Internal sensors regularly monitored the temperature and humidity of the room and communicate this information to the outside unit. Temperature adjustments were automatically made with the series of heating and cooling units. The room also had fans that helped to circulate the air and therefore maintained an acceptable temperature. The relative humidity of the chamber was controlled with a steam generator located within the fan system. Constant monitoring and adjustments made by the chamber control system allowed the room to be kept at a constant temperature of 74 ⁰ F and a constant relative humidity of 50 ± 4%.

CHAPTER IV

TEST RESULTS

4.1 INTRODUCTION

This chapter presents the test results of the testing performed on the NJ Turnpike

HPC mix. Mechanical properties (compressive strength, splitting tensile strength and modulus of elasticity) are presented followed by the results of the free shrinkage and restrained shrinkage testing.

4.2 MECHANICAL PROPERTIES

4.2.1 Compressive Strength

Figure 4.2.1 and Table 4.2-1 depict the change in compressive strength of the mix over time. As can be seen in the figure the 28-day requirement of a compressive strength of at least 5,000 psi was met in accordance with NJDOT standards.

Compressive Strength of Concrete 9000

8000

7000

6000

5000 (psi) c

f ' f 4000

3000

2000

1000

0 0 10 20 30 40 50 60 70 80 90 100 Time (Days)

Figure 4.2.1. Compressive Strength of Concrete Mix over Time

Table 4.2-1. Compressive Strength of Concrete Mix over Time

Mechanical Properties

TIME (Days) σc (psi) 0 0 3 4634 7 5609 14 6026 28 6929 56 7637 91 7531

4.2.2 Tensile Splitting Strength

The tensile strength of the mix can be seen in Figure 4.2.2 and Table 4.2-2.

Splitting Tensile Strength of Concrete 800

700

600

500

400 f 'c (psi) 'c(psi) f 300

200

100

0 0 10 20 30 40 50 60 70 80 90 100 Time (Days)

Figure 4.2.2. Splitting Tensile Strength of Concrete Mix over Time

Table 4.2-2. Compressive Strength of Concrete Mix over Time

Mechanical Properties

TIME (Days) σt (psi) 0 0 3 492 7 507 14 621 28 648 56 731 91 657

4.2.3 Elastic Modulus

The modulus of elasticity of the mix can be seen in Figure 5.2.3 and Table 5.2-3.

The rapid rise during the first seven days followed by a slow decrease over time is due largely to the curing history of the specimens. As the specimen are wet cured the pore network within the samples is constantly filled with water. At the culmination of curing the water gradually becomes replaced with air and the modulus ceases it increase and occasionally, as with the mix in this study, the modulus will even decrease slightly.

Elastic Modulus of Concrete 7000000

6000000

5000000

4000000 (psi) (psi) c

f ' f 3000000

2000000

1000000

0 0 10 20 30 40 50 60 Time (Days)

Figure 4.2.3. Elastic Modulus of Concrete Mix over Time

Table 4.2-3. Compressive Strength of Concrete Mix over Time

Mechanical Properties TIME (Days) Е (psi) 0 0 3 5669060 7 5568868 14 5372478 28 5211666 56 5025342 91 -

4.2.4 Free Shrinkage

The free shrinkage of the mix can be seen graphically in Figure 4.2.4 and is summarized in Table 4.2-4. The values obtained are close to the expected values provided by the concrete plant (around -470 ε at day 56).

Free Shrinkage of Concrete 0 0 20 40 60 80 100

-100

-200

-300

Strain (Microstrain) Strain -400

-500

-600 Time (Days)

Figure 4.2.4. Free Shrinkage of Concrete Mix over Time

Table 4.2-4. Free Shrinkage of Concrete Mix over Time

Free Shrinkage TIME (Days) ε(ε ) 14 0 15 -57 16 -117 17 -157 18 -203 35 -380 42 -403 49 -430 56 -443 91 -507

4.3 LABORATORY TEST RESULTS

4.3.1 Shrinkage Rings

The shrinkage ring specimens were monitored for 91 days. Throughout that period visual inspections were performed every two to three days and any crack observations were recorded. Cracking was typically found to begin along the outer circumference of the ring before penetrating the top or bottom. Each sample was divided into six regions based on location of the VWSGs along the top of the rings. Crack history was monitored by recording date of cracking, crack length and crack width. The final crack mappings for sample 1 can be seen in Figures 4.3.1 and 4.3.2. The FSGs and

VWSGs applied to the concrete specimens proved unreliable and all findings were based on visual analysis.

