SPECIAL REPORT

Lessons Learned from the Kobe Earthquake A Japanese Perspective

Hiroshi Muguruma This report presents an overview of the Ph.D. performance of reinforced and precast, Professor Emeritus Department of prestressed concrete buildings during the Architectural Engineering Kyoto University Hyogoken-Nanbu earthquake (also known as Kyoto, Japan the Great Hanshin earthquake) of January 17, 1995, situated in and around the city of Kobe, Japan. The performance of pile foundations is also examined. Highway bridges, rapid transit structures, and other special structures are covered elsewhere. The assessment of damage is related to the Minehiro Nishiyama evolution of design code provisions for Ph.D. concrete building structures in Japan. Department of Preliminary reports indicate that precast, Architectural Engineering Kyoto University prestressed concrete structures performed Kyoto, Japan remarkably well during the earthquake, especially those designed with recent seismic code provisions. The probable causes of the damage are examined, although it should be emphasized that several investigations are currently being carried out to determine more comprehensive causes of structural failures Fumio Watanabe, Ph.D. by many researchers, engineers, the Professor Architectural Institute of Japan (AIJ), the Department of Architectural Engineering Japan Prestressed Concrete Engineering Kyoto University Kyoto, Japan Association (JPCEA), and other organizations.

28 PCI JOURNAL t precisely 5:46 a.m. in the N early morning of January 17, A 1995, a devastating earthquake struck Japan, imparting a trail of de­ ® ~ Severely damaged area struction across a narrow band extend­ ing from northern Awaji Island through the cities of Kobe, Ashiya, Nishinomiya and Takarazuka (see Fig. 1). The 7.2 Richter magnitude registered was one of the strongest earthquakes ever recorded in Japan. Initially, Sumoto City on Awaji Is­ land and Kobe City were assigned a Shindo 6 intensity. However, later the Japan Meteorological Agency (JMA) revised the Shindo intensity level from 6 to 7 for parts of the cities of Kobe, (1 gal= 1 em/sis) Ashiya, Nishinomiya and Takarazuka, J. .I and parts of northern Awaji Island. Skm The Shindo intensities 1 to 7 corre­ spond to the Modified Mercali Inten­ sity Scale of I to II, II to IV, IV to V, Fig. 1. Area map of severe earthquake damage and recorded accelerations. V to VII, VII to Vill, Vill to IX, and IX to XII, respectively. The Hyogoken-Nanbu earthquake GROUND MOTIONS (also called the Great Hanshin earth­ The focal depth of the earthquake quake) will hereafter be referred to as was approximately 14 km (8.6 miles). the Kobe earthquake. The epicenter was located at 34 o 36.4' The earthquake resulted in 5502 north latitude, 135° 2.6' east longitude. deaths. More than 24,000 people were Three faults are believed to have rup­ injured in the Hyogo Prefecture alone. tured during the main shock. A hori­ As of June l, about 40,000 people still zontal displacement of up to 1.6 m live in temporary shelters. The esti­ (5.25 ft) was found at the Nojima Fault mated property damage ranges from on Awaji Island. The severely dam­ $95 to $140 billion. aged area consists of a narrow band The earthquake caused significant from the northern part of Awaji Island damage not only to old buildings de­ to the city of Takarazuka. The cities of Period (sec) signed according to former design Kobe, Ashiya, and Nishinomiya are in­ codes but also to modern buildings cluded in this region (see Fig. 1). Fig. 2. Velocity spectrum of that conformed to current design codes The Preliminary Reconnaissance earthquake recorded by Kobe Maritime Meteorological Observatory. and regulations. The performance of Report of the 1995 Hyogoken-Nanbu building structures during the earth­ Earthquake published by the Archi­ quake has been studied by numerous tectural Institute of Japan (AIJ)' states around 0.25 to 0.4 seconds. researchers, engineers, and organiza­ that the characteristics of the ground 5. Ground motions were affected by tions. At this time, several clues to the motions recorded may be summarized local soil conditions and topography. causes of the devastating damage have as follows: The preliminary reconnaissance report 5 been found. '· 1. Peak ground accelerations were of AIJ states that ground motions were In this report, the damage to rein­ large in both the horizontal and verti­ most likely amplified in the plains forced concrete buildings and precast, cal directions. The peak accelerations near the mountains between the cities prestressed concrete buildings is as­ observed at several sites are summa­ of Kobe and Nishinomiya. sessed. Several possible causes for typi­ rized in Table 1 and in Fig. L. cal damage are presented. Also, the 2. Fig. 2 shows the velocity spec­ performance of precast, prestressed trum recorded by the Kobe Maritime GEOLOGICAL ASPECTS concrete piles is discussed. This report Meteorological Observatory. Because the geological aspects of is limited to building structures. Special 3. The duration of strong shaking the region are described elsewhere by structures, such as highway bridges and was 10 to 15 seconds. experts in the field, this section will rapid transit structures, are covered 4. The predominant period was 0.8 only briefly mention the highlights of elsewhere by other researchers or engi­ to 1.5 seconds. A second predominant the AIJ report. 1 The area of Shindo 7 neers in the civil engineering field. period was, at times, observed to be intensity is approximately 20 km (12.3

July-August 1995 29 Table 1. Peak accelerations and soil conditions (Ref. 1).

