196
DIAPHRAGMS FOR LATTICE STEEL SUPPORTS
Working Group 22.08
April 2002 196
DIAPHRAGMS FOR
LATTICE STEEL SUPPORTS
Working Group 22.08
MARCH 2002
Copyright © 2002 “Ownership of a CIGRE publication, whether in paper form or on electronic support only infers right of use for personal purposes. Are prohibited, except if explicitly agreed by CIGRE, total or partial reproduction of the publication for use other than personal and transfer to a third party; hence circulation on any intranet or other company network is forbidden”.
Disclaimer notice “CIGRE gives no warranty or assurance about the contents of this publication, nor does it accept any responsibility, as to the accuracy or exhaustiveness of the information. All implied warranties and conditions are excluded to the maximum extent permitted by law”. 196
DIAPHRAGMS FOR LATTICE STEEL SUPPORTS
SC22 WG08 Task Force Members: G. Gheorghita (Task Force Leader) (Romania), JBGF da Silva (Brazil), DA Hughes (UK)
During the preparation of this report, WG08 comprised the following: Members : JBGF da Silva (Convenor, Brazil), DA Hughes (Secretary, UK); S Kitipornchai (Australia), J Rogier (Belgium), RC de Menezes (Brazil), L Binette (Canada), K Nieminen (Finland), B Rassineux (France), R Paschen (Germany), E Thorsteins (Iceland), PM Ahluwalia (India), S Villa (Italy), G Nesgård (Norway), GA Copoiu (Romania), J Diez Serrano (South Africa), J Fernandez (Spain), R Jansson (Sweden), L Kempner (United States), JA Pardiñas (Venezuela).
Corresponding Members: H. Hawes (Australia), RP. Guimarães (Brazil), K Mito (Japan), TJ. Ploeg (Netherlands), G. Gheorghita (Romania), KE. Lindsey (United States), C Garcia (Venezuela).
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CONTENTS
Page
1 INTRODUCTION 1
2 THE QUESTIONNAIRE 3
2.1 SUMMARY OF RESPONSES TO THE QUESTIONNAIRE 4 2.1.1 Question 1 : “Who is responsible for the choice of the principles 4 employed for diaphragm design”? 2.1.2 Question 2 : “Are there any special loading cases considered for 6 diaphragm design not addressed by standards”? 2.1.3 Question 3 : “Is the Consultant / Designer’s experience 8 considered adequate to establish the levels / heights at which diaphragms should be employed”? 2.1.4 Question 4 : “What is the maximum distance allowed between two 9 horizontal diaphragms”? 2.1.5 Question 5 : “Are any special design rules employed to check the 10 permissible sag of the diaphragm members when erected”? 2.1.6 Question 6 : “During fabrication, is it necessary to adjust member 11 flange angles to ensure a flush connection between the flanges of the diaphragm and the horizontal support face members”? 2.1.7 Question 7 : “What forces are employed for the member design of 12 diaphragms”? 2.1.8 Question 8 : “Are different design philosophies employed for the 13 diaphragms in supports to be tested as opposed to those not to be tested”? 2.1.9 Question 9 : “Are there any differences in the design and or type 13 of diaphragm employed for heavy supports (angle, terminal) than for the relatively lighter suspension supports”? 2.1.10 Question 10 : “Are you using the geometry described in the 14 Figure”? 2.1.11 Question 11 : “Who is responsible for the choice of principles 15 employed for diaphragm design”? 2.1.12 Question 12 : “What is the maximum distance allowed between 15 two horizontal diaphragms”?
3 STANDARDS / GUIDES FOR DIAPHRAGMS 17 3.1 INTERNATIONAL AND NATIONAL STANDARDS 17 3.1.1 ASCE Manuals and Reports on Engineering Practice No. 52, 2nd 17 Edition “Guide for Design of Steel Transmission Supports” 3.1.2 ANSI/ASCE 10-97 “Design of latticed steel transmission 19 structures”.
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3.1.3 EUROPEAN STANDARD EN 50341-1 : 2001 Overhead electrical 19 lines exceeding AC 45kV, Part 1 General Requirements - Common Specifications 3.1.4 CENELEC prENV1993-3-1 : 1997 Overhead electrical lines 21 3.1.5 BS 8100 Lattice towers and masts; Part 3 : Code of practice for 22 strength assessment of members of lattice towers and masts : 1999 3.1.6 DIN VDE 0210 /12.85 : Planning and Design of Overhead Lines 26 with rated voltages above 1 kV 3.1.7 DIN 18 800 Part 2 : Steel Structures, stability, buckling of bars 27 and skeletal structures. 3.1.8 DIN 18 800 Part 1 : Steel Structures, stability, buckling of bars 28 and skeletal structures. 3.1.9 ASTM A6/A6M-00a : Standard Specification for General 29 Requirements for Rolled Structural Steel Bars, Plates, Shapes, and Sheet Piling 3.1.10 Australian Standard AS 3995 - 1994 : Design of steel lattice 30 towers and masts 3.1.11 Australia : HB C(b)1 1999. Guidelines for Design and 30 Maintenance of Overhead Distribution and Transmission Lines 3.1.12 Romania Departmental Rule PE 105 : 1990. Methodology for 30 OHTL Steel Tower Design 3.1.13 Slovenian Regulation On Technical Norms Covering Construction 33 Overhead Power Lines of 1kV to 400kV, U.l. SFRJ 65/88 3.1.14 Japanese Electrical Code JEC-127 : Overhead Transmission 33 Lines 3.1.15 ECCS Publication No 39. “Recommendations for angles in lattice 44 transmission towers”
3.2 INDUSTRY STANDARDS AND PRACTICES 45 3.2.1 Companhia Energética de São Paulo Technical (CESP) 45 Specification ET-EMTL-100/91 3.2.2 Centrais Elétricas do Norte do Brasil S.A. (ELETRONORTE) 45 Technical Specifications PEL-000-10001 3.2.3 Companhia Hidro Elétrica do São Francisco (CHESF) Technical 45 Specification ET/DET 132 -Rev 1 - 1995 3.2.4 Japan : Tohoku Electric Power Co. Inc. Standard For 46 Transmission Towers 3.3 OTHER REFERENCES 49 3.3.1 CIGRÉ paper 22-102 “Assessment and upgrading of 49 Transmission towers”. S Kitipornchai and F Alobermani.
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3.3.2 “Investigation into damage of tower horizontal plan bracing during 50 high intensity winds”. Introduction 50 Overview of Damage 51 Local Wind Effect Research 53 Design Specification 55 Member Failures 56 Material Testing 59 Conclusions 60 3.4 COUNTRY SPECIFIC COMMENTS 61
4 USE AND APPLICATION OF DIAPHRAGMS 63 4.1 STRUCTURAL CONSIDERATIONS FOR DIAPHRAGM 64 LOCATIONS 4.1.1 Torsion distribution 64 4.1.2 Support geometry 64 4.1.3 Stability 65 4.1.4 Maintenance Loading 66 4.2 ERECTION CONSIDERATIONS 66 4.3 OTHER BENEFITS OF DIAPHRAGMS 75 4.3.1 Support Upgrading 75 4.3.2 Support Repairs 75
5 CONCLUSIONS 79 5.1 REQUIREMENTS FOR DIAPHRAGMS 79 5.1.1 Reasons for Diaphragms 79 5.1.2 Necessary locations for diaphragms 80 5.1.3 Design conditions for diaphragms 80 5.1.4 Arrangement of Diaphragms 81 5.2 RECOMMENDATIONS 81
ACKNOWLEDGEMENTS 83
REFERENCES 85
ANNEX 1 : Questionnaire Issued ANNEX 2 : Detailed Responses to Questionnaire ANNEX 3 : Calculation example for plan bracing (J. Short) ANNEX 4 : Dimensioning of diaphragm members (Brazilian experience)
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LIST OF FIGURES AND TABLES Figure 3.1 : ACSE Figure 3.1 Model of Simplified Tower 18 Figure 3.2 : prEN 50341-1 (Figure J.9) - Typical plan bracing. 20 Figure 3.3 : prEN 50341-1 (Figure J.10) - Typical plan bracing 21 Figure 3.4 : prENV - Figure 5.3 - Typical plan bracing 22 Figure 3.5 : BS 8100 Figure 6 - Typical plan bracing 24 Figure 3.6 : BS 8100 Figure 7 - K bracing horizontals without plan bracing 24 Figure 3.7 : Table 1– Applied force as percentage of leg load, F 26 Figure 3.8 : Figure 1 DIN VDE 0210. 27 Horizontal loads acting on the tower body resulting from a torsional moment Figure 3.9 : DIN 18 800 Figure 2 – Initial bow imperfections of 27 member in the form of a quadratic parabola or sine half wave Figure 3.10 : PE 105 Figure. 8.2a 31 Figure 3.11 : PE 105 Figure. 8.2b 32 Figure 3.12 : PE 105 Figure. 8.2c 32 Figure 3.13 : JEC-127 Figure 34 34 Figure 3.14 : JEC-127 Figure 35 : Example of plan trusses 34 Figure 3.15 : JEC-127 Figure 37 35 Figure 3.16 : JEC-127 Figure 1 (Section 6) 37 Figure 3.17 : JEC-127 Figure 2 (Section 6) 37 Figure 3.18 : JEC-127 Fig. 3 (Section 6) Settlement force determined 39 from damaged tower members Figure 3.19 : Tokoku Diaphragms 47 Figure 3.20 : Tokoku A-Plans 48 Figure 3.21 : Tokoku B,C-Plans 48 Figure 3.22 : Figure 7 : Case Study 3 - Upgrading 400 kV suspension 50 Figure 3.23 : Report Figure 1 : Plan Configuration that failed 51 Figure 3.24 : Report Photo1 : View of damaged support 52 Figure 3.25 : Report Photo 2 : Underside view of damaged support 53 54 Figure 3.26 : Report Photo 3 : Scale topographical model for wind tunnel test Figure 3.27 : Report Table 1 : Comparison of loads arising from test 55 Figure 3.28 : Report Table 2 : Loads arising from test at 10-15 m 56 Figure 3.29 : Report Table 3 : Peak Gust speeds due to local effects 56 Figure 3.30 : Report Figure 2 : Modified diaphragm arrangement 57 Figure 3.31 : Report Figure 3 : Angle of member wind incidence 59 Figure 3.32 : Report Table 4 : Revised design wind pressures 60 Figure 4.01 : Typical locations for diaphragms (self supporting structure) 63
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Figure 4.02 : Typical torsion moment philosophy 64 Figure 4.03 : Diaphragms at change in leg slope 65 Figure 4.04 : Triangulation philosophy 65 Figure 4.05 : Collapse of inadequately supported leg extension 67 Figure 4.06 : Bending failure of main leg. 68 Figure 4.07 : Legs unsupported 69 Figure 4.08 : Temporary erection supports 69 Figure 4.09 : Temporary guys for initial stability 70 Figure 4.10 : Legs faces installed with temporary guys 70 Figure 4.11 : Location of diaphragms for erection purposes 71 Figure 4.12 : Main member joint located above diaphragm level 71 Figure 4.13 : Diaphragm providing stability to partially erected support 72 Figure 4.14 : Stability provided by two diaphragms 72 Figure 4.15 : Unsupported main members above bottom panel 73 Figure 4.16 : Two diaphragm levels 74 Figure 4.17 : Upgrading by adding a diaphragm 75 Figure 4.18 : 400 kV Suspension support with deformed structure 76 geometry Figure 4.19 : 400 kV Support with deformed geometry. Input data for 77 analysis
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1 INTRODUCTION
Following a review of available standards the Working Group determined there was little definitive documentation that addressed the application, design or requirement for horizontal diaphragms (plan bracing) within lattice steel supports. The Working Group assumed the task to establish the current state-of-the-art. The following objectives were set for the Task Force:
• To establish what is contained in current and past standards and or documentation; • To determine what philosophies are currently employed within the industry for the utilization, application and design of horizontal diaphragms; • To establish the reasons for the use of diaphragms, • To determine the possible different arrangements were available or possible and to provide guidelines for their application in new supports.