Figure 4.3.1. Final Crack Mapping of Ring Specimen 1 (Side Profile)

Figure 4.3.2. Final Crack Mapping of Ring Specimen 1 (Top/Bottom Profile)

Of importance to note are the cracking widths of the observations. While there were a few cracks measuring 0.002mm and one measuring 0.003mm the majority were less than 0.001mm. As a result accurate measurements could not be given due to the limits of the cracking microscope scale (the smallest possible measurement was

0.001mm). Nearly all cracks were difficult to see with the human eye and the microscope was needed extensively to make observations.

The observations of ring specimen 1 show forming initial cracking to occur on day 40 with a crack width of less than 0.001mm (side profile, VWSG region 4, (5/24)).

The largest crack formation was noted on day 84 with a cracking width of 0.003mm (top profile, VWSG region 5, (7/7)). This was one of only two cracks to have penetrated mostly or entirely through the ring to the steel core.

The final crack mapping for sample 2 can be seen in Figures 4.3.3 and 4.3.4.

Figure 4.3.3. Final Crack Mapping of Ring Specimen 2 (Side Profile)

Figure 4.3.4. Final Crack Mapping of Ring Specimen 2 (Top/Bottom Profile)

Similarly, first cracking was observed around day 40 with a crack width of 0.002mm

(side profile, VWSG region 5, (5/24)). The largest crack occurred on day 76 with a crack width of 0.004mm (side profile, VWSG region 5, (6/29)). Three cracks reached the steel inner core all with crack widths of 0.002mm and each occurred on day 105.

As mentioned previously, while the crack mappings clearly show the details of the cracking the specimens looked nearly flawless to the human eye. To provide a better understanding of the distribution of cracking widths Table 4.3-1 is provided.

Table 4.3-1. Crack Width Distribution Over All Samples

Crack Width Distribution Over All Samples Based on 156 Observed Crack Widths Crack Width Total No. % < 0.001 mm 64 41.03 0.001 mm 74 47.44 0.002 mm 16 10.26 0.003 mm 1 0.64 0.004 mm 1 0.64

As seen in the distribution, 88.47% of all crack widths were 0.001mm or less and only

1.92% of all cracks reached the steel ring. This statistic expresses the difficulty it was to view most of the cracking. The smooth surface of the samples provided easier observation than the more porous surfaced concrete experienced in the field.

4.4 FIELD TEST RESULTS

4.4.1 Field Strains

To assess the accuracy of the AASHTO ring test it was important to analyze the field data both during the early age as well as longer term. The early age period considered in this study was defined as the first 36 hours beginning from the start of the pour. The concrete strain observed along the deck at mid-span can be seen graphically in

Figures 4.4.1 – 4.4.4. The earliest cracking strain was determined from laboratory testing at three days. The cracking strain curve presented on the following figures was interpolated from this data.

Figure 4.4.1. Longitudinal Mid-span Strains at Top of Deck During Early Age of Concrete

Figure 4.4.2. Transverse Mid-span Strains at Top of Deck During Early Age of Concrete

Figure 4.4.3. Longitudinal Mid-span Strains at Bottom of Deck During Early Age of Concrete

Figure 4.4.4. Transverse Mid-span Strains at Bottom of Deck During Early Age of Concrete

As can be seen in Figure 4.4.1 – 4.4.4 all strains present along mid-span were compressive and remained so immediately after the pour was completed (at 6-8 hours) as well as during the following 24 hours. The spike in strain during the first hour in Figures

4.4.1 and 4.4.2 can be explained by the workers and equipment moving along mid-span while preparing and pouring the deck.

Similarly, the transverse strains were compressive throughout the pour as well as the rest of the 36-hour early age period. These strains have been plotted in Figure 4.4.5 and 4.4.6.