Peak ground acceleration (gal)

Soil Measured North- East- Up- Recorded point Location condition level south west down

JMA- Kobe Chou Ward, Kobe City Diluvial IF 818 617 332 JMA - Osaka Chou Ward, Osaka City Oil uvial B3F 81 66 65 - ~ Rock foundation 272 265 232 MTRC Kita Ward, Kobe City - Rock GL-15 m 208 213 11 6 - A Building Chuo Ward, Kobe City Diluvial B3F 223 208 292 B Building Kita Ward, Osaka City Alluvial GL ' 1:1 182 267 302 C Building Kita Ward, Osaka City Diluvial B4F 155 157 193 -- Oil uvi al GL >52 49 46 SiteT Minamikawachi District Diluvial GL-IOOm 23 - 16 -- Diluvial GL .. 43 50 49 Site Y Kita Ward, Osaka City - Oil uvial GL-60m 24 49 I - - Obayashi building Chou Ward, Osaka City Diluvial B2F X: 139 Y: 87 Z: 210

• c, M apartment Miyakojima Ward Alluvial IF X: 60 Y: 86 Z:42 house Osaka City p Osaka mechanical Taisyo Ward '"""· Alluvial GL 195 140 122 material center Osaka City ~ .. ~ I Abiko apartment Sumiyoshi Ward f, Alluvial IF I· 107 115 92 house Osaka City ' Takami Tall Konohana Ward Alluvial IF 156 178 176 • residence Osaka City Alluvial GL .. · 222 267 255 Fill-in GL f3 11 7 85 53 Takatsuki Campus Reizenjicho ground ~ of Kansai University Takatsuki City Sandstone GL 67 61 36 Sandstone GL-13 m 66 49 39 ~ - Abeno Ward I Point A Oil uvial GL-3 m 76 - 26 Osaka City I Alluvial IF 129 I 103 91 1: PointD Asahi Ward, Osaka City Alluvial GL 189 155 126 ,,, Oil uvial GL-25 m 129 113 8 1 ·~· ·~ Note: I m = 3.28 ft; I gal = I em/sis; I em = 0.39 m. miles) long and reaches from Kobe to Mount Rokko consists primarily of earthquake in 1923. The seismic de­ Nishinorniya. Mount Rokko lies north granite and is crossed by many faults. sign procedures in Japan have been re­ of Kobe, extending in an east-west di­ The southern side of the mountain has vised every time a significant earth­ rection. The plains are within a narrow step-like slopes, consistent with down­ quake occurs and causes severe band of land between Osaka Bay and ward displacements at the fault scarps damage. The evolution of the seismic the mountains. relative to the north. Near the ground design codes is described below. surface, the granite has weathered into decomposed granite soil. A simplified Historical Review of 1. SOkrn 0.35krn profile of the ground cross section in the north-south direction in Kobe City Seismic Design Provisions is illustrated in Fig. 3. for Reinforced Concrete Buildings in Japan EVOLUTION OF In the seismic provisions of the Building Standards Law of 1950, the DESIGN CODES seismic design load applied to each The damage to buildings caused by floor, Vj, was calculated by the follow­ the Kobe earthquake is closely related ing equation: to the design methods adopted. The Vj = [0.2 + O.O l(H; - 16)/4]w; (1) Fig. 3. Simplified profile of ground first seismic design provisions were cross section in Kobe City. established just after the Great Kanto where

30 PCI JOURNAL ...,._ 0.26 ...,. 0.25 Subsoil II/ (Flexible) 1(; =0.8 0.24 rf: ...,. ....: ...,. 0.23 c: 0.22 - ~ ...,. ~ 0.21 Q) ...,. 0 0.20 (.) ~ E (3

July-August 1995 31 where C is the standard base shear ~ b· 0 ~ h4 (J . =_L coefficient and for the second phase I hj ~-l. design C = 1.0. The term Ds is a re­ 62 'h 0 '2- 1 duction factor that depends on the c, h2 (Jj type and the ductility of the structure. ~- Rsj The factor Ds ranges from 0.3 for duc­ h1 -I-1[" 1] tile frames to 0.55 for buildings in n i=l8i which a large portion of the lateral load is assigned to the walls and Note: braces. This factor is primarily based 8j Interstory drift of j-th story under seismic design load on the equal energy concept in which of the first phase design. the energy absorbed by a building that bj Interstory displacement. yields with elasto-plastic characteris­ hj Story height. tics is assumed to be equal to that of a R