To establish what Industry practices and documentation was available (standards, guidelines, papers, etc.) the initial investigation assumed the form of a questionnaire issued to the Working Group members, representatives of CIGRÉ SC22 member countries and invited specialists worldwide. This questionnaire is given in Annex 1 and the responses, as received, are presented in Annex 2. The responses are collated, analysed and summarized in Chapter 2.
Chapter 3 presents the documentation considered relevant (or partly relevant) by the questionnaire respondents. Other references identified by the Working Group during the course of preparation of this brochure are also identified.
The usage and application of diaphragms are discussed in Chapter 4 and presents typical diaphragm arrangements, design examples and techniques. Annex 3 and 4 present the design methods adopted by some of the respondents.
Chapter 5 presents the conclusions of the report and concludes with a summary of the design conditions and application rules currently adopted by the industry.
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2 THE QUESTIONNAIRE
A questionnaire was circulated to CIGRÉ member countries and Working Group members who, in turn, requested the additional input of recognised overhead line support design specialists. The questionnaire issued requested responses to the following general questions. These questions were applicable to both self-supporting and guyed steel supports.
• Who is responsible for the choice of the principles employed for diaphragm design? • Are there any special loading cases considered for diaphragm design not addressed by standards? • Is the Consultant and or the designer’s experience considered adequate to establish the levels (heights) at which diaphragms should be employed? • What are the maximum distances allowed between two horizontal diaphragms? • Are there any special design rules employed? (i.e. to restrict the vertical sag of the diaphragm members when erected?) • During fabrication, is it necessary to adjust member flange angles to ensure a flush connection between the flanges of the diaphragm and the horizontal support face members? • What forces, if any, are employed for the member design of diaphragms? • Are different design philosophies employed for the diaphragms in supports to be subject to prototype test as opposed to those not to be tested? • Are there any differences in the design and or type of diaphragm employed for heavy supports (angle, terminal) than for the relatively lighter suspension supports? • Do you use any of the typical diaphragm arrangements shown in Figure 1 (see Annex 1)?
The complete questionnaire as issued is contained in Annex 1. A total of 21 responses from 13 countries were received as follows.
Country No of Country No of responses responses Australia 2 Italy 1 Belgium 1 Norway 1 Brazil 2 Romania 3 England 4 South Africa 1 Finland 1 Turkey 1 France 1 USA 1 Iceland 1 Venezuela 1
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2.1 SUMMARY OF RESPONSES TO THE QUESTIONNAIRE
The responses and conclusions made, as applicable, are summarized in the following Sub-Sections.
Questions 1 to 10 relate to self-supporting supports and 11 to 12 to guyed supports.
2.1.1 Question 1 : “Who is responsible for the choice of the principles employed for diaphragm design”?
Answers: The Consulting Engineer … Nil ( 0%) The Support Designer 13 (72%) Both 8 (38%)
From the additional comments provided it is noted three of those respondents who identified the support designer as the responsible party also stated the consulting engineer played a part in the overall design process.
When this aspect is incorporated the result becomes:
The Consulting Engineer … Nil ( 0%) The Support Designer 10 (48%) Both 11 (52%)
It is clear from the comments that the question itself was not well phrased.
For those respondents that answered “Both” the responsibilities were clarified as follows:
• The Designer selects and the Consultant accepts • The Consultant is responsible for special supports only (crossings, high antenna); the Support Designer is responsible for normal overhead line supports. • The Consultant’s specification normally indicates the conductor and earth-wire attachment heights and requirements for leg & body extensions. This leaves little option to the support designer to allocate diaphragm levels • The Consulting Engineer should specify minimum requirements for support stability and maximum distances between the diaphragm levels. The Designer should provide all necessary bracings to guarantee the support stability.
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Those respondents that replied “The Support Designer” commented as follows:
• The Consulting Engineer may require supplementary levels of horizontal bracings to provide transverse stability. • In many cases the levels where the horizontal bracings shall be installed are defined by specifications and/or Consultant Engineers. • Usually the Support Designer is the one who chooses the levels for the horizontal bracings based on the good practice and design standard. • The design needs to consider stability during support erection and under extreme wind loading. • The person who looks at the results. • Standards do not explain why horizontal bracings are needed therefore it is usually the support designer who is in charge. The Consulting Engineer should check the result. • There is no standard. • If a support is successfully tested, I don’t believe that the Engineer can reasonably insist that the Support Designer add additional bracings.
It seems the general opinion is that initial selection of the design principles is the responsibility of the designer and final responsibility and or acceptance is jointly shared between the designer and the consultant.
The respondents also share a general opinion that there is a lack of specific standards or guidelines available to be adopted by the consultant or purchaser and this, in turn, results in a situation where each support design has to be considered on an “as and when” basis with respect to diaphragms.
The main pertinent comments on the diaphragm location philosophy were:
• For the majority of overhead line supports, the location of diaphragms is in the main dictated by the requirement for and range of leg and body extensions required; • Dictated by transverse (and longitudinal?) stability requirements during high wind conditions; • Structure stability during erection.
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2.1.2 Question 2 : “Are there any special loading cases considered for diaphragm design not addressed by standards”?
Answers: Yes 10 48% No 11 52%
Respondents who answered “Yes” provided the following explanation of the additional rules:
• As per ECCS “Recommendations for angles in Lattice Transmission Towers” January 1985; European prEN 50341, ENV 1993-3-1 and draft for CLC/TC (SEC) 50 (CENELEC) the diaphragm should be calculated to resist: - a hypothetical local horizontal load of 1.5 kN multiplied by the distance (in m) between the main legs, applied to the centre of the outside angles (practical rule), in order to stabilize the diaphragm against overall buckling for wind forces applied on the diaphragm; - Deflection under this force shall not exceed L/1000; • Each angle of diaphragm is designed to resist a vertical load (weight of a man) applied in its centre. Values quoted were 1 kN and 1.4 kN. • As per ASCE Manual Pub. No. 52. • Diaphragm to resist the wind acting directly on horizontal members and attached redundant members. A gust factor appropriate to the largest dimension of the relevant area of the support face should be used. A case controlling the stiffness of the horizontal members may also be necessary. • Diaphragm to resist ice/snow and temperature loading. Ice and snow increase the bending and torsion moments on horizontal members (Japanese code). • The diaphragm needs to resist the wind loading on the frame it stabilizes. • The diaphragm must be designed to resist forces imposed by men whilst erecting or mounting it. • Diaphragm to resist torsion under broken wire conditions. • The diaphragm to be able to resist the force in the supported members.
Respondents who answered “No” provided the following comments:
• Extreme wind loading conditions; • There should be – and probably will be after this exercise.
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After further clarification with the respondents the loading following conditions are suggested to be worthy of additional consideration:
• Ice / snow loading which increase the bending and torsion moments on horizontal members. These respondents referred to Japanese Electrical Code JEC-127. • Wind loading; - Typical design software considers wind on a support panel to be resisted at the corners of the panel as discrete point loads. In fact the diaphragm may have to transfer the wind load imposed on the face panel bracing and redundant members both above and below the diaphragm to the corner leg reaction points. - Extreme wind loading plus appropriate gust should be considered on all adjacent panel members including redundant members and the diaphragm members themselves. - A case controlling the stiffness of the horizontal members may also be necessary. - Wind acting on snow / ice covered members. • stiffness of the diaphragm members; against vertical and horizontal deflection. • torsion loads arising from broken wire assumptions. • Safety / maintenance rule: the weight of a man plus equipment walking on the diaphragm members. Generally this is applied without reference to loads induced from other sources (i.e. wind) however this may not be considered safe • Maximum slenderness ratio limits. • Fabrication and detailing requirements.
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2.1.3 Question 3 : “Is the Consultant / Designer’s experience considered adequate to establish the levels / heights at which diaphragms should be employed”?
Answers: Yes: 17 (80%) No: 4 (20%)
In general, the respondents consider the designer’s experience is adequate for the determination of the required levels for the diaphragms and many quoted the same levels as already contained in the standards, i.e.
• at every bend-line; • at horizontals where the width is too great to allow a horizontal of reasonable size to span across the face.
Other comments received indicated the following aspects were necessary to be considered to determine levels for diaphragms:
• Temporary support stability during erection; • Erection method employed.
Other pertinent comments:
• Testing experience has indicated a slim support requires intermediate support every 15m. • If the support is designed employing non-linear analysis the designer’s experience is not adequate; • When looking at supports designed by different organizations to the same standard it is obvious different designers employ different criteria when determining the requirement for and location of diaphragms. Previous experience is only adequate providing a similar support has been previously tested.
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2.1.4 Question 4 : “What is the maximum distance allowed between two horizontal diaphragms”? Answers: According to your standard: 3 (14%) According to your experience: 15 (71%) Both 3 (14%)
Most respondents stated the diaphragm levels are established in accordance with the normal rules, i.e.
• at each cross-arm level to resist torsion induced by broken conductor conditions; • at each bend line of the main leg; • at the top of “K” bracings (where necessary for stabilization); • at the support waist; • at the top of the bottom panel (for hillside leg extensions).
The following general distances between two diaphragm levels were quoted:
Number of respondents Separation (m) From standard From Comments experience 8 – 12 1 average 10 8 – 12 (1) avg 10 (ice areas) 10 – 12 1 average 11 12 1 1 15 3 15 – 18 1 Average 16.5 18 1 15 –20 1 Average 17.5 20 2 22.5 1 1 23 2 Average 20 14.5 Ignoring Ice
One respondent stated consideration is be given to the requirement to maintain face alignment and general support stiffness;
One respondent considered a definite rule is inappropriate but gave no reasons.
It is clear that there is a very wide dispersion in the “experience” values for the maximum separation between diaphragms, ranging between 8 and 20 m. It is clear however that the separation quoted in the standards,
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22.5 m, is considered by nearly all respondents to be high. The average response is 14.5 m.
It is worthy of note that the separation between diaphragms distance normally is only applicable to very tall supports, typically above 60 m. For supports less than 60 m the general rules for placement of diaphragms normally provides separations less than 12 to 15 meters.
2.1.5 Question 5 : “Are any special design rules employed to check the permissible sag of the diaphragm members when erected”?
Answers: Yes: 9 (43%) No: 12 (57%)
The special design rules referred to by the “Yes” respondents are:
• The deflection of the edge member must be lower than 1/1000 of each length. • According to ASTM A6. • ASCE 10-90 Redundant members • Check calculations to be provided if dead load sag exceed “e” angle profile value. • Horizontal members should meet the “normal” stiffness checks for all other members. Limiting slenderness to 200 is usually sufficient. • Usually we use the limit on the slenderness (λ < 250) and a resistance check under a loading condition with a man walking on the horizontal member. • 1/500, DIN 18800. • The bracing should resist to a contract value depending on the internal force in the connect member, connection type, support use.