Figure 4.4.5. Longitudinal Quarter-span Strains at Top of Deck During Early Age of Concrete

Figure 4.4.6. Transverse Quarter-span Strains at Top of Deck During Early Age of Concrete

Apart from the early age period it is also important to oversee the longer term after the concrete has had enough time to cure. Previous researchers have found that

HPC mixes typically exhibit long time to cracking (over 56 days). To analyze this the longer term for this study was defined as a 91-day period beginning at the end of curing.

The concrete strains present at mid-span throughout this period are plotted in Figures

4.4.7-4.4.10. Dotted sections of the plots are a result of a faulty battery which caused lost data for that duration of time. Those dotted sections are simply linear connections between the last point before data loss and the first point after a new battery was replaced.

Figure 4.4.7. Transverse Strains Along Top of Mid-span During Long-Term Period

Figure 4.4.8. Transverse Strains Along Top of Mid-span During Long-Term Period

Figure 4.4.9. Longitudinal Strains Along Bottom of Mid-span During Long-Term Period

Figure 4.4.10. Transverse Strains Along Bottom of Mid-span During Long-Term Period

Review of the previous figures shows that all strains along the mid-span of the instrumented region of the bridge deck remained in compression from the beginning of the pour through the 91-day Long-Term period.

Figures 4.4.11 and 4.4.12 graphically show the transverse strains during the Long-

Term period.

Figure 4.4.11. Longitudinal Strains Along Top of Quarter-span During Long-Term Period

Figure 4.4.12. Longitudinal Strains Along Top of Quarter-span During Long-Term Period

As with the mid-span the quarter-span strains remained largely in compression throughout the Long-Term period. The spike in transverse strain seen in Figure 4.4.12, while not more than cracking strain, is likely the result of adjacent traffic having moved over the section of the bridge open to traffic while readings were being taken. This possibility is discussed in further detail in the following section on bridge deck cracking.

4.4.2 Bridge Deck Cracking

Several trips were made to the field throughout the duration of this study to monitor any cracking on the surface of the bridge deck. It was important to detail crack locations and widths for correlation with the restrained shrinkage ring data acquired in the lab. The final crack mapping of the bridge deck (Figure 4.4.13 and Table 4.4-1) details the locations of cracks as well as sensor locations. The first cracks (cracks A, B and H) were observed at day 29. The largest crack (H) began near mid-span at the border of Stage I and extended close to 8 feet into Stage II. Crack A formed eight feet from the west abutment, beginning roughly four feet from the border of Stage I and extending 24 inches. Crack B formed a few inches from the end of crack A and had a cracking length of 39 inches.

Figure 4.4.13. Crack Mapping of Bridge Deck with Sensor Locations

Table 4.4-1. Crack Map Details

Crack Details Crack Crack Length Crack Width Letter (in) (mm) A 24 0.002 B 39 0.002 C 71 0.003 D 16 0.002 E 21 0.002 F 20 0.002 G 42 0.002 H 96 0.004

Cracks C, E and F were observed on day 37 and were located near the quarter- span of the bridge deck. Crack C formed seventeen feet from the west abutment and had a cracking length of 71 inches. Cracks E and F were observed twenty-seven feet from the west abutment and were found to have cracking lengths of 21 and 20 inches, respectively.

Cracks G and D were observed on day 57, the final day of observation possible.

Crack G was located thirty-four feet from the west abutment and had a cracking length of

42 inches while crack D was found to be 16 inches long and was located fifteen feet from the west abutment.

Most crack widths were 0.002mm with the only exceptions being crack C

(0.003mm) and crack H (0.004mm). Measurements were taken using a cracking microscope (Figure 4.5.14). The apparatus magnified the surface of the deck and crack widths were recorded using the metric scale engraved on the instrument lens.

Figure 4.4.14. Crack Microscope Used in Deck Crack Mapping

The deck crack map shows all cracking to have occurred on the non-instrumented half of the structure. One half remained un-instrumented because of economic concerns

80 and as a result no data for the cracked half of the deck was recorded that could be used to analyze this unusual phenomenon. To do so finite element analysis with review of the concrete mix design was employed for further investigation.