Table 2. Coefficients F, and F with regard to the eccentricity ratio Re and 5 where F, is a coefficient that is related stiffness ratio R , respectively. 5 to the eccentricity ratio Re in each R, F, Rs F. story and ranges between 1.0 and 1.5 s; 0.15 1.0 ~0.60 1.0 (see Fig. 6). The term F, is a coeffi­ 0.1 5-0.30 Linear interpolation 0.30-0.60 Linear interpolation cient that is dependent on the stiffness ratio Rs in each story and ranges be­ ~0.30 1.5 s; 0.30 1.5 tween 1.0 and 1.5 (see Fig. 7). Table 2 summarizes these coefficients, F,, F, and F.s· Therefore, F.s varies between shear force Q; given by Eq. (2). The soils, respectively. The coefficient R 1 1.0 and 2.25. The factor was intro­ interstory drift of each layer obtained ranges between 1.0 and 0.25 and is ex­ F.s duced to provide an extra strength in by linear elastic analysis under the pressed schematically in Fig. 5. The the case of buildings with unsymmet­ above load combination shall be less term A; is the distribution factor of lat­ rical arrangements of the seismic load than or equal to 1/z oo: eral shear forces along the height of resisting elements and/or with ex­ the building and is given by Eq. (4): tremely flexible stories, compared to the other stories. 1 where C is the basic seismic coeffi­ A=1 1 + [---a1- ]_I!___ (4) 0 {a; 1+3T cient of 0.2. The symbol Z denotes the Historical Review of seismic hazard zoning coefficient and where a; is the ratio of the gravity Seismic Design Provisions for varies between 0.8 and 1.0. The sec­ load above the ith layer to the total Prestressed Concrete Buildings ond symbol, R1, is the design spectral gravity load above the level of im­ in Japan coefficient, which depends on the sub­ posed lateral ground restraint. soil profile and the natural period of The design procedure for pre­ stressed concrete structures was first vibration of a building and is given by Second Phase Design the following equation: issued in 1960 when the Standard for The lateral load resistance in each Structural Design and Construction story is calculated using inelastic of Prestressed Concrete Structures T

32 PCl JOURNAL stressed concrete buildings are not prestressed. For example, columns are usually not prestressed. They may be designed according to the codes for reinforced concrete. When the design procedure was first introduced, the design seismic load applied to each floor was calculated using Eq. (1). The maximum height was four stories or 16 m (52 ft) . The basic seismic coefficient, therefore, was 0.2 regardless of the height of the building. The load combination was ------. 1.2(D + L) + 1.5£. In 1973, the height limitation for No prestressed concrete building struc­ tures was extended to 31 m (102ft), which was the same for buildings of other structural types. In addition, the No design seismic load applied to each floor was calculated using Eq. (1). The load combination was D + L + 1.5£ for flexure by the strength design method and D + L + 2.0£ for shear by the allowable stress design f!Jethod. The design stresses were calculated by linear elastic analysis. The method of calculating the ulti­ mate shear strength of prestressed Special approval by and non-prestressed members had Minister of not been established at that time. Consttuction Therefore, a relatively large design load combination was specified and the allowable stress design method was used. Reinforced concrete mem­ bers should be designed to fail in ~ flexure. The maximum spacing of Route CD Route ® 1,2,3 Route 3a Route 3b transverse reinforcement was 10 em Fig. 8. Flowchart for design of prestressed concrete building structures. (3 .9 in.) at the ends of columns and 15 em (5 .9 in.) elsewhere. After the 1981 code revisions came into force, prestressed con­ crete buildings could be designed according to either the pre-1981 de­ D Slight or no damage sign method described above or the 1982- [] Minor damage new seismic design procedure aimed "8 h======:_-l [] Moderate damage primarily for conventionally rein­ - ~ forced concrete buildings (see Fig. c: mn Collapse or severe :5 72-81 W!l damage 8). The revised seismic design load distribution and intensity given by ~c: Eqs. (4) and (5), respectively, are 8 used . However, buildings higher -1971 than 31 m (102 ft) and lower than or equal to 60 m ( 197 ft) must be 0 5 designed according to the latter de­ Number of buildings sign method. In addition, the design of concrete buildings higher than Fig . 9. Damage level of reinforced concrete structures with 31 m ( 102 ft) requires approval by respect to the year of construction (Ref. 1). the Minister of Construction.

July-August 1995 33 Fig. 10. A pre-1981 apartment building that collapsed at the Fig. 11. A pre-1981 apartment building that collapsed at the soft first story. soft first story.