Some of the “No” respondents referred to the “non special” rules as follows:
• Check bending stresses due to weight (presumably self weight and or maintenance weight of man and equipment). • Vertical point load of 1.5 kN on all horizontals. This condition effectively prevents the selection of members that sag excessively. • A maximum slenderness ratio is imposed. Values quoted were 200 and 250 • Members to be able to withstand a vertical centre point load to represent the mass of a man and equipment.
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• The plan bracing must be “locked off” – i.e. not a mechanism. • Member sags checked during prototype assembly and testing. • Only the limits placed on the L/r values: 200 for compression members in a horizontal bracing system.
2.1.6 Question 6 : “During fabrication, is it necessary to adjust member flange angles to ensure a flush connection between the flanges of the diaphragm and the horizontal support face members”?
Answers: Yes: 16 (76%) No: 5 (24%)
It is interesting to note that none of the “No” respondents gave any reason why it was not necessary to ensure a ‘flush’ connection was detailed. Of those the replied “Yes” the following additional comments were provided:
• A “flush” connection is obtained by providing a bent gusset plate when the vertical angle exceeds 3º. • Necessary if the difference exceeds approximately 2º to vertical. • During detailing it is expected the member connections are flush. Normally a 2° “error” is allowed before bending is introduced. • For supports with double slope over 35-40% (correlated with the flange size of the main post angles). • In some cases it is necessary to bend the angle leg to connect the (horizontal) bracing. • It is easier to open the flange of the outside horizontal member than to close the flanges of all internal horizontal members. • Sloped washers are used to correct for the orientation of connected members.
One respondent commented that the diaphragm was required for erection/construction purposes to provide temporary stability during the erection as well as to provide a check for the setting of the foundation construction.
In general the respondents consider it is necessary for the detailing to maintain a flush connection between the diaphragm members and the horizontal face members of the support. The maximum tolerance is considered to be between 2 and 3º.
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2.1.7 Question 7 : “What forces are employed for the member design of diaphragms”?
Answers: Calculated: 14 (56%) Assumed: 8 (32%) Neglected: 3 (12%)
One respondent gave a positive answer to all three categories. Another stated the forces were both “Calculated” and “Neglected”; for this respondent it has been assumed that the forces may be neglected after confirming the results of a calculation and consequently this response has been considered as “Calculated”. Four respondents gave positive answers to “Calculated” and “Assumed”. When these are considered the responses become:
Calculated: 10 (50%) Assumed: 4 (20%) Calculated and assumed : 4 (20%) Neglected: 2 (10%)
In retrospect it is clear that the question was not clearly presented in the questionnaire. Many of the respondents were confused between “calculating” the forces from “assumed” design forces. This is clearly demonstrated by the breakdown of comments provided:
The “Calculated” forces quoted were: • Torsion forces arising from broken wire conditions. (at crossarm levels); • To resist extreme wind loading conditions; • The shear / torsion forces; • Horizontal components arising from changes in the main members slope; • Transverse brace force; • Diaphragm members are included in the analysis and as such are calculated; • To resist snow / ice forces calculated according to JEC-127.
The assumed forces were: • Hypothetical local horizontal wind load of 1.5 kN multiplied by the distance (in m) between the main legs, applied to the centre of the outside angles.
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• Assumed vertical load equivalent to the weight of a man and equipment applied in member centre. Values quoted were 1 kN, 1.5 kN • Assumed force in the horizontal face member equal to 2 to 2.5% of the force in the main member.
2.1.8 Question 8 : “Are different design philosophies employed for the diaphragms in supports to be tested as opposed to those not to be tested”?
Answers: Yes : Nil No : 21 (100%)
There was universal agreement on this point except one respondent qualified his “No” answer with the statement “Unless the support is a short one without body extension”.
2.1.9 Question 9 : “Are there any differences in the design and or type of diaphragm employed for heavy supports (angle, terminal) than for the relatively lighter suspension supports”?
Answers: Yes : 8 (36%) No : 14 (64%)
One respondent answered both “Yes” and “No”.
The differences between the diaphragms for suspension and angle / terminal towers were quoted to be as a result of :-
• In angle / terminal towers the forces in the edge horizontal members are larger than for those of suspension towers. Consequently the diaphragm members have to be designed to resist and provide support to those members with the proportional larger forces; • The member sizes are usually bigger when the dimensions of the support at the diaphragm level are larger; • Due to the larger size of the supports the same cannot be used; • Because internal forces in members are more important in tension supports than in suspension supports;
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Generally it appears the same design techniques are employed for both type of supports.
2.1.10 Question 10 : “Are you using the geometry described in the Figure”? For Figure see Annex 1 Page 4 Figure 1
Answers: Yes : 20 (77%) No : Nil ( 0%) Others : 6 (23%)
Five respondents gave positive responses to both “Yes” and “Others”.
The usage of the proposed diaphragm arrangements by the respondents is summarised as follows:
Diaphragm Number of Percentage type respondents who use 1 16 76% 2 16 76% 3 14 66% 4 4 19% 5 12 57% 6 13 62% 7 9 43% 8 8 38% 9 8 38% 10 2 9% 11 4 19% 12 7 33%
Pertinent comments relating to the diaphragm arrangements are as follows:
• When the width of the support becomes large it is sometimes more economic to introduce a crack or bend into the main diagonals. This has the effect of reducing the length and size of the redundant but produces high stresses in the members meeting at the bend and necessitates transverse support at the joint. • In all cases diaphragms must be rigid (not deformable). • The dimensions indicated for the proposed diaphragms are not treated as absolute maximum or minimum values. • The design condition for the weight of a man and his equipment will normally rule the member sizing (this assuming it is not taken in association with other loading conditions).
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2.1.11 Question 11 : “Who is responsible for the choice of principles employed for diaphragm design”?
Answers:
The Consulting Engineer: Nil ( 0%) The Support Designer: 12 (57%) Both: 4 (19%) Nobody 5 (24%)
Four respondents stated guyed supports are either not allowed or used in their countries.
Of those that have experience of guyed supports 75% stated the designer is responsible and 25% that the responsibility is shared between the Consultant and the designer.
Reasons given why the designer should be responsible were:
• due to construction (erection) reasons. • because the diaphragm position has an effect on the design.
2.1.12 Question 12 : “What is the maximum distance allowed between two horizontal diaphragms”?
Answers: According to your standard: Nil ( 0%) According to your experience: 16 (76%) No response 5 (24%)
It is clear from the above however that there is no standard or guidelines available for the maximum separation between diaphragms in the mast(s) of a guyed tower. All those responding stated the separation was based solely on experience.
The following separations were quoted :
• 3-6 meters • 5 meters • 6 to 8 meters • 6 to 7 meters • 9 meters
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• 10 to 12 meters • 33 meters
It is suggested the 33 meter value can be considered an ‘outlier’. Nevertheless there still remains significant dispersion between the minimum ‘experience’ separation of 3 meters and the maximum, 12 meters. The average response is approximately 7.5 meters.
The required location of the diaphragms within the mast were quoted to be at every trunk jointing / joint level (this presumably would be governed by the maximum member length available). Erection considerations were also an important aspect for the positioning of the diaphragm.
Reference was made to Russian industry practice however no references were provided. The limit in Russia is taken as 12 m where it is considered there is a reduction in mast capacity of 1 to 2% for each 1 meter of plan separation exceeding 10 m.
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3 STANDARDS / GUIDES FOR DIAPHRAGMS
The following Standards, Guides or Rules were identified by the Working Group and proposed for inclusion in this brochure as being relevant to the design, use or application of horizontal diaphragms in steel supports.
Those documents identified are given below and where possible are quoted verbatim. For compatibility the references to figures, etc., have been changed to suit the layout of this brochure. References to figures in this brochure are quoted in italics.
The documents are split into the following categories:
• International or national Standards, Codes and or Guides • Industry standards and practices • Actual experience with diaphragms, their use and or application for specific events.
3.1 INTERNATIONAL AND NATIONAL STANDARDS
3.1.1 ASCE Manuals and Reports on Engineering Practice No. 52, 2nd Edition “Guide for Design of Steel Transmission Supports” Chapter 2 : geometric Considerations Item 2.4 : Other Design Considerations Paragraph 2.4.1 Horizontal (Plan) Bracing
“In some structures horizontal bracing is required to distribute shear and torsional forces. Horizontal bracing is also used in square and rectangular towers and masts to support horizontal struts and to provide stiffer structures to assist in reducing distortion caused by oblique wind loads. Horizontal bracing is normally used at levels where there is a change in the slope of the tower leg to assist the bracing system in resolving the horizontal component.
In square and rectangular towers it is not unusual for the structure to extend 75 ft from the foundation to the first panel of horizontal bracing. The cross section of the tower, the stiffness of the lacing members, and the torsional load distribution normally determine how often horizontal cross bracing is required.
For structures with a square or rectangular configuration greater than 200 ft high, or heavy dead-end towers, it is suggested that horizontal bracing be installed at intervals not exceeding 75 ft. The spacing of horizontal bracing is dictated by general stiffness requirements to maintain support geometry and face alignment. Factors which affect this determination are type of
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bracing system, the face slope, the dead load sag of the face material, and erection considerations that affect splice locations and member lengths.
Commentary on Chapter 4 (Design of Members) 4C.4 Slenderness ratios
Damaging vibration of steel, members in latticed towers usually occurs at wind speed less than 20 mph since a nearly constant velocity is required to sustain damaging vibrations. Tests on a number of shapes with L/r values of 250 show that the possibility of damaging vibration is minimal (Carpena and Diana 1971; Cassarico et al. 1983). Tension-hanger members are prone to vibration, but L/r values as large as 375 have been used successfully.
In areas of steady winds over extended periods, such as mountain passes or flat plains, allowable L/r values may need to be reduced. Where severe vibration is a concern, careful attention must be given to framing details. The practice of blocking the outstanding leg of angles to facilitate the connection should be avoided.
Chapter 3 : Methods of Analysis Item 3.2 : Tower Model Paragraph 3.3.2 Practical Analysis Features “……. The diaphragm in section D-D (figure 3.1 – Figure 3.1) is a mechanism in the absence of member 8 (shown as a dotted line)”
Figure 3.1 : ACSE Figure 3.1 Model of Simplified Tower
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3.1.2 ANSI/ASCE 10-97 “Design of latticed steel transmission structures”.
Paragraph 4C.3 Geometric Configuration “In most structures, horizontal bracing is required to distribute shear and torsional forces. It is normally used at levels where there is a change in the slope of the structure leg. Horizontal bracing is also used in square and rectangular configuration structures to support horizontal struts and to provide a stiffer system to assist in reducing distortion caused by torsional and/or oblique wind loads.
For structures which are taller than 200 ft (61 m), or heavy dead end towers, it is suggested that horizontal bracing be installed at intervals not exceeding 75 ft (23 m). The spacing of horizontal bracing is dictated by general stiffness requirements to maintain tower geometry and face alignment. Factors which affect this determination are: type of bracing system, face slope, dead load sag of the face members, and erection considerations that affect splice locations and member lengths.