Review of the mix design yielded no additional information until the contractor was contacted in the process of the investigation. As a result of that conversation it was

lbs. 3 learned that while the mix design called for 6.6 /yd the concrete delivered on the day of

lbs. 3 the pour had used the maximum allowable 18 /yd , three times the design amount. This, would lead to a delay in the set time of the concrete which could have resulted in a much lower f’ c than expected. When traffic patterns returned to normal nine hours after completing the pour this weakened concrete would have been more susceptible to cracking. The contractor further explained that at the time normal traffic patterns opened the concrete was still wet and soft to the touch. While more time was requested for the alternative traffic pattern this was denied.

To verify the possibility the increased retarder could have delayed the set time the temperatures of the concrete at gauge locations (Figure 4.5.15) was analyzed.

Figure 4.5.15. VWSG Temperatures During Initial 36 Hour Period

Figure 4.4.15 is divided into separate regions by four dotted lines of differing colors. The purple line marks the start of the pour (beginning at the east abutment). The red line signals the end of the pour (culminating at the west abutment). The green line specifies the increase in concrete temperature associated with heat of hydration and setting. Finally, the blue line shows the time at which traffic was returned from the construction pattern (single lane, farthest from Stage II) to its normal pattern (both lanes open to alternating traffic).

As is seen in the figure the heat of hydration was delayed for roughly five hours after completing the pour and nearly thirteen hours from the beginning. As a result, despite the additional nine hour delay in returning to normal traffic patterns, the extra

82 retarder pushed the set time of the concrete back between 4-6 hours. This allowed the concrete delivered during the first half of the pour (the un-cracked, instrumented section) an average of 6 hours of set time before being subject to adjacent traffic loads. Similarly, the second half of the deck (cracked, non-instrumented section) was allowed to set for an average of only 3 hours before subjection to adjacent traffic loads.

Verification of this scenario was performed with the use of ABAQUS, a finite element program. The findings of that analysis are detailed in the following chapter.

CHAPTER V

FINITE ELEMENT MODELING

5.1 INTRODUCTION

Finite element programs are commonly used to study the cracking behavior of bridge decks. For this study the general purpose software ABAQUS was used. The program allows for definition of material behavior and properties, boundary conditions, reinforcement and bond behavior among other variables. It is critical the user have an understanding of the model elements and constraint conditions for maximum accuracy to be gained. A description of each model element used as well as the constraint/release conditions is provided in the following section.

5.1.1 Model Element Types

While the program’s libraries contain various elements beam and shell elements have been determined to be the most reliable in the analysis of bridges. Although solid element analysis is possible in ABAQUS it requires increased computer resources without any rise in accuracy of results. It was this reason that beam and shell elements were used in this research.

5.1.1.1 Beam Element

The beam element is useful when modeling stringers and plate girders as it is a one dimensional element that typically cannot deform out of its plane. This restriction allows for plane sections to remain plane throughout analysis.

5.1.1.2 Shell Element

Shell elements are most commonly used for modeling concrete bridge decks because of the thickness of deck slabs being significantly thinner than the other two dimensions of the slab. ABAQUS possesses many shell elements in its libraries though the four node shell element is the most common for bridge deck modeling. The four node shell element is a fully integrated, general purpose, finite membrane-strain shell that allows in plane bending as well as transverse shear deformation in accordance with thick- shell theory. In short, as a shell thickness is increased the thin shell behavior it displays

(as determined by the Kirchhoff-Love hypothesis) decreases. Each node has six degrees of freedom and there are four integration points for each element. An example of a four node shell element and its integration points can be seen in Figure 5.1.1.

Figure 5.1.1. Four Node Shell Element Detailed with Integration Points

5.1.1.3 Steel Reinforcement

All steel reinforcement was modeled using the rebar element. The rebar element is an ABAQUS template that allows embedment within beam or shell elements. Within a shell it is possible to orient the rebar in layers both transversely as well as longitudinally.

The rebar can only be oriented along the longitudinal length. For all analysis grade 60 steel was specified throughout the model.