PERFORMANCE OF REINFORCED CONCRETE BUILDINGS Damage to buildings by the earth­ quake was much more severe in build­ ings built before the 1971 code revi­ sion took effect. The investigation conducted by the AIJ Kink.i Branch re­ vealed that in the Chuo Ward of Kobe City, the center of Kobe, 18 reinforced or steel-encased reinforced concrete buildings constructed before 1971 col­ lapsed or suffered severe damage (see Fig. 9). On the other hand, only two of those buildings built between 1971 and 1981 were found collapsed or severely damaged. No concrete buildings built after the 1981 revision collapsed. Fig. 12. A post-1981 apartment building Fig. 13. A post-1981 apartment building that collapsed at the soft first story. that collapsed at the soft first story. Collapse of Soft First Story Many buildings that were con­ structed with open retail space or quirements collapsed in the open first systems were calculated based on a parking on the first floor collapsed. In story (see Figs. 12 and 13). The col­ different base shear coefficient but the old buildings designed and con­ lapse calls attention not only to a uni­ elastic stiffness of the layer was the structed before the 1981 revisions, the form distribution of story stiffness same. The A; distribution was used as collapse can be attributed to a more along the height of buildings but also a shear force distribution over the flexible and/or weaker story than the to an excessively weak story, com­ height of the systems. other ones, and inadequate transverse pared to the other stories, even if it has The envelope curve model for shear reinforcement in terms of its amount greater story shear strength than that force-interstory drift of each layer in and detailing (see Figs. 10 and 11). specified by the code. the case of the base shear coefficient

Since 1981, an excessively flexible The damage tends to concentrate of C0 = 0.35 is shown in Fig. 15 as an story, compared with the other stories into the weakest story, as shown in example. The standard El Centro NS in a building, has been restrained or Fig. 14. The figure was obtained by 1940 earthquake wave record was has been required to have extra dynamic response analyses on lumped used. It was amplified to the maxi­ strength. This was realized by the in­ multi-mass shear systems to observe mum velocity of 50 cmls (20 in. per troduction of the stiffness ratio, Rs. In how large a deformation was concen­ sec). Fig. 14 shows the maximum in­ addition, the detailing of transverse re­ trated to weakest stories when a col­ terstory drifts of the systems analyzed. inforcement has been improved. umn sidesway mechanism formed. The results include the response of: However, several buildings that The systems consisted of eight (1) the linear elastic system; (2) the conform to the current design code re- masses. The yield capacities of the systems that were designed using base

34 PCI JOURNAL shear coefficients of 0.25, 0.35 and 0.45; and (3) the systems designed ba­ sically using base shear coefficients of 0.45 and 0.35, but the story shear ca­ pacity of the second (Case 1), the 8 fourth (Case 2) or the sixth (Case 3) layer was provided from the base shear coefficient of 0.25. 6 Therefore, an interstory drift dis­ placement was expected to concen­ trate into the weakest layer. If the

shear capacity based on C0 = 0.25 is assumed to be required, the layers other than the weakest layer had re­ 2 serve strength. The ratios of the pro­ vided strength to the required strength were 1.8(0.45/0.25) and 1.4(0.35/0.25), Gr. respectively. 0 0.5 1 1.5 2 2 As expected, the interstory drift con­ lnterstory drift (10 · rad.) centrated in the weakest story. Table 3 summarizes the analytical results of the maximum interstory drift angles in l ! ! Case2 0.01 radian. The column of the weak­ 8 est story of each system was sur­ rounded by double lines and included 6 the corresponding ductility ratio. In the 1981 revisions, a stiffness factor was introduced to prevent an excessively flexible story. It is, how­ : ~ 1- · - r·· --¢- c.,=0.35 ever, based on the elastic stiffness. In r-...... ,.. // \'.,'., ~·<>, ...... , ...... -b. - C - 0.45 n1 0 order to avoid an excessively weak !-...... ~ ... , , ...... [...... - •·- C - 0.45 (4F-C - 0.25) 2 0 0 story, a distribution of story shear

- ...... jo .. ~ ...... ) ...... ---• -- · C0 - 0.35 (4F-C0 - 0.25) strength along the height of the frame should be considered. I I i I Gr. A non-ductile frame that does not 0 0.5 1 1.5 2 rely on plastic deformation of the mem­ 2 lnterstory drift (10 · rad.) bers can be designed if D, = 0.45 is used. The reduction factor D, for a duc­ tile frame is 0.3. Because the maximum value of F, is 1.5, a non-ductile open ! ! ! Case3 8 first story that conforms to the current design requirements is realized if a story shear strength of 0.675W is pro­ vided with the story in which the provi­ '- sion for a stiffness ratio is not satisfied. 0 0 The term W is the weight of the build­ LL:4 ing. However, several post-1981 build­ : . . I : - <> - c -0.3:, ings that collapsed in the first story re­ r-...... 'f 1\~]> ...... 1...... - b. - c:.. o .45 vealed that a story shear strength of that 2 ,...... p ... ~ ...... 1...... - • -- c - 0.45 (6F-c - 0.25) quantity may not be enough if one or _...... j ch .. ~! ...... +...... ---• --- c:- o.35 (6F-c:.. o.25J more columns fail in a brittle manner Gr. i i before the shear strength of each col­ 0 0.5 1 1.5 2 umn in the first story should develop. lnterstory drift (10 ·2rad.) In the first phase design of the cur­ rent design code, the story shear force­ inters tory drift curve of each layer based on the elastic stiffness should pass over the coordinates [1h oo in inter­