Paragraph 4C.4. Methods of Analysis
“……. The diaphragm in section D-D (figure 4C.1 - same as Figure 3.1 above) is a mechanism in the absence of member 8 (shown as a dotted line)”
3.1.3 EUROPEAN STANDARD EN 50341-1 : 2001 Overhead electrical lines exceeding AC 45kV, Part 1 General Requirements - Common Specifications provides the following recommendations on bracing patterns: Clause J.7 Additional recommendations on bracing patterns Clause J.7.1 Horizontal edge members with horizontal plan bracing (Figure J.9)
Due care should be taken in respect of the following: 1) Where the length of the horizontal edge members becomes large, for example when the slenderness ratio is greater than the one proposed in J.6.3.5.(3) or J.7.2.(5), or to secure the tower against partial instability, it is normal to provide a horizontal plan bracing. 2) The geometric length of the horizontal member for buckling is the distance between intersection points in the plan bracing for buckling transverse to the frame, and the distance between supports in the plane for buckling in the plane of the frame. 3) Care is needed in the choice of the vv or rectangular axes for single angle members, and the vv axis should be used unless suitable restraint by bracing is provided at or about the mid-point of the buckling length.
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4) The horizontal plan bracing needs to be stiff enough to prevent partial buckling. In case of doubt a good practice design rule is as follows: - the horizontal plan bracing, as indicated in Figure J.9 (Figure 3.2), has to resist a concentrated horizontal load F = 1.5 L, in kN, placed in the middle of the horizontal member, where: L = length of the horizontal edge member in m. - the deflection of the horizontal bracing under this load is limited to L/1,000.
Figure 3.2 : prEN 50341-1 (Figure J.9) - Typical plan bracing.
J.7.2 Horizontal edge members without horizontal plan bracing. (the following clause relates to the condition where a plan bracing is not provided. This is included as being relevant for the additional design checks necessary when diaphragms are omitted).
(1) For small widths of supports and for masts, plan bracing may sometimes be omitted. (2) As the horizontal members usually have compression in one half of their length and tension in the other, the effective length kL of the horizontal transverse to the frame shall be determined from Figure J.10 (Figure 3.3) depending on the ratio of the tension load, P2, to the compression load, P1 as given by the following formula
k = 0,085 x R2 - 0,316 R + 0,730
where: R = |P2/P1| and 0 R 1
(3) The radius of gyration about the yy axis (iyy) shall be used for buckling transverse to the frame except that for single angle members, either restraint by secondary bracing at intervals along the length shall be provided or the radius of gyration about the vv axis (ivv) shall be used.
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(4) For selection of buckling curve the member shall be considered as discontinuous at both ends. (5) The overall slenderness of the horizontal edge member on the transverse axis should be less than 250 (normative value).
Figure 3.3 : prEN 50341-1 (Figure J.10) - Typical plan bracing.
3.1.4 CENELEC prENV1993-3-1 : 1997 Overhead electrical lines (Although this previous draft version of EN 50341 has now been superseded it is included as containing other design requirements relevant to diaphragms)
Clause 5.6.3.8 Horizontal face members with horizontal plan bracing 1) Where the length of the horizontal face members becomes large, plan bracing may be introduced to provide transverse stability. 2) The system length of the horizontal member for buckling should be taken as the distance between intersection points in the plan bracing for buckling transverse to the frame, and the distance between supports in plan for buckling in the plane of the frame. 3) Care should be taken in the choice of vv or rectangular axes for single angle members, and the vv axis should be used unless suitable restraint by bracing is provided at or about the mid-point of the system length. 4) Selection of the effective slenderness factor k should be based on the end conditions of the member under consideration in accordance with 5.7.
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5) Where the plan bracing is not fully triangulated, additional allowance should be made for the bending stresses induced in the edge members by loads, such as wind transverse to the frame, (see Figure 5.3 - figure 3.4). 6) To avoid buckling, where the plan brace is not fully triangulated:- - the horizontal plan bracing should be designed to resist a concentrated horizontal force of 1.5L kN applied at the middle of the member where L is the length of the member in m; - the deflection (in m) of horizontal plan bracing under this force should not exceed L/1000.
Figure 3.4 : prENV - Figure 5.3 - Typical plan bracing
3.1.5 BS 8100 Lattice towers and masts; Part 3 : Code of practice for strength assessment of members of lattice towers and masts : 1999 Clause 5.3.3.3 Discontinuous cross bracing with continuous horizontal at centre intersection
“The diagonal members should be designed in accordance with 5.3.3.2. The horizontal member should be sufficiently stiff in the transverse direction to provide restraints for the load cases where the compression in one member exceeds the tension in the other or both members are in compression. This criterion will be satisfied by ensuring that the horizontal member without plan bracing withstands (as a strut over its full length Lh on
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the rectangular axis) the algebraic sum of the load in the two members of the cross brace resolved in the horizontal direction.
NOTE A maximum effective slenderness ratio for the horizontal of 180 is good practice (as a strut over its full length on the rectangular axis).”
Clause 5.3.4.2 Horizontal members with plan bracing “Where the length of the horizontal edge members becomes large, it is normal to subdivide them as part of the plan bracing which provides a convenient basis to do this. (See 5.3.5.).
The system length of the horizontal members is taken between intersection points in the plan bracing for buckling transverse to the face of the structure, and between supports in plane for buckling in the plane of the frame.
Care is needed in the choice of the vv or rectangular axes for single angle members. The vv axis should be used unless suitable restraint by bracing is provided in one plane at or about the mid-point of the system length. In this case buckling should be checked about the vv axis over the intermediate length and about the appropriate rectangular axis over the full length between restraints on that axis”.
Clause 5.3.4.3 Horizontal members without plan bracing “For small widths of towers and for masts, plan bracing may sometimes be omitted. [See also 5.5.1d) for reduction factor K1.] The rectangular radius of gyration should be used for buckling transverse to the frame over system length Lh (see Figure 7 – Figure 3.6). In addition for single angle members, the radius of gyration about the vv axis should be used over Lh2 [see Figure 7a)] unless restraint by secondary bracing at intervals along the length is provided in which case the system length is Lh1 [see Figure 7b)]. However, buckling about the rectangular axis will be critical except in the case of an unequal angle.
Additional allowance should be made for the bending stresses induced in the edge members by loads transverse to the frame, e.g. wind which can produce significant point loads”.
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Figure 3.5 : BS 8100 Figure 6 - Typical plan bracing
Figure 3.6 : BS 8100 Figure 7 - K bracing horizontals without plan bracing
Clause 5.3.5 Plan bracing “Plan bracings are required at bend lines of the leg, provide lateral restraint to long horizontal members and to distribute local loads from ancillaries, etc.
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Plan bracing is also required to maintain the shape of square towers, particularly where K bracing is used, and to distribute eccentric loads. The non-triangulated systems shown in Figure 6 (Figure 3.5) do not satisfy the first requirement.
Where the plan is not fully triangulated, additional allowance should be made for the bending stresses induced in the edge members by loads, e.g. wind, transverse to the frame whereby the main diagonal can impose a point load arising from the summation of the distributed loading on the various face members.
Attention should be given to vertical bending due to the self weight of plan bracing. Support from hip bracing can be used to reduce long spans. The design should be detailed to eliminate any face slope effect which promotes downward displacement of inner plan brace members”.
Clause 5.4 Secondary Members “In order to design secondary members it is necessary to apply a hypothetical force acting transverse to the leg member (or chord if not a leg) being stabilized at the node point of the attachment of the secondary member. This force varies with the slenderness of the leg member being stabilized and is expressed as a percentage of the leg load, F for which the following two checks should be carried out. a) Values of this percentage of the leg load for various values of the slenderness of the leg should be taken from Table 1 (Figure 3.7). The slenderness ratio to be used should be that of the restrained member, which will normally be L/rvv, where L is the length between nodes and rvv is the minimum radius of gyration.
The effect of applying this force in the plane of the bracing at each node in turn should be calculated. b) When there is more than one intermediate node in a panel then the secondary bracing systems should be checked separately for 2.5% of the leg load shared equally between all the intermediate node points. These loads should be assumed to act together and in the same direction, i.e. at right angles to the leg and in the plane of the bracing system.
In both cases a) and b) the distribution of forces within the triangulated secondary bracing panel should then be determined by stress diagrams or linear elastic analysis.
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Slenderness ratio, λ Applied force (percentage of leg load, F) % 0 to 40 1.02 45 1.15 50 1.28 55 1.42 60 1.52 65 1.60 70 1.65 75 1.70 80 1.75 85 1.80 90 1.85 95 1.92 100 2.00 105 2.08 110 2.16 115 2.23 120 2.30 125 2.38 130 2.46 >132 2.50 NOTE 1. The tabulated values assume minimum axis restraint of the leg by two such forces each in the plane of the bracing. Where the critical direction is in the plane of the bracing (e.g. round or cruciform sections) the values should be multiplied by a factor of √2. NOTE 2. These loads are not additive to the existing forces on the tower or mast. If the main member is eccentrically loaded or the angle between the main diagonal of a K brace and the leg is less than 258 then this figure may be unsafe and a more refined value should be obtained by taking into account the eccentricity moment and secondary stresses arising from leg deformation. NOTE 3. In general if the bracing strength requirements are met, the stiffness requirements should be sufficient.” Figure 3.7 : Table 1– Applied force as percentage of leg load, F
3.1.6 DIN VDE 0210 /12.85 : Planning and Design of Overhead Lines with rated voltages above 1 kV Clause 8.1.1.3 Erection and maintenance loads
“The erection and maintenance loads of 1.5 kN acting vertically in the centre of a member shall be assumed, however, without any other loads, for all members which can be climbed and are inclined with an angle less than 35º to the horizontal line.
In this case the permissible stresses for exceptional loading conditions apply”.
Clause 8.4.2 Calculation of permissible stresses Clause 8.4.2.1 Determination of member forces
“If a torsional moment Md related to the axis of the tower body results from a horizontal load Z, the horizontal forces may be determined according to
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Figure 1 (Figure 3.8). For the horizontal forces, each individual tower face may be treated as a plane framework.
When using this approach the ratio a/b shall not exceed 1.5. The shape of the tower must be prismatic or correspond to a truncated pyramid. At all crossarm levels and at changes of slope of leg members, horizontal bracings shall be provided and their adequacy shall be proven.
Figure 3.8 : Figure 1 DIN VDE 0210. Horizontal loads acting on the tower body resulting from a torsional moment
3.1.7 DIN 18 800 Part 2 : Steel Structures, stability, buckling of bars and skeletal structures. Chapter 2 Imperfections Clause 2.2 (204) Bow Imperfections
Individual members, members making up non-sway frames and members as specified in Item 207, shall generally be assumed to have the initial bow imperfections given in figure 2 (Figure 3.9) and table 3
Figure 3.9 : DIN 18 800 Figure 2 – Initial bow imperfections of member in the form of a quadratic parabola or sine half wave
Bow imperfections need not be assumed if members satisfy the criteria in item 739 of DIN 18 800 Part 1.
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Table 3. Bow imperfections Bow Type of member Imperfection wo, vo Solid member of cross section with following buckling curve 1 a l/300 2 b l/250 3 c l/200 4 d l/150 Built-up members, with 5 l/500 analysis as in sub-clause 4.3
Chapter 3. Solid Members Clause 3.1 General Element 302 Lateral buckling
Since the analysis of lateral buckling specified in sub-clauses 3.2 to 3.5 already includes both types of imperfection and second order effects, the initial forces and moments from first order theory shall be taken as a basis for calculations. Note 1 In the literature, the combination of equations (3), (24), (28) and (29) is referred to as first order elastic analysis with sway mode effective length (equivalent member method, for short). Note 2 Sub-clauses 3.4.2.2, 3.5.1 and 5.3.2.3 shall be taken into consideration when applying the equivalent member method to members notionally singled out of the frame.