5.1.1.4 Shear Studs

When modeling shear studs two element types can be selected – breakable bond or unbreakable bond. Breakable bond shear stud elements allow for failure of the shear stud to occur while unbreakable bond elements restrict the stud’s ability to fail during analysis.

5.1.1.5 Boundary Conditions

Boundary conditions within ABAQUS are idealized to simulate piers and abutments. No effect from live load was assumed (no occurrence of side-sway or settlement within the model). Pinned supports are created by restricting the x and y planes of the girders cross section while roller supports are produced by restricting only the y plane of the cross section.

5.1.1.6 Constraint and Release Elements

The ABAQUS model is merely an assembly of individual structural components

(beams, shells, studs, etc.). Until these elements are connected to form the entire bridge

86 structure analysis cannot be performed. To bind elements to one another constraint elements must be used. The most common constraint element used is known as a multi- point constraint (MPC). Beam MPCs provide rigid connections simulating a beam between two nodes and are mainly used in slab and beam elements to cause composite action. The displacement and rotation of one node is constrained to the other. Pin MPCs are used in pinned connections and are typically reserved for stringers and floor beams as well as other non-composite connections.

Some components within the model require sharing nodes such as connecting diaphragms to stringers. In this case rotation of the starting and ending nodes (of the diaphragm) must be released with the use of a release element.

5.1.2 Material Properties

GSP overpasses typically employ the use of concrete and steel. For accurate results the properties of these materials must be determined and input into the program.

Steel has been used for decades and as a result accepted values can be used for the elastic modulus (29,000 ksi) as well as Poisson’s ratio (0.30). To perform plastic analysis the yield strength and ultimate strength of the steel must be known. Therefore a full stress-strain curve of the steel used is needed for acceptable accuracy.

For this study two types of steel were used – structural steel and reinforcing steel.

For the structural steel A36 carbon steel and A572 high strength, low alloy carbon steel was used for the girders and diaphragms. A36 steel typically exhibits a minimum yield strength of 36,000 psi and an ultimate of between 58,000 and 80,000 psi. A572 steel possesses a minimum yield strength of 50,000 psi and an ultimate of between 70,000 psi

87 and 100,000 psi. Typical stress-strain curves for the two types of steel used in FE analysis can be seen in Figure 5.1.2.

Figure 5.1.2. Typical stress-strain curves of structural steel (Salmon and Johnson, 1997)

Unlike steel, the strength of concrete can range from 100 psi to 35,000 psi.

Typically in bridge decks the design strength falls between 4,000 and 6,000 psi with HPC exhibiting a higher compressive strengths and lower water to cement ratios that normal concrete. For this reason the deck slab was considered to have a compressive strength of

5,000 psi.

The elastic modulus of concrete varies due to several factors such as aggregate size, properties of cement, and compressive strength among others. Under the ACI

Building Code (ACI 318 Article 8.5.1, 2005) the elastic modulus for concrete can be taken as:

1.5 [Eq. 5.1]

lb 3 where: = Unit weight of concrete ( /ft ) lb 2 = Compressive strength of concrete ( /ft )

The tensile strength of concrete can be approximated as 10-20% of the compressive strength. If the element is subject to bending, as opposed to tension alone, the modulus of rupture should be used in analysis instead. The modulus of rupture can be estimated under ACI 318 Article 9.5.2.3 and AASHTO LRFD Article 5.4.2.6, both of which specify (for normal weight concrete):

[Eq. 5.2] 7.5

lb 2 where: = Compressive strength of concrete ( /ft )

5.2 FINITE ELEMENT ANALYSIS RESULTS

5.2.1 Bridge Deck Analysis

Before any analysis it was required the model be verified with results from field testing. Strain data was collected throughout the day of the pour to do so. Weigh-in- motion testing was not possible due to the construction schedule limitations and as a result no truck weights were recorded in the field. To determine an accurate truck weight for analysis the model was run with a typical HS20-44 vehicle loading and strain results from the model were compared with the field data. Initial comparison showed the HS20 loading was too high and the axle weights were adjusted accordingly until the model

89 strain agreed reasonably with field strains. A comparison of these values can be seen in

Table 5.2-1.