Fi g. 14. Dynamic response of buildings with one story weaker than the other stories. story drift angle, 0.2ZR1Aiwi in story

July-August 1995 35 transverse reinforcement of 9 to 10 mm Shear force (0.35 to 0.39 in.) diameter was pro­ vided in the spacing of 20 to 30 em (7 .8 to 11.8 in.). Due to the 1971 revisions, transverse reinforcement was required Oco=0.35 to have a spacing of 10 em (3.9 in.) or 0.004E less in the column end regions. The observed damage to the columns Oco=0.25 in the Kobe earthquake and other past earthquakes indicated that a 90-degree Oco=0.175 hook followed by relatively short ex­ tensions cannot prevent columns and walls from failing in shear. A 135-de­ gree hook of transverse reinforcement and sufficient extensions must be pro­ lnterstory drift vided, as required in the current codes. Use of closed ties and cross-ties is rec­ Fig. 15. Shear force-interstory drift envelope model. ommended, especially when the col­ umn section is large and ductility de­ mand is high. Even a 135-degree hook shear] as shown in Fig. 16. Beyond the codes no matter how large inter­ was found ineffective in some cases be­ this point, no consideration of dis­ story drift may be attained. If a story cause of spalling of cover concrete. placement is required. Each layer of shear strength of 0.675W is attained at Brittle fracture at the bent was ob­ the building is required to have a story so large a displacement that a second­ served in the transverse reinforce­ shear capacity greater than specified in order geometric effect should be con­ ment. This may be due to the poor sidered, the structure would become quality of 9 to 13 mm (0.35 to 0.51 in.) unstable, which would lead to collapse. diameter bars.

Shear Failure of Collapse of a Midheight Story Columns and Walls A conspicuous mode of failure of Numerous columns and walls were reinforced concrete buildings in the observed to fail in shear. Such failures earthquake is the story collapse at a were pointed out in past earthquakes. rnidheight story (see Figs. 20, 21 and This kind of damage can be attributed 22). Several reasons described below to short columns, insufficient shear re­ are potentially responsible for these inforcement, no cross-tie or supple­ collapses: mental ties, and inadequate construc­ 1. Unless a building structure is de­ lnterstory drift 0 11200 tion (see Figs. 17,18 and 19). signed so that a certain collapse mech­ Old buildings constructed before anism is intentionally formed, damage Fig. 16. Story shear force-interstory 1971 had relatively little transverse re­ may concentrate in any story. drift curve. inforcement in their columns. The 2. Damage can concentrate at a

Table 3. Maximum interstory drift angle (0.01 radian). Elastic I c. = 0.45 I c. = 0.35 c. =0 .45 c. =0 .35 c. =0.45 c. =0 .35

Floor response C0 = 0.25 c. = o.35 I c. = o.45 (2F - C25) (2F - C25) (4F -C25) (4F- C25) (6F - C25) (6F- C25) 8 0.849 1.532 1.298 1.11 6 0.832 1.256 1.037 1.274 1.007 1.292 7 0.920 1.248 1.11 2 1.030 0.963 1.038 0.928 1.013 0.950 1.052 1.502 0.920 6 0.909 0.716 0.724 0.787 0.79 1 0.703 0.729 0.716 I (1.81) (1 .1 09) 5 0.879 0.636 0.6 11 0.657 0.56 1 0.567 0.632 0.626 0.501 0.583 l.l75 0. 794 4 0.822 0.598 0.625 0.653 0.5 19 0.612 0.56 1 0.600 (1.4 1) (0.965) 3 0.784 0.576 0.692 0.675 0.55 1 0.622 0.6 15 0.628 0.561 0.650 1.370 2 0.758 0.535 0.722 0.775 0.649 0.69 1 0.661 0.667 I I (1.65) (1.008~ I 0.786 0.538 0.709 0.794 0.638 0.725 0.639 0.660 0.683 0.650