3.1.8 DIN 18 800 Part 1 : Steel Structures, stability, buckling of bars and skeletal structures. Clause 7.5 Ultimate limit state analysis Clause 7.5.1 Criteria and detailing Element 739 Lateral buckling
Linear members and frames shall be verified for lateral buckling as specified in DIN 18 800 Part 2.
The effect on equilibrium of deformations occurring by second order theory may be disregarded if there is no more that a 10% increase in the significant bending moments as opposed to when deformations are determined by first order theory. This may be deemed to be the case if any of the following apply:
a) the axial forces, N, in the structure do not amount to more than 10% of the axial forces associated with the ideal buckling load, Nki,d, in the system (in plastic hinge theory, based on the statical system directly before formation of the final plastic hinge);
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_
b) the non-dimensional slenderness, λ K , is not more than 0.3 f y,d /σ N , _ where σN is equal to N/A, λ K is equal to λK/λa, λK is equal to sK/i,
and λa is equal to π E / f y,ki
c) the characteristics of all the linear members, ε = N /(E − I )d ,
multiplied by the coefficients of effective length, β = sK/l, are not greater than unity.
In cases where cross sections or axial forces vary, (E + I), NKi and sK shall be determined at the point in the member on which the ultimate limit state analysis is focused. In cases of doubt, more than one point shall be examined.
Note. In a), b) and c), N shall be assumed to be positive as specified in DIN 18 800 Part 2 (cf. item 314).
3.1.9 ASTM A6/A6M-00a : Standard Specification for General Requirements for Rolled Structural Steel Bars, Plates, Shapes, and Sheet Piling Clause 12.3.2 :
"The permitted variation in dimensions shall not exceed the applicable limits in Tables 16 to 25 [Annex A1, Tables A1.16 to A1.25], inclusive. Permitted variations for special shapes not listed in such tables shall be as agreed upon between the manufacturer and the purchaser.
Tables 21 and A1.25 refer to Permitted Variation in Straightness for S, M, C, MC, L, T, Z, and Bulb Angle Shapes, and recommend:
Nominal ASTM A6/A6M Variable Permitted variation Shape size ¼” in any 5’ or ¼ x (total Table 21 Camber < 3” length in feet /5)” Table 21 Camber >= 3” 1/8 x (length in feet/5)” Table A1.21 Camber < 75 mm 4 mm x total length in meters Table A1.21 Camber >= 75 mm 2 mm x total length in meters
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3.1.10 Australian Standard AS 3995 - 1994 : Design of steel lattice towers and masts Appendix G Guidance for structural analysis and design (informative). Clause G1 : Freestanding Lattice Tower.
For square or rectangular tower structures over 60 m high, it is recommended horizontal plan bracing be installed at intervals not exceeding approximately 20 m.
3.1.11 Australia : HB C(b)1 1999. Guidelines for Design and Maintenance of Overhead Distribution and Transmission Lines
The following clause is employed as the basis of diaphragm design.
Clause 3.8 Maintenance and construction loads
Maintenance Loads: 1.1Gs + 1.5Gc + 2.0Q + 1.5Ft
A factor of 1.1 has been applied to Gs (Gs = vertical dead loads resulting from non-conductor loads) which is accurately defined. A factor of 1.5 has been applied to loads which are static and reasonably well defined, e.g. intact conductor tensions (Ft) and conductor dead loads (Gc). A factor of 2 should be applied to Q (maintenance loads) which includes dynamic loads or loads that may be variable and not so well defined, e.g. weight of men and equipment or conductor tensions affected by the work being undertaken.
The conditions should be based on the worst weather conditions under which maintenance will be carried out. The limiting wind velocity for maintenance work is generally taken as 10 m/sec. This has minimal effect on an intact structure designed for the preceding loading conditions and may be neglected in this load case.
3.1.12 Romania Departmental Rule PE 105 : 1990. Methodology for OHTL Steel Tower Design Chapter 8.2 OHTL Lattice steel Towers:
Clause 8.2.3. Over the height of the tower horizontal reinforcements with function of diaphragm shall be provided. The distance between two levels of diaphragms (horizontal reinforcements) shall not be more than 8 to 12 m and one shall be provided as mandatory at the top of the bottom panel of the towers erected by bascule. [bascule is a rotation erection method - a common method employed in Romania for welded structures in the late 1970’s]
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Horizontal diaphragms are to be provided as mandatory at all changes of leg slope and at the level of all cross-arms. The diaphragm members (where no twist appear) shall be verified for a conventional compression load (fig. 8.2.a – Figure 3.10) equal to:
Q1 = 0.1 ∆Q + 0.00267 N
Where: ∆Q is the difference in bracing horizontal shear load on the portion between the two diaphragms acting in the diagonal; N is the axial load at the tower portion where the diaphragm is provided
For the condition where the changing in slope of the main members, the diaphragm shall be checked for the load:
Q2 = 0.1 ∆Q + 0.00267 N + 0.5 H
Where: H is the horizontal load arising from the change in leg inclination (fig. 8.2.b – Figure 3.11)
In the case of existing the twist moment M, the diaphragm shall be checked supplementary for a horizontal load Q3 for each support face (fig. 8.2.c – Figure 3.12) of:
Q3’ = M / 2a and Q3” = M / 2b Where: a and b are the tower widths at the level of diaphragm
Figure 3.10 : PE 105 Figure. 8.2a
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Figure 3.11 : PE 105 Figure. 8.2b
Figure 3.12 : PE 105 Figure. 8.2c
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3.1.13 Slovenian Regulation On Technical Norms Covering Construction Overhead Power Lines of 1kV to 400kV, U.l. SFRJ 65/88 (unofficial translation of the original version undertaken on behalf of the Working Group by dr. B Zadnik of IBE Consulting Engineers)
Article 246. “The torsion couple of binding forces on lattice towers of rectangular section can be replaced by two equal couples of forces acting in mutually parallel sides of the section in the plane of the torsion couple activity. Such mode of calculation is used in case the ratio of the section sides is 1.5 at the most, if the tower is of truncated pyramid form and if in the plane of the torsion couple activity, horizontal plan bracings are built in the tower structure.”
3.1.14 Japanese Electrical Code JEC-127 : Overhead Transmission Lines Selected extracts – unofficial translation of the original Japanese version undertaken on behalf of the Working Group by Mr. Norio Masoka and his colleagues of Tomoe Corporation).
Clause 5.1.2 Restrictions on the slenderness ratio of compression members The slenderness ratio (l/r) of compression members shall be subject to the following restrictions:
(a) 200 or under for main leg members (b) 220 or under for compression members other than main leg members (c) 250 or under for redundant members used to reinforce compression members
Where : l : is the length (cm) of the member between supporting points and r : is the radius of gyration (cm) of the cross-section around the buckling axis.
Note: In cross-sections A and B shown in Figure 34(c) and 34(d), (Figure 3.13) plan members, e.g. horizontal redundant members, shall be used to maintain the rigidity of the whole truss.
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Figure 3.13 : JEC-127 Figure 34
Figure 35 (Figure 3.14) shows examples of plan trusses.
Figure 3.14 : JEC-127 Figure 35 : Example of plan trusses
In general, compression members of the truss are reinforced three dimensionally by redundant members, as shown in Figures 34(c) and 34(d) (Figure 3.13), to reduce the buckling length. When the rigidity of the redundant members is small, the buckling curve of the compression members may follow Figure 37(b) rather than Figure 37(a) (Figure 3.15) because of the deformation of the redundant members and results in loss of their reinforcement role.
Thus, sufficient rigidity and strength are required for redundant members, especially for the three-dimensional supporting points of the main compression leg members.
However, rigidity and proof stress of redundant members change due to the influence of eccentricity, which is structurally unavoidable, and it is difficult to theoretically determine these values. In practice, as indicated in field tests
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on actual towers, they are calculated by assuming 1-2% of the main member stress is induced to the redundant members.
Figure 3.15 : JEC-127 Figure 37
Clause 6 : Snow loading on towers There are two classifications for snow loads acting on a structure:
1. Settlement force due to the sinking of accumulated layers of snow, 2. Creep and glide pressure due to the movement of the snow layer along a slope.
Many studies have been carried out on snow pressure and the following describes those which have been published or actually used in designs to date that required the design to consider the effect of snow load on tower structures.
1. Settlement force 1.1 Settlement force acting on an independent horizontal beam The settlement force depends on the shape of the beam receiving the pressure, size of the beam, slope of the beam from the horizontal axis, height of the beam in relation to the maximum depth of snow, weight and quality of the snow, etc.
The settlement force acting on the independent horizontal beam has, however, been studied considerably, and is expressed by the following experimental equation 1:
2 Fmax = 2.3 × Hw (1)
Where : Hw = Hs × ρ
Fmax: max settlement force (t/m)
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2 Hw: max snow weight (t/m )
Hs: max snow depth (m) ρ: average density of snow (t/m3)
This experiment was demonstrated in areas where the average density of all layers of accumulated snow was 0.26 t/m3 and the maximum depth of snow was 3.0 m, however it can be applied in areas where the maximum snow depth is more than 3 m (approx. 5- 6 m).
Nakamata suggested a practical equation based on the observed settlement force on a horizontal beam (Equation 2). The conventional beam is considered as infinite length, but the beam has an effective length in Equation 2, considering the edge effect and height of facilities:
πR 2 2R Fmax = + (b + l) + b H'γ ' (2) l l
Where : Fmax : settlement force (t/m) R : load area radius (m) (0.5-0.6 m) L : length of pressure receiving side (m) b : width of receiving side (m) H’ : snow depth at the top of facility γ’ : snow density (t/m3)
1.2 Design of tower leg against settlement force
For structures like tower legs, since they have a sophisticated 3-D truss structure, settlement force varies both due to the inclination of the members and the mutual interaction between the members and is complicated as compared to independent simple beams. There are few observations of them, but from past statistical analysis of damaged tower leg members we will derive the settlement force and apply it to design the required sectional area of the member.
(a) Snow resistance value method Legs of truss towers generally consist of triangles, and the settlement force which acts on each member triangle is proportional to the area. By engineering assumption, many members which actually suffered damage are statistically considered, and the limit is determined as the limit of the snow resistance value.
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For example, a and b members shown in Fig. 1 are defined as follows:
Z Σa = a (snow resistance value of a member, non-dimensional) (3) qa 2bsin β
Z Σb = b (snow resistance value of b member, non-dimensional) (4) qab2 sinα
Where: 3 Za, Zb : section modulus of members a and b (cm ); a, b : length of members a and b (cm); α, β : tilt angle of members a and b; q : area ratio; 1 for triangle, 2 for rhombus
Figure 3.16 : JEC-127 Figure 1 (Section 6)
Figure 3.17 : JEC-127 Figure 2 (Section 6)
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By obtaining the snow resistance value from this equation for many members buried in snow, it was obvious that members with values around 1-1.5 × 10-6 (value calculated for general structural steel SS41) are the limit to be damaged.