Table 5.2-1. Strain Validation

Steel Strain Comparison Girder 3 Girder 4 Girder 5 FE Model Strain 0.000075 0.00006 0.000041 Field Strain 0.000078 0.000066 0.000044

Additionally, the model required a minimum of three regions: the older concrete deck open to traffic, the first half of the new deck and the second half of the new deck.

The new deck required separate regions in order to attribute different properties to account for the disparity in set time. It was determined that a higher accuracy could be obtained if the new concrete were to be split into six regions (as opposed to two) in order to simulate the gradient of concrete properties that resulted from the delay in set time.

The model layout can be seen in figure 5.2.1 (pre-existing concrete is green, new concrete is multi-colored).

Figure 5.2.1. ABAQUS FE Model Layout

Concrete properties for the newer concrete sections were interpolated from the laboratory data and are presented in Table 5.2-2.

Table 5.2-2. Concrete Properties for New Concrete Sections

Region Color Age at Traffic Opening (Hours) f' c (psi) E (psi) fR (psi) 1 Red 6.00 386.18 472421.67 38.62 2 Blue 5.50 353.99 433053.20 35.40 3 Yellow 4.58 294.99 360877.66 29.50 4 Purple 3.82 245.83 300731.39 24.58 5 Orange 3.18 204.86 250609.49 20.49 6 Pink 3.00 193.09 236210.83 19.31

Eight loading scenarios were applied to the model in order to analyze the various live load strains the deck may have endured. These eight cases can be seen in Figure

5.2.2. Cracking strain was assumed to be one-third of the calculated allowable tensile strain based on the principle that one-third of tensile strain comes from live load, one- third from shrinkage and one-third from temperature change.

Figure 5.2.2. Loading Cases Considered in FE Analysis

There is a lack of substantial literature on the strength of concrete at very early ages and as a result the early compressive strength was determined through the interpolation of laboratory testing. The tensile strength used in analysis was considered to be 10% of the respective compressive strength at the age in consideration. Analysis was performed for each loading scenario and the results were compared with the cracking strain to determine the effect of adjacent traffic on the new concrete. To better understand the FE diagrams presented in this chapter Figure 5.2.2 has been included to detail the orientation of Stage I and Stage II construction within the FE results.

Figure 5.2.3. FE Diagram Index (Stage I Concrete Not Outlined, First Section of Stage II Poured Outlined in Orange, Second Section Poured Outlined in Pink)

The strain results at each crack location experienced with Case 1, in which a truck passes in the outermost lane, are presented in Table 5.2-3 and can be seen visually in Figure

5.2.4.

Table 5.2-3. FE Model Strains Resulting from Case 1 Loading

Case 2

Crack ε MAX A 1.09E-05 B 1.01E-06 C 2.25E-06 D 2.46E-06 E 2.77E-06 F 2.41E-06 G 2.48E-06 H 3.92E-05

ε Cracking 2.82E-05

ε Plastic Limit 1.09E-05

Figure 5.2.4. Deck Strains Due to Case 1 Live Load

Case 1 loading could only have occurred ten hours or more after the pour concluded. As detailed previously, in Figure 4.4.16, this allowed the fresh concrete to set for 3-6 hours depending on the region under consideration.

Review of Figure 5.2.4 shows little strain at the locations of cracks A – G (crack distribution shown previously in Figure 4.4.13). The strain concentrated at mid span due to the loading exceeds that of the allowable plastic limit (considered to be 40% of f’ c).

While this may not have resulted in immediate cracking of the deck the plastic deformations that occurred during curing likely led to the eventual cracking observed at mid-span (Crack H).

Case 2 was the traffic pattern experienced from the beginning of the pour until the reopening of normal traffic patterns. These results suggest altering traffic to this specific pattern during construction was adequate in reducing the potential for early age cracking due to adjacent traffic though a further reduction in strain is desirable. The strain results

94 at each crack location experienced with Case 2, in which a truck passes in the outermost lane, are presented in Table 5.2-4 and can be seen visually in Figure 5.2.5.