36 PCI JOURNAL Fig. 17. Shear failure of short columns. Fig . 18. Shear failure of walls.

story in which the story shear strength Torsional Failure Resulting and/or stiffness changes abruptly be­ From Eccentricities of tween adjacent stories. Several build­ Stiffness and Mass ings were found collapsed at the story One building had a structural wall on where the structural system changed one side of the perimeter in the first floor. from steel-encased reinforced concrete The other three sides were open. The (SRC) to reinforced concrete. In an­ building sustained damage in columns on other case, the amount of structural the open sides due to torsional response. walls in the collapsed story was found Members susceptible to larger force and to be much less than the other stories. deformation demands due to plan eccen­ 3. The seismic design load distribu­ tricity need to be designed recognizing tion over the height used in the old de­ their actual stiffness and strength proper­ sign codes is different from current ties and the impact of these properties on codes. Although the codes cannot be torsional response. Buildings should be compared directly due to differences be­ as regular as possible. tween the design procedures, the propor­ tion of design story shear was smaller at the middle stories in the old codes than Failure of Gas-Pressure Welded the current ones, as shown in Fig. 23. Reinforcement Splices Fig . 19. Shear failure of a column due 4. Large vertical accelerations may to inadequate transverse reinforcement. have generated large compressive and In Japan, reinforcement splices in tensile axial loads in the columns, buildings and bridges recently have al­ which resulted in ductility and shear most always been made by a process then fused together by heat and pressure strength reductions. The interaction of known as gas-pressure welding. In this applied by mechanical devices, causing horizontal and vertical accelerations process, the bars to be joined are the bars to flare out at the splice. Pres­ may also be a reason. aligned and butted together; the bars are sure welded splices were observed to

Fig. 21 . A hospital that collapsed at the fifth floor.

July-August 1995 37 Collapse Mechanism Several new buildings were designed such that a certain collapse mechanism, especially a beam sidesway mecha­ nism, is intentionally formed. Fig. 25 shows a ten-story apartment building constructed in 1991. Fig. 26 shows the elevation of a structural frame and the plan. Plastic hinges at the ends of the beams in the second to seventh stories were fo und as intended in the design (see Fig. 27). The residual drift at the top of the building could not be ob­ served. Because non-structural walls suffered damage, the building is sched­ uled for repair. The typical damage described above was also observed in the 1968 Toka­ cbioki earthquake and the 1978 Fig. 22. An apartment building that collapsed at the th ird story due to torsion resulting from eccentricities of stiffness and mass. Miyagikenoki earthquake. Old build­ ings designed according to the old de­ sign codes should be analyzed and fracture in the earthquake (see Fig. 24). strengthened. Investigation is needed to identify the Concrete of approximately 20 MPa causes of the fracture. (2.9 ksi) in design compressive strength is generally used in Japan. 12 r----.----~---,----~ Performance of High Rise Higher strength concrete should be re­ Reinforced Concrete Buildings quired. Higher strength concrete not only improves the seismic perfor­ No damage was found in high rise mance of buildings but also leads to reinforced concrete buildings in the re­ better and more careful construction. gion of severe ground motions due to careful design and construction and use 2 of high strength concrete. Acceleration o L___ J_~==~~~~ ~ PERFORMANCE OF responses recorded in a 31 -story rein­ 0.1 0.2 0.3 0.4 0.5 Seismic Shear Coefficient PRECAS~PRESTRESSED forced concrete building located about CONCRETE BUILDINGS 43 km (26 miles) east of the epicenter Fig. 23. Compari son of seis mic design indicated that the maximum accelera­ load between old and current design The inspection carried out by the tions in horizontal and vertical direc­ codes. Japan Prestressed Concrete Engineering tions on the 31st floor are 1.14 and 1.7 Association (JPCEA)2 reported that times those on the ground, respectively. there are 163 precast, prestressed con-

Fig. 24. Fail ure of gas-pressure welded reinforcement Fig. 25. An apartment building designed such that a beam splices. sidesway mechanism was intentionall y formed .

38 PCI JOURNAL The number includes buildings that have some precast, prestressed concrete members. Eleven of these buildings are precast, prestressed concrete buildings, 49 buildings which had non-structural precast, prestressed members, 89 cast­ in-place prestressed concrete buildings and 14 buildings which had non-struc­ tural cast-in-place prestressed mem­ bers. Buildings with unbonded tendons are excluded. Use of unbonded tendons for primary earthquake resistant mem­ bers is prohibited in Japan. Most of the precast, prestressed con­ crete buildings performed remarkably Piles I I I I I I I I well in the earthquake. The reasons 5600 5600 5800 5800 why little damage was found in pre­ cast, prestressed concrete buildings are Elevation of structural frame summarized below: 1. Seismic design loads assigned to precast, prestressed concrete buildings ~-: ~N are larger than those of buildings of _r---L_..:.-.r------~ Lr--"'-- different structural types, although the design methods are different. There­ fore, resistance to earthquake loads of precast, prestressed concrete buildings is expected to be higher than that of ~ the other types of building structures. 2. Precast, prestressed concrete buildings are generally regular struc­ tures with a symmetrical shape in plan and a uniform distribution of masses Standard floor plan and stiffening elements. 3. High strength and quality con­ Fig . 26. Structural frame and plan of ten-story reinforced concrete apartment crete is usually used, resulting in building pictured in Fig. 25. higher shear resistance and careful construction. crete buildings in the region of Shindo which is one of the regions of Shindo 4. Precast, prestressed concrete intensity of 6 and 7: Kobe, Ashiya, intensity of more than 6, there are three buildings are relatively new. Nishinomiya, Takarazuka, Itami, Ama­ prestressed concrete buildings. These Among these buildings, only three gasaki and Kawanishi. On A waji Island, buildings sustained no damage. sustained severe structural damage.