Therefore, when based on this method, the limit of the snow resistance are:
Σa or Σb ≥2 × 10-6
(b) Evenly distributed loading method
For calculating the sectional area of a leg member, the above- mentioned “snow resistant value” method can be applied to main trusses, but not to the parallel redundant members, etc. Generally, the following evenly distributed loading is used in many cases. In Figure 1 (Figure 3.16) when the settlement force acting perpendicularly unrelated to a member triangle is considered with the evenly distributed loading method, the bending moment, M, due to settlement force W is:
Wl 2 cos2 θ M = (5) K
where W: settlement force (kg/cm); l: member length (cm), θ: tilt angle of member; K: coefficient determined according to the support conditions of the member ends (K = 8 when both ends are pined, K = 12 for both ends fixed)
Bending stress σ (kg/cm2) occurring in a member with section modulus Z (cm3) from imposed moment M, is,
M σ = (6) Z
The equation which determines settlement force W can be derived from equations (5) and (6) as follows:
KσZ W = (7) l 2 cos2 θ
Example for equation 7: with the support conditions of the member ends taken as semi-fixed (K = 10) and a maximum bending stress σ of 4,100 kg/cm2 (SS41 steel) the ultimate settlement force W that the member can bear without bending and breaking, ▲* can be
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estimated by damage for the combined force arising from the settlement force and the creep and glide pressure
The results of investigations offered from Electric Power Companies for damage to members by the settlement force (when the max snow depth is 6.0 m or under) is shown in Figure 3 (Figure 3.18)
Figure 3.18 : JEC-127 Fig. 3 (Section 6) Settlement force determined from damaged tower members
The ultimate settlement force, W, which can be borne without bending and breaking is calculated and plotted for many members buried in snow. The solid area is for members which actually suffered damage. As can be seen from the damaged members in the figure, the settlement force for tower leg members is approximately 700 kg/m maximum.
Therefore, when calculating a safe sectional area of a member against settlement force the yield point stress should be used instead of steel bending strength in the following equation (No 8):
Wl 2 cos2 θ Z ≥ (8) Kσ Y
where : Z : required section modulus (cm3) W : settlement force, 7 kg/cm (= 700 kg/m)
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K : coefficient determined by support conditions at the member ends When each member is attached by one bolt, K should be 8 When each member is attached by more than one bolt, K should be 10 2 σy : yield point stress intensity of steel (kg/cm )
In addition, Equation 7 can be used to calculate the sectional area of the redundant member where axial force is not applied. However the following equation 9 should be used for members such as main leg members, bracing members and crossarm members which have axial forces.
σ ca N M • + ≤ σY (9) σ ka A Z
where: 2 σca : allowable compressive stress (kg/cm ) 2 σka : allowable buckling stress of the member (kg/cm ) N : axial stress of member (kg) A : sectional area of member (cm2) M : bending moment by settlement force (kg.cm)
2 2 M = Wl cos θ K Z : section modulus of member (cm3) 2 σy : yield point stress intensity (kg/cm )
2. Creep and glide pressure
Leg extensions of towers located on mountain slopes receive moving pressure due to snow creeping and gliding either between snow layers or between the snow and the ground. This pressure depends on the inclination of the slope, surface condition of the ground, snow quality, types of facilities, the layout, etc. When the slope is not steep and the ground surface has severe irregularities, the creep and glide pressure in most cases comes from snow creeping between layers and the pressure does not reach a high value, and ordinary strengthened anti-snow towers can be used.
Of course, when the slope is very steep and the ground surface is smooth, ground snow glide may occur and cause a far stronger creep and glide pressure compared to creep between layers, which may develop into an avalanche.
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Since it is impossible to design a tower for such a case, these areas should be avoided as much as possible. When a tower must be built for uncontrollable circumstances, using indirect protection equipment including avalanche prevention piles, stepped terraces or retaining walls may be a good idea. Although the problems associated with creep and glide pressure developing into an avalanche have been just started to be studied, experimental equations and examples of observations which are already published are shown below.
2.1 Experimental equation Furukawa determined an experimental equation from the observation value for an individual pole on bare ground with a slope of about 40°:
2 P = 4Hw (10) where P : creep and glide pressure of snow acting on a single pole (t) 2 HW : Snow density (t/m )
2.2 Example of observation For creep and glide pressure which acts directly upon leg members, Electric Power Development Co., Ltd., used a simple over-voltage meter installed on a part of the legs of the Tadami Trunk Line and the Miboro Trunk Line in the winter of 1962-63. A maximum 380 kg/m creep and glide pressure was obtained for a unit length of the member using a main leg member where the max snow depth is 2.7 m and slope angle is about 30°.
Tokyo Electric Power Company prepared an experimental slope with a 37° angle in Yuzawa-machi, Niigata Prefecture, where they observed the creep and glide pressure of snow applied to 4 piles built at 1 m intervals, in the first to fourth winters. In the first and the second winters, turf vegetation was on the slope. In order to promote glide at snow melting time, they watered with preinstalled pipe on the bottom of the snow coverage, but glide did not occur. Total creep and glide pressure of snow applied to 4 piles was 5.7 ton (max snow depth 3.3 m) in the first winter and 6.9 ton (max snow depth 3.8 m) in the second year. In the third winter, thatching over a 12 m wide slope and watering the base led to glide and a creep and glide pressure of 18.7 t (max snow depth 2.0 m) was measured. In the fourth winter, thatching over the width of 25 m resulted in glide developing into an avalanche by watering the base. Creep and glide pressure of snow was 26.3 t (max snow depth 2.6 m) immediately before an avalanche.
Thus, since the value of the creep and glide pressure changes remarkably with snow depth, gradient and vegetation, when designing, it is necessary to appropriately set the creep and glide pressure considering the geographical
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features of the construction site, vegetation, etc. by referencing these survey results.
2.3 Snow pressure equation in Switzerland As above, many studies have been performed on creep and glide pressure of snow in our country. In Switzerland, Haefely and others proposed a snow pressure equation and standardization was achieved as shown below:
(a) Snow pressure component parallel to a slope:
H 2 SN = γ KN(t / m) (11) 2 Where :
SN : snow pressure component parallel to a slope (t/m) γ : snow density ratio (t/m3) H : designed snow depth (m) K : creep coefficient (Ref. Table 6.1) N : glide coefficient (Ref. Table 6.2)
(b) Snow pressure component perpendicular to a slope:
α SQ = S (12) N tanφ N
Where :
SQ : snow pressure component perpendicular to a slope (t/m) α : ratio of snow quality 1 − 2ν α = c 2(1 −ν ) c
vc = 0.4 γ N : glide coefficient
SN : snow pressure component parallel to a slope (t/m) φ : slope angle
(c) Creep and glide pressure of snow
2 2 R = S N + SQ (13)
The above equation indicates snow pressure per unit width for long sideways facilities, such as avalanche prevention fences, and may change substantially depending on the influence from edge effects like tower legs.
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Snow density: 0.2 0.3 0.4 0.5 0.6 γ k/sin φ 0.70 0.76 0.83 0.92 1.05 Table 6.1 Creep coefficients
Glide coefficient (N)
Ground surface conditions
North-facing South-facing slope slope I Slope of cobble (>30 cm Ø), rugged slope of large rock 1.2 1.3 II Slope of gravel (<30 cm Ø), bushy place (1 m over height) , rugged slope 1.6 1.8 (>50 cm) III Small bushy place, rugged slope (>50 cm), grassy place 2.0 2.4 IV Smooth place, rock ground, long leaf grassy place, swamp 2.6 3.2 Table 6.2 Glide coefficients
Tokyo Electric Power Company obtained a value of about 1 m of influence which should be taken into consideration as an edge effect from the end pile to the outside of the experiment slope (Yuzawa- machi, Niigata) with an inclination of 37˚, using the observations in 1969-1974 and the snow pressure equation from Haefely. However, since this data is also only an example, it is necessary to evaluate appropriately by field observation, etc., when needed. In addition, when designing, since such settlement force and creep and glide pressure indicate max values when snow melts at the beginning of spring, snow depth, snow density, etc., at this time should be obtained by field measurement.
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3.1.15 ECCS Publication No 39. “Recommendations for angles in lattice transmission towers” Chapter 6 Redundants
“In order to reduce the effective length of the main legs and sometimes that of bracings it is frequently necessary to introduce stabilizing members which are normally unstressed, referred to as redundants. In order to produce satisfactory sizes of these members it is necessary to introduce a hypothetical force acting transverse to the member being stabilized and is expressed as a percentage of the leg load (or other member if not a leg). Values of this percentage are for various values of the slenderness ratio of the leg:
λ <40 45 50 55 60 65 70 75 80 85 90 95 100 % 1.02 1.15 1.28 1.42 1.52 1.60 1.65 1.70 1.75 1.80 1.85 1.92 2.00
This slenderness ratio is usually L/ivv where L is length between nodes and ivv = minimum radius of gyration. This load shall be applied at each node in turn and in the plane of the bracing. The bracing shall also be checked for 2.5% of the leg load equally shared between all the node points along the length of the leg in a panel, excluding the first and last, all these loads acting together and in the same direction, i.e. at right angles to the leg and in the plane of the bracing. It should be noted these loads are not additive to the existing loads on the tower.
If the main member is eccentrically loaded the stabilizing force quoted above may be unsafe, and a more refined is obtained taking into account the eccentricity moment.
Bracing aims at the maintenance of alignment of compression members in order to allow them to attain maximum strength. A real column is not straight and the fit of the brace is not exact : so the column deflects under load and the brace force depends on the stiffness of the brace and its design stiffness should be sufficient to ensure the that the design ultimate strength on the column can be reached. The brace should have sufficient strength to transmit the maximum brace force. Good initial alignment of main members due to good fabrication will reduce the loads in braces.
Common design practice confirmed by test experience is to design transverse braces for 1 to 2.5% of the maximum compression load of the column and to limit its slenderness at a maximum: in general if the brace strength requirements are met, stiffness will be sufficient. See also (10) and (11).
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3.2 INDUSTRY STANDARDS AND PRACTICES The following excerpts are taken from standards prepared by utilities and contractors.
3.2.1 Companhia Energética de São Paulo Technical (CESP) Specification ET-EMTL-100/91
Extract: The horizontal bracing must be placed on the upper side of the leg extensions and at bend lines of the main members of self-supporting supports. If it is necessary to guarantee stiffness, stability or for any another reason, the supplier must include additional horizontal bracing at the appropriate sections.
3.2.2 Centrais Elétricas do Norte do Brasil S.A. (ELETRONORTE) Technical Specifications PEL-000-10001
Extract: To guarantee stiffness and stability of the support, the supplier must consider the following: - Horizontal bracing must be provided at connections between the support body and the leg extensions or body extensions and on the lowest section of the support cross arms. - If it is necessary to guarantee tower stiffness, stability or any other reason, the supplier must include additional horizontal bracing at the appropriate sections. - The supplier must use bracing on the diagonal plan of the support leg extensions, when this procedure increases the stiffness and the stability of the support. - The purchaser may demand additional bracing be provided for the supports to ensure the desired stiffness.
3.2.3 Companhia Hidro Elétrica do São Francisco (CHESF) Technical Specification ET/DET 132 -Rev 1 - 1995
Extract: The horizontal bracing must be placed on the upper side of the leg extensions and at the waist of the supports. Horizontal bracing must also be placed at the appropriate sections of the guyed supports masts.
If it is necessary to guarantee stiffness, stability or any other reason, the supplier must indicate additional horizontal bracing at the appropriate sections.