Table 5.2-4. FE Model Strains Resulting from Case 2 Loading

Case 1

Crack ε MAX A 2.00E-05 B 1.66E-05 C 9.75E-06 D 3.385-06 E 8.28E-06 F 8.76E-06 G 4.45E-06 H 2.28E-08

ε Cracking 2.82E-05

ε Plastic Limit 1.09E-05

Figure 5.2.5. Deck Strains Due to Case 2 Live Load

Of importance to note is that although the model strains did not exceed plastic or cracking limits the stress distribution is in relative agreement with the crack locations.

The crack mapping shown in Figure 4.4.13 has been aligned with Figure 5.2.5 to detail this observation (Figure 5.2.6).

Figure 5.2.6. Comparison of Stress Distribution to Crack Orientation (Orientation Point Included (Arrows))

As can be seen in Figure 5.2.6, the strain distribution begins at the right corner of the newest concrete near the abutment. Similarly, all cracks but that located at mid-span propagate diagonally outward from the abutment. This observation suggests the cracking pattern is related to the strains resulting from adjacent traffic.

Case 3 considered the passing of two trucks, one in each lane, at mid-span. This loading scenario could only have occurred following the return of traffic to normal patterns. Strain results at each crack location occurring as a result of a Case 2 loading are presented in Table 5.2-5 and can be seen visually in Figure 5.2.7.

Table 5.2-5. FE Model Strains Resulting from Case 3 Loading

Case 3

Crack ε MAX A 2.55E-05 B 1.75E-05 C 1.06E-05 D 4.98E-06 E 1.01E-05 F 1.01E-05 G 6.26E-06 H 2.50E-06

ε Cracking 2.82E-05

ε Plastic Limit 1.09E-05

Figure 5.2.7. Deck Strains Due to Case 3 Live Load

While the strains produced by a Case 3 live loading generally follow the distributions seen in Cases 1 and 2, no locations of observed cracking experience strains in excess of allowable plastic and cracking limits.

Case 4 analyzed two trucks traveling in the lane furthest from Stage II construction. The results of this analysis are presented in Table 5.2-6 and can be seen visually in Figure 5.2.8.

Table 5.2-6. FE Model Strains from Case 4 Loading

Case 4

Crack ε MAX A 3.26E-05 B 2.69E-05 C 1.60E-05 D 5.55E-06 E 1.37E-05 F 1.45E-05 G 7.35E-06 H 2.28E-08

ε Cracking 2.82E-05

ε Plastic Limit 1.09E-05

Figure 5.2.8. Deck Strains Due to Case 4 Live Load

Strain distributions from a Case 4 loading are similar to those seen in Case 2, the only difference being higher strain values because of the additional live load.

Case 5 considered two vehicles in series traveling along the lane closest to the freshly poured concrete. Strain distribution was again similar to a single truck loading with higher strain values at mid-span due to the increased load. Strain values at each crack location occurring as a result of a Case 5 loading are presented in Table 5.2-7 and can be seen visually in Figure 5.2.9.

Table 5.2-7. FE Model Strains from Case 4 Loading

Case 5

Crack ε MAX A 1.60E-05 B 1.47E-06 C 3.44E-06 D 3.91E-06 E 4.41E-06 F 3.81E-06 G 3.97E-06 H 6.71E-05

ε Cracking 2.82E-05

ε Plastic Limit 1.09E-05

Figure 5.2.9. Deck Strains Due to Case 5 Live Load

The results of the Case 5 loading show strain at mid-span clearly in excess of the plastic limit and approaching the tensile limit of the deck. This loading was observed on several occasions during field visits and provides an explanation for the largest crack recorded in the field analysis. While Cases 1 – 5 do not provide strains over allowable limits for all crack locations the strain distributions provides strong evidence in support of adjacent live loads being the cause of the observed cracking. The FE analysis did not include any lane load and the presence of additional passenger vehicles would further increase the strains presented.

Cases 6 – 8 were not regularly observed during field observations and are considered rare cases. Strain values and distributions are similar to previous cases and are provided for cases 6, 7 and 8 in Tables 5.2-8, 5.2-9 and 5.2-10, respectively. Visual analysis for those cases can also been seen in Figures 5.2.10 – 5.2.12.