Fig. 27. Cracks at the beam-to-column interface- an .... indication of the formation of beam sidesway mechanism Fig. 28. Prestressed concrete building with extension of steel (building pictured in Fig. 25). tumbled down .

July-August 1995 39 Fig . 29. Shear failure of a column in the second floor of the Fig . 30. Damage to a column that supported a prestressed building shown in Fig. 28. concrete beam.

One building sustained architectural 1973. The building had an extension depth was 2 m (6. 6 ft). No damage damage in its precast non-structural made of steel in the front, half of was observed in the beams. Several elements. which completely tumbled down (see small cracks were fo und but th ey The most devastating damage of a Fig. 28). The beams of the fo urth story could not be identified as cracks concrete structure was found in a four­ are 37.2 m (122ft) cast-in-place pre­ caused by the earthquake. They are story bowling arena that was built in stressed concrete. The total beam considered to have been caused by the

Fig. 31. A gymnasium of an elementary school in which Fig . 32 . Seven pieces of the precast, prestressed concrete precast, prestressed concrete roof shells fell down. roof shells fell down.

Fig. 33. Damage to a column of the gymnasium shown in Fig. 31 . Fig . 34. Damage to a pile.

40 PCI JOURNAL introduction of prestressing. Almost half the inner reinforced concrete columns and several periph­ eral columns in the second story failed in shear (see Fig. 29). The column sec­ tion was 1000 x 1000 mm (39.4 x 39.4 in.) with a clear height of 4 m (13 ft). The third story had columns only on the peripheral frame. Therefore, the vertical load of the fourth story was mainly transmitted to the peripheral columns in the second story. A small column axial load resulted in reduction of shear resistance. This, as well as insufficient transverse rein­ Fig. 35. Another example of damage to a pile. forcement, may be a reason for the column failure. Several severely dam­ aged columns were observed in the fasten them. Seven pieces fell down several reinforced concrete buildings. first floor as well. The spacing of the on to the ground (see Fig. 32). In some Six years after the earthquake, new transverse reinforcement of D 13 of the cases, one end of the panel was left on seismic design provisions for founda­ columns was 150 mm (5.9 in.). the top peripheral beam. The columns tion piles were mandated by the Min­ The most typical type of damage to on the third floor failed in flexure at istry of Construction. In the code, prestressed concrete building struc­ the place where some of the longitudi­ piles are required to be designed to re­ tures was found in this building: a col­ nal reinforcing bars terminated and the sist elastically the loads from the su­ umn failure prior to yielding of the number of the bars was reduced (see perstructure as large as 0.2 W, where W prestressed concrete girders and Fig. 33). is the weight of the superstructure. beams that frame into the column. A relative movement between the In Japan, three grades of precast, Fig. 30 shows the top of a column in panel and the pillow beam is consid­ prestressed spun concrete piles are the third floor that a prestressed con­ ered to be about 100 to 150 em (39 to currently being produced with an aver­ crete beam was framed into. Immedi­ 59 in.). The failure of the reinforced age prestress of 4, 8, and 10 MPa ately below the beam, the column had concrete columns would trigger off the (580, 1160, and 1450 psi), respec­ its cover concrete spalled off. drop of the roof panels. The precast tively. Until 1984, only piles with a Prestressing tendons are usually panels were installed to absorb the dis­ prestress of 4 MPa (580 psi) were pro­ provided to cancel or reduce flexural tance change between the supports. duced. However, since the seismic moments due to dead and live loads. However, the reinforced concrete code provisions for piles were en­ This results in much more beam frame that supported the panels was so forced in 1984, piles with a prestress strength than required for the actions flexible that the movement seemed to of 8 and 10 MPa (1160 and 1450 psi) due to design seismic loads. Plastic exceed the margin. have mainly been produced. hinges are expected to form in the A similar type of collapse was In the Kobe earthquake, large columns rather than the beams. In the found in the Northridge earthquake in ground settlements and landslides ex­ worst case, this would result in story 1994.4 At least three precast concrete posed damage to foundations and piles collapse. Structural designers should parking structures partially collapsed (see Figs. 34 and 35). An effort to pay attention to these characteristics. due to a lack of ties connecting precast identify the damage to piles has also The columns should be provided with floor elements. Another defect was been made. Excavations for the in­ sufficient transverse reinforcement that the combination of large lateral spection of piles with 8 and I 0 MPa and careful design is needed. deformations and vertical load caused (1160 and 1450 psi) prestress levels Another two buildings sustained the crushing in poorly confined columns revealed shear failures at pile caps. same structural damage: hyperbolic that were not designed to be part of Nondestructive examinations by im­ precast, prestressed concrete shell roof the lateral load resisting system. pact wave propagation indicated that panels fell down on to the ground. the failure or severe cracks probably They were gymnasiums designed and occurred at the middle or tip portion of constructed in 1972 and 1974. PERFORMANCE OF the piles. It is anticipated that piles The newer one was a gymnasium of PILE FOUNDATIONS with 4 MPa (580 psi) average pre­ an elementary school (see Fig. 31). stress will reveal severe damage, al­ The building had 17 precast, pre­ Damage to foundations is, in gen­ though their inspection has not started stressed concrete panels as the roof. eral, invisible. However, in the as of this date. The roof panels were supported by the Miyagiken-oki earthquake of 1978, se­ Piles are currently designed to resist pillow beams at both ends through a vere damage to precast, prestressed elastically the seismic design load of rubber plate. Steel bolts were used to spun concrete piles was found under 20 percent of the total weight of the