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3.2.4 Japan : Tohoku Electric Power Co. Inc. Standard For Transmission Towers The following clause extracts have been provided
Minimum Size of Member are as follows:
Type of member Angle Tower Pipe Tower ≤ 77 kV L 60 x 5 ∅89.1 x 3.2 Main leg member 154 kV L 65 x 5 101.6 x 3.2 Crossarm member L 70 x 6 ≤ 77 kV L 45 x 4 Hanging Member 154 kV large tower L 50 x 4 Crossarm side ≤ 77 kV L 60 X 5 member 154 kV L 65 X 6 Horizontal and diagonal L 60 X 5 ∅89.1 x 3.2 member K lattice member 1st horizontal redundant L 50 X 4 ∅48.6 x 2.4 member 154 kV large tower L 50 X 4 Other member pipe member - ∅48.6 x 2.4 others L 45 X 4
Minimum size limit for horizontal members in tower body is determined by the bending inspection and slenderness ratio as follows:
4 ⋅σ ⋅ Z l = W where: l : Allowable Length (cm) W : Concentrated load (man weight) = 100 kgf σ : Allowable Bending Stress Z : Section Modulus
Lattice patterns Comparing the weight of panels, the patterns of diaphragms (here diaphragm refers to the support face bracing panel) and plans are to be determined as described in the following figures. Nevertheless, the pattern for any panel under the lowest crossarm must be the same as the upper one or one in the descending order.
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Figure 3.19 : Tokoku Diaphragms
Note : When the width of the tower at (point) 1 is more than 5 m the A-II pattern is adopted for the A-Plan (Figure 3.20)
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Figure 3.20 : Tokoku A-Plans
Figure 3.21 : Tokoku B,C-Plans
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3.3 OTHER REFERENCES
3.3.1 CIGRÉ paper 22-102 “Assessment and upgrading of Transmission towers”. S Kitipornchai and F Alobermani. Paper presented at CIGRE Paris session Sept ’00.
In their paper the authors describes a computer simulation technique for predicting the ultimate structural behaviour of self-supporting and guyed lattice transmission supports under static loading conditions. The simulation method employs non-linear analytical techniques and models the structures as an assembly of beam-column elements. The paper describes how this method was applied to three case studies to:
• Verify a new tower design. • Strengthen existing towers. • Upgrade existing towers.
The following is an extract from the paper relating to the upgrading which is relevant to this brochure:
4.3 Case Study 3 : Upgrading old towers “An existing 400 kV transmission line was designed and constructed almost 45 years ago. The line has performed its function and suffered no tower failure. The non-linear analysis was conducted to determine if the capacity of the towers in this line could be upgraded to carry larger and heavier conductor loads and to devise appropriate practical upgrading schemes for the towers.
One of the towers analysed is shown in figure 7(a) (Figure 3.22). It has a square base of about 9.5 m x 9.5 m and a height of about 49 m. The tower self-weight was calculated to be 127 kN. Seven loading conditions were specified based on the revised wind and incorporating larger conductor loads. The tower was modelled using 1245 elements, 660 nodal points and 3960 degrees-of-freedoms. In the non-linear analysis, the vertical loads were applied first up to 100% of their specified values, followed by the incremental application of the transverse and longitudinal loading.
Results from the non-linear analysis indicated that the tower did not reach the new ultimate design loads in four out of the seven loading conditions. Load factors at failure ranged from 0.87 to 1.32. Based on results from the non-linear analysis including the failure pattern, the tower was upgraded by adding a horizontal diaphragm as shown in Figure 7(b) (Figure 3.22). The upgraded tower was re-analysed using the same seven loading conditions. With this modification, the tower was able to carry the increased loads without any difficulty with the lowest load factor being 0.99.”
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Figure 3.22 : Figure 7 : Case Study 3 - Upgrading 400 kV suspension
3.3.2 “Investigation into damage of tower horizontal plan bracing during high intensity winds”. Alan King, Gordon Humphreys, Balfour Beatty Power Networks This report was prepared especially for the Working Group and as such is reproduced here in full.
Foreword This article summarizes several detailed reports prepared by W. R. Charman and D. C. J. Williams in 1983 after tower damage had been caused by high intensity cyclonic winds.
Introduction A tropical depression which occurred in 1983 intensified into a cyclone with maximum estimated surface sustained winds exceeding 33 meters per second.
The cyclone was ranked as one of the most intense ever. The local meteorological recording station recorded it’s highest ever wind speeds with a magnitude of 46.4 m/s mean hourly with a peak gust speed of 66.0 m/s. The previous highest speeds measured at this recording station were 36.6 m/s and 64.4 m/s respectively.
As a result of the cyclone major damage occurred to two towers and minor damage to several others in a 400 kV overhead transmission line. At the location, two lines, ‘A’ and ‘B’, run parallel to each other. At the time of the cyclone, line ‘A’ was in operation whilst line ‘B’ was not energized.
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This paper addresses the nature of the cyclone, the local effects of the wind loading, the consequential damage to the towers and the steps taken to strengthen the design.
Overview of Damage The two severely damaged towers, A2 and B1, were both suspension towers with 18m body extensions.
The main damage to the two towers appeared to be in the plan bracing of the 18m extension of the towers. The bracing at this level is in a configuration as shown in figure 1 (figure 3.23).
Figure 3.23 : Report Figure 1 : Plan Configuration that failed
Tower B1 suffered a partial collapse which was caused by the failure of the four horizontal members of the plan bracing (see photos 1 and 2- Figures 3.24 and 3.25).
Tower A2 had the same four horizontals in the 18 m extension fractured. A further five towers had minor damage to the horizontal members of plan bracings at various levels.
During the survey of the remaining towers in the area, it was noticed that several bolts were either loose or missing. All of these bolts were either tightened or replaced at the time and research was carried out by Professor A. G. Davenport at the Boundary Layer Wind Tunnel Laboratory of The University of Western Ontario, Canada, and also at Painter Brothers Ltd in Hereford to find out how this could have happened.
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Figure 3.24 : Report Photo1 : View of damaged support
All the tower bolts utilized palnuts. As both the contractor and the client supervised the erection of the towers thoroughly, it was felt that there was very little chance that the bolts were installed incorrectly. It therefore appeared that the bolts had worked loose whilst in service.
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Figure 3.25 : Report Photo 2 : Underside view of damaged support
Local Wind Effect Research Detailed meteorological research was also carried out by Professor Davenport.
A 1:2500 scale topographical model was constructed for use in a wind tunnel to investigate the local effects caused by the land surrounding towers A2, A3, B1 and B13 (see photo 3 – Figure 3.26).
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Figure 3.26 : Report Photo 3 : Scale topographical model for wind tunnel test
A previous topographical model built for determining the relationship between wind speeds at anemometer height to those at surface level (10 m), over open water and at gradient height on a nearby island was used to calculate the gradient winds experienced in the transmission line area. Since the two locations lay in almost exactly the same position with respect to the track of the storm, the gradient winds experienced at both locations would have been almost identical but with a time difference of around 3 hours.
Therefore, the new topographical model, used in conjunction with the predicted gradient wind speeds from the model testing enabled the tower loadings to be assessed through the course of the storm.
Both towers A2 and B1 are built on borrow areas with the site of B1 complicated by the presence of a rock pile.
The topographical model covered the whole area including the location of tower B13, where bolt loosening was found. The effects of the island upstream, the borrows, rock pile and local secondary ridges in the area were included in the model.
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Mean wind speed and turbulence intensities were measured at varying levels at several locations across the surface, namely the tower locations and conductor span midpoints adjacent to the towers of interest. The lowest measurement was at around 6ºm, the next at 15ºm which corresponds to the lowest main horizontal cross brace on towers A2 and B1. Several other heights were examined to a sufficient level where free stream conditions were approached. Mean speeds were then calculated as ratios of the gradient speeds, and root mean square values expressed as fractions of local wind speeds.
The strongest cyclone winds occurred at an angle of 130º. The transmission lines A and B were at 105º orientation. In addition to a wind angle of 130º, angles of 120º and 140º were investigated to check the possibility that the maximum wind speeds at the tower sites may have occurred at slightly off- peak cyclone winds. Measurements were also made with the wind in the direction of the line, irrespective of the model orientation, to find whether substantial differential loading could occur between the front and rear faces of the towers at 15 m level.
The maximum loads derived from the wind tunnel testing are shown in tables 1 and 2 (Figures 3.27 and 3.28).
Height Specification Cyclone (m) Load (kPa) Load (kPa) 10 5.6 4.4 25 6.0 4.9 50 7.05 5.6 72 7.65 6.15
Figure 3.27 : Report Table 1 : Comparison of loads arising from test
The difference in loading for the local effects versus the overall tower effects is due to the local topography. All loading was amplified by the presence upstream of another island.
Design Specification The specification asked for design wind pressures as detailed in table 1 (Figure 3.27), showing overall loads on the towers. The actual wind pressures recorded are also shown. It can be seen that the specification loads exceeded those experienced during the cyclone at all levels.
Table 2 (Figure 3.28) shows the specified local loads in the 10 to 15 m range in relation to the actual gust loads experienced at 15 m above ground level for each of the three towers, derived as described previously. It can be seen that for the two collapsed towers, the specification load exceeded that experienced during the cyclone whilst the actual loading on tower A3 was in excess of the specification.
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Tower Specification Cyclone Load (kPa) Load (kPa) B1 7.28 6.7 A2 7.28 7.25 A3 7.28 9.15
Figure 3.28 : Report Table 2 : Loads arising from test at 10-15 m
In light of the severity of the storm and the loadings predicted to have occurred, it appears that the approach taken during the design of the towers, whereby overall loading requirements are adopted except where hill crest loading analysis predicted larger loads, remains an appropriate approach.
The wind tunnel testing shows that the intense local loads between 10 m and 15 m are strongly fluctuating and can give rise to significant differential loading. This effect can be amplified by the resonance of the flexible plan bracing at this level. It appears that this loading effect exceeded the capacity of the bracing system on towers A2 and B1. For simply supported bracing members not subject to differential loading the factor of 1.3 for local effects given in the specification appeared adequate for towers in flat, open terrain.
The study suggested that, in general, local effects will increase the loading on the tower (see table 3 – Figure 3.29) although the specification loads remain conservative in most cases when compared to the cyclone loads. It is only in exceptional circumstances where higher local amplification factors combine with a particularly exposed site that higher loads are experienced.
Peak Gust Speeds m/s Level B1 A2 A3 10m 59.3 27.6 52.1 15m 73.2 76.1 85.5
Figure 3.29 : Report Table 3 : Peak Gust speeds due to local effects
Member Failures The investigation following the failures showed that tower B1 partially collapsed when a plan bracing member failed through bending at midpoint due to the loading from high speed buffeting. When this member failed, the adjacent side horizontal members buckled followed by the legs, which were no longer restrained. When the initial failed plan bracing member was inspected it showed signs of both high stress, low cycle and low stress, high cycle fatigue. The member would have failed by ductile rupture when the fatigue cracks reduced the cross sectional area, and therefore the bending resistance, to a point below the critical level.
Tower A2 also had all four horizontals in the 18 m extension fractured but this did not result in the failure of the whole tower.
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A third suspension tower, A3, experienced higher loads (see table 2 – Figure 3.28) but suffered no damage. Tower A3 did not have the 18 m body extension which only reinforces the conclusion that the weakness is in the plan bracing of the 18 m extension.
Elastic analyses were carried out to determine the maximum bending moments that could occur in the centre of the main horizontal member in the 18 m body plan bracing. The maximum moment this member could theoretically sustain at point of yield was calculated to be 4.9 kNm. Two loading cases were considered; front face fully loaded with back face unloaded and both faces equally loaded. These cases were considered to be boundary conditions, where the actual loading arrangement would be somewhere between the two.
The actual wind pressures required for member failure under the two boundary limits respectively were found to be 1.28 kN/m2 and 3.66 kN/m2.