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Table 5.2-8. FE Model Strains from Case 6 Loading

Case 6

Crack ε MAX A 4.19E-05 B 2.78E-05 C 1.82E-05 D 8.60E-06 E 1.73E-05 F 1.74E-05 G 1.07E-05 H 1.23E-05

ε Cracking 2.82E-05

ε Plastic Limit 1.09E-05

Figure 5.2.10. Deck Strains Due to Case 6 Live Load

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Table 5.2-9. FE Model Strains from Case 7 Loading

Case 7

Crack ε MAX A 3.86E-05 B 2.78E-05 C 1.70E-05 D 7.17E-06 E 1.56E-05 F 1.60E-05 G 9.20E-06 H 2.28E-08

ε Cracking 2.82E-05

ε Plastic Limit 1.09E-05

Figure 5.2.11. Deck Strains Due to Case 7 Live Load

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Table 5.2-9. FE Model Strains from Case 7 Loading

Case 8

Crack ε MAX A 2.93E-05 B 1.76E-05 C 1.17E-05 D 6.48E-06 E 1.17E-05 F 1.16E-05 G 7.77E-06 H 3.05E-05

ε Cracking 2.82E-05

ε Plastic Limit 1.09E-05

Figure 5.2.11. Deck Strains Due to Case 7 Live Load

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5.5.2 FE Analysis Conclusion

After completing the analysis the cause of cracking on only one half of the deck became clearer. The observations of the contractor with the FE results point towards the extended delay time as the prime factor in the deck cracking. The crack orientations line up almost identically with the stress distributions of vehicles passing in the far lane.

Vehicles passing in the nearest lane were determined to generate concentrated strains at the mid-span boundary of the new and pre-existing concrete – precisely the location of the longest crack. These findings along with the lack of tensile strains from the field instruments suggest the cracking observed on the bridge deck was a result of the excessive retarder used in the mix. The return of traffic to normal patterns prior to adequate concrete set time likely led to crack formation as well. A follow-up with the contractor provided further support for this conclusion as well. The northbound interchange 67 overpass, identical in construction to the bridge focused on in this study, was poured two weeks later with concrete possessing one third the retarder of the southbound bridge. After an identical two-week curing period the northbound bridge had not exhibited visible cracking of any kind during the first four weeks after pouring.

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CHAPTER VI

SUMMARY AND CONCLUSIONS

6.1 SUMMARY AND CONCLUSION

The purpose of this study was to discover the accuracy of the AASHTO Passive

Ring Test’s ability to simulate the shrinkage cracking mechanics of HPC bridge decks, commonly used by the New Jersey Department of Transportation (NJDOT). A high performance concrete bridge deck was instrumented with various sensors to monitor the structure throughout the construction process and furthermore once the bridge began its service life. Samples were taken during the pour which included two AASHTO PP 34-06

Passive Ring Specimens. These samples were compared with the information obtained from the field instrumentation and incorporated with a finite element analysis to make the following conclusions:

(a) The AASHTO PP 34-06 Passive Ring Test simulated the field cracking due to

shrinkage reasonably well as 98% of all ring sample cracking was determined

to be micro-cracking. Field surveys of the bridge deck showed no signs of

cracking credited to restrained shrinkage (micro-cracks were unobservable on

the deck due to the coarse surface of the deck itself).

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(b) The cracking noted on the bridge was most likely the result of three times the

design amount of retarder being incorporated in the mix that was delivered to

the site. This led to a delay in time of setting of roughly 6 hours or more after

completion of the pour. This delay led to weaker concrete than expected once

traffic patterns were returned to normal.

retarder (and therefore weaker concrete at the reopening of normal traffic

patterns) led to concrete strains in the deck more than the allowable plastic

limit. Deformations from adjacent live load during setting the deck likely led

to the eventual appearance of cracking shortly after the completion of

construction.

6.2 SCOPE OF FUTURE WORK

Future research is needed to research the ability of the AASHTO Passive Ring

Test to accurately simulate the HPC deck cracking mechanics of continuous span bridge decks. Replication of this study with stricter regulations on mix design parameters would likely provide further insight into the accuracy of the test as well.

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