July-August 1995 41 superstructure, while superstructures cause they are relatively regular ACKNOWLEDGMENT are required to be designed plastically and/or new structures with higher The authors wish to express their against the seismic design load corre­ strength and quality concrete than or­ appreciation to Dr. Y. Ohno of Osaka sponding to a base shear coefficient of dinary reinforced concrete structures. University and Dr. I. Takewaki of 0.3 if they are ductile frames. There­ Their resistance to earthquake loads Kyoto University for permission to fore, piles should be designed in the is considered to be higher than that of use their photographs in this article. same way. Presently, precast, pre­ the other structural types of concrete The authors also wish to thank H. stressed spun concrete piles are not buildings. Sato for preparing several of the provided with transverse reinforce­ 4. Design procedures for piles drawings. ment for resisting shear and for con­ should be revised in the same way as finement of compressed concrete. It is for superstructures. Use of ductile also recommended that ductile piles piles with transverse reinforcement with transverse reinforcement be used. is recommended. REFERENCES 5. One reinforced concrete building that had 37.2 m (122 ft) long cast-in­ 1. Preliminary Reconnaissance Report of CONCLUSIONS place prestressed concrete beams on the 1995 Hyogoken-Nanbu Earthquake The following conclusions can be the third floor suffered devastating (English Edition), Architectural Insti­ drawn on the basis of the field obser­ damage in ordinary reinforced con­ tute of Japan, April 1995. vations and investigations resulting crete columns on the second and first 2. Preliminary Reconnaissance Report on from the earthquake: floors. No damage was found in the Damages to Precast, Prestressed Con­ 1. Damage to buildings in the Kobe prestressed concrete beams. crete Structures of the Hyogoken­ earthquake was much more severe in 6. Precast, prestressed concrete shell Nanbu Earthquake (in Japanese), buildings built before the 1971 revision roof panels fell down in two gymna­ Japan Prestressed Concrete Engineer­ of the reinforcement requirement of the sium buildings. The cause of this fail­ ing Association, April 1995. Building Standards Law. Old buildings ure was attributed to support frames 3. The Standard for Structural Design should have been strengthened. that were too flexible and poor detail­ and Construction of Prestressed Con­ 2. The typical damage to reinforced ing to install the panels on to the sup­ crete Structures (in Japanese), Archi­ concrete buildings was collapse of the port girders. tectural Institute of Japan, 1987. soft first story, shear failure of Lastly, the findings reported 4. Preliminary Reconnaissance Report on columns and walls, collapse of a mid­ herein are a preliminary assessment Northridge Earthquake, January 17, height story, torsional failure due to of the causes of damage resulting 1994, Earthquake Engineering Re­ eccentricities of stiffness and mass, or from the Kobe earthquake. More de­ search Institute, Oakland, CA, 1994. failure of gas-pressure welded rein­ tailed reports will be forthcoming 5. Nishiyama, M., Seismic Response and forcement splices. after comprehensive investigations Seismic Design of Prestressed Con­ 3. The majority of the precast, pre­ have been carried out by researchers, crete Building Structures, Doctoral stressed concrete buildings performed engineers, the AIJ, the JPCEA and Thesis, Kyoto University, Kyoto, remarkably well in the earthquake be- other organizations. Japan, 1993.

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