Physical testing was carried out to confirm the capacity and likely failure mechanism of the plan bracing of the 18m body extension from the tower. A full scale plan brace was constructed with the main corner points on pin supports and the vertical bracing supports on roller bearings. In addition to testing the existing plan bracing system, a reinforced plan brace was also tested (see figure 2 – Figure 3.30).
Figure 3.30 : Report Figure 2 : Modified diaphragm arrangement
Six different tests were carried out: • Ultimate loads applied to reinforced plan system on both windward and leeward faces. • Ultimate loads applied to original plan system on both windward and leeward faces. • Ultimate loads applied to reinforced plan system on windward face only.
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• Incremental loads applied to windward face only to determine deflected shape and enable required cyclic loading magnitudes to be found. • Cyclic loads applied to windward face of original plan until fatigue failure occurred. • Cyclic loads applied to windward face of reinforced plan until fatigue failure occurred.
The results showed that the bracing in the original state could withstand the ultimate loading but suffered a permanent deflection vertically of all the main internal plan members. When the loading was reversed the deflected members returned to their original geometry.
In the reinforced state, all the members returned to their original geometry when unloaded.
Under cyclic loading, using a load derived from the previous test, a member on the windward face failed after 3869 cycles in the original frame. This member was then replaced with the proposed reinforced member and the frame re-tested. After 6000 cycles, no sign of fatigue or bolt loosening was visible and the test was stopped.
When only the windward face of the reinforced frame was loaded to ultimate conditions, the structure held 90% of the load but failed as it was increased to 100%.
Further wind tunnel research carried out by Professor Davenport focused on the local dynamic wind effects on long slender members giving rise to aerodynamic galloping. Large amplitude motion at the members natural frequency in a direction roughly perpendicular to the wind can occur when the wind approaches or exceeds a critical level. The mechanism is related to steep changes in the lateral lift coefficient coupled with an angle of attack that is such that motion perpendicular to the wind extracts energy from the wind through negative aerodynamic damping.
This phenomenon is well known for equal angles and, through wind tunnel testing on a sample member mounted on spring supports, it was shown to occur on the unequal angle sections used on the towers. The galloping effect can be expressed in terms of aerodynamic damping which, for high wind speeds, can be expressed as;
ρ V a ς a = C eqn. 1 ρ s fD
Where C is a constant relating to the geometry of the member, ( a/ s) is the ratio between the density of air and the structural material, V is the mean wind speed, f is the frequency of oscillation and D the characteristic dimension of the section.
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Instability arises if this aerodynamic damping, combined with the structural damping, reduces the total damping to near zero.
The parameters specific to individual structural members can be identified through testing which enables the development of a simple criteria for the identification of members susceptible to large amplitude oscillation.
If the parameter C is taken as a representative value of -0.5, as is the case for the tested member, then it can be shown that the most vulnerable members are those with a slenderness ratio greater than 246. If these members are inclined in such a way that the angle of attack, , of the wind could be near to the critical value of 15º to 40º (see figure 3 – Figure 3.31) to the horizontal then steps should be taken to decrease the slenderness.
Figure 3.31 : Report Figure 3 : Angle of member wind incidence
Material Testing The suitability of the materials used in the construction of the towers was determined from samples taken from the collapsed structures.
Three of the four fractured members from each of towers A2 and B1 were tested. Chemical analyses and tensile testing was carried out on the samples. The fractures were examined both macroscopically and microscopically and some calculations performed to relate the features in the fractured members to the service conditions that caused the failure.
Through the chemical testing, all materials were found to meet the requirements of grades 43A and 50B as appropriate.
The tensile testing showed that the yield strength of the fractured members was in excess of that required for Grade 50B designation.
Examination of the fracture areas showed that the fatigue cracking was of an order that could be shown to have occurred solely as a result of the cyclone. However, this did not preclude the possibility that the fatigue was initiated prior to the cyclone by lower velocity winds and it was only during the storm that they grew to critical size.
The examination was hampered by the presence of corrosion on the fractured face due to the time delay in the dismantling of the towers after the cyclone.
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Conclusions Following the various investigations into the tower collapses the following conclusions were made:
• The original design specification was adequate with regards to maximum gradient wind speeds but did not cover local effects on members adequately. • The original tower design was adequate given the design specification. • The plan bracing reinforcement measures implemented on the remaining and rebuilt towers was adequate for a repeat of the loads experienced during the cyclone. • Bolt loosening can occur in very specific circumstances when members with high slenderness ratios can be subject to high winds at an angle of attack between 15º and 40º. This effect can be eliminated by decreasing the slenderness below a critical level or the problem removed by adopting effective measures to prevent loosening.
The design specification was revised to take account of the findings. Localised loading wind pressures, in the absence of other loads, were increased to more adequately cover the situations identified, as shown in table 4 (Figure 3.32).
Panel Wind pressure on projected area Height of windward face members (m) (KN/m2) 9 12.78 15 10.07 21 9.03 27 8.15 40 8.32 50 8.90 60 9.95
Figure 3.32 : Report Table 4 : Revised design wind pressures
The other changes apply to differential loading between the front and rear faces, the resonant frequency of plan bracing systems and the localised loading on individual panels, including hip bracings.
Three differential load cases must now be included in the analysis to allow for the various distributions of loading between the front and rear faces. When considering differential loading, if the natural frequency of the plan bracing system is less than 6 Hz then the specification requires that a further load factor, up to 1.2 for 4 Hz, must be included. Plan bracing systems with natural frequencies less than 4 Hz are not permitted.
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Tower plan bracings modified to meet this revised specification have to date provided satisfactory service on these particular lines.
The Working Group extends their thanks to the authors and Balfour Beatty Power Networks for writing this paper especially for this brochure.
3.4 COUNTRY SPECIFIC COMMENTS
The following countries advised there were no specific country or utility standard that specifically addressed diaphragms and they generally followed one or more of the standards previously described in this section.
Spain; No design requirements for diaphragms.
Switzerland; The prevailing Ordinance Relating to Electrical Lines only addresses diaphragms in terms of their function to resist torsion. 4.4.3. and 4.4. of Annex 14, which both read as follows (translation): " The whole of the unbalanced horizontal load at 0°C, inclusive of the supplementary load of two conductors acting in the same direction, or of a multiple conductor, shall act on the support structure at those supporting points with the most unfavourable torsional loading, whereby no reduction factors shall be applied." Designs are performed according to the relevant Swiss or European standards for civil engineering.
Malaysia : Tenaga Nasional Berhad mainly uses the American ASCE Standard for the design of OHL Support. IEC standards are also employed when performing tower testing. There is no specific Malaysian standard covering diaphragms.
South Africa ; Eskom’s Code of Practices for construction refers to ASCE manual 52.
Italy ; Designs are performed according to the relevant European standards for civil engineering.
United Kingdom ; The National Grid Transmission Standards states 'The strength members of lattice towers shall be calculated using the BS Code of Practice DD133 with a partial safety factor on design strength to be taken as 1.15. There shall be compliance with BS8100 Clause 2.3.2 Class A towers'. DD133 has now been superseded by BS8100-3:1999: Lattice towers and masts - Part 3: Code of practice for strength assessment of members of lattice towers and masts.
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Australia ; Australia does not have a compulsory standard for the design of lattice structures for transmission lines. The closest to a standard is a guideline HB C(b)1 Guidelines for the Design and Maintenance of Overhead Distribution and Transmission Lines which refers to an Australian Standard AS 3995 Design of Steel Lattice Towers and Masts. This latter document uses the compression design curves from ASCE Manual 52 etc. The tensile capacity and load calculations are based on various Australian Standards and practices.
AS 3995, does not have any compulsory requirements for plan bracing but in an informative section recommends that towers over 60m tall have plan bracing at intervals not exceeding approximately 20 m. Because there has not been a compulsory Standard in use in Australia for the last 70 years, each asset owner has had their own standard on plan bracing or relied on the supplier to decide and the range of practice on plan bracing has been extreme. It is common to find that the only plan bracing is at the change in leg slope at the waist and at crossarm levels. There is a greater acceptance by designers that plan bracing is beneficial. Later towers often have plan bracing at the top of the legs as well (3-9 m above the ground). Practice over the last 15 years is to provide plan bracing at the top of the legs, every 12-15m, at the waist, at crossarm levels and any change in leg slope.
Because modern towers are often designed for higher voltages and are often wider than the early towers, it is appreciated that not only does plan bracing assist with load distribution and maintenance of shape of cross- sections, it also supports the centre of the tower face. This latter use is critical in large towers where the long members often do not have moment continuity across bolted connections.
Sweden ; There are no rules for horizontal plans/diaphragms in the current Swedish Standards however the Cenelec EN-standard will become valid in Sweden.
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4 USE AND APPLICATION OF DIAPHRAGMS
From the preceding Chapters it is clear diaphragms have traditionally been employed for the purposes of distributing torsion loads or as a “redundant panel” to reduce the slenderness ratios and therefore weight of main members.
Traditionally, for self-supporting supports, diaphragms are installed at the locations shown in Figure 4.01 and these are summarised as follows: • within the earth wire peak and the cross arm levels; • at any change in the slope of the support main leg members; • at the base of the support body extensions and or at the top of the leg extensions.
Figure 4.01 : Typical locations for diaphragms (self supporting structure)
Key to Figure 4.01: 1,2 Levels at which torsion is introduced to the support 3 Levels where there is a change in the slope angle of the main leg members
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4 Intermediate levels to provide rigidity 5 Interfaces between the basic support & body extensions 6 Interface between leg extensions and support body 7 Levels where there is a change of member geometry and additional out-of-plane support is required
Practical experience shows the role diaphragms play in a structure has been generally underestimated. This is evidenced by the number of collapses for which the identified cause can be directly attributable to the lack of a diaphragm or, if provided, one with an unsuitable arrangement or of inadequate strength. Diaphragms are now generally recognised to have a positive contribution to the overall structural integrity of a support against the imposed forces. Additionally they are necessary for both structural integrity and to provide necessary support during construction. These aspects are addressed below.
4.1 STRUCTURAL CONSIDERATIONS FOR DIAPHRAGM LOCATIONS
4.1.1 Torsion distribution
Diaphragms are required where torsion is introduced into the support. For diaphragms at levels 1, 2 and 3 of Figure 4.01 are required to be designed to provide resistance against torsion loads. These diaphragms transmit torsion moments typically from the conductor and earth wire cross arms into shear forces in the support faces (Figure 4.02).
Figure 4.02 : Typical torsion moment philosophy.
4.1.2 Support geometry
When there is a change in the main leg member slope angle diaphragms are typically provided (Figure 4.03).
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Figure 4.03 : Diaphragms at change in leg slope
Traditionally only the face members (horizontals) of the diaphragms have been designed as they are required to absorb the components of horizontal load arising from the change of slope of the main members (although if omitted from the analysis they theoretically can also be omitted). However when omitted double member compression effects on the bracing diagonals immediately below the leg slope change are encountered.
The internal diaphragm members have been considered as unloaded members and subject only to slenderness ratio restrictions and, when required, designed to resist a centre point load representing the mass of erection or maintenance personnel and their tools (see item 4.1.4).
4.1.3 Stability
Diaphragms provide rigidity to the structure. The report from Balfour Beatty Power Networks clearly demonstrates proper attention needs to taken of the forces induced by the wind on the support. Supports are susceptible to deformation from wind loads on the support face bracing members. Appropriate application of diaphragms is required to ensure:
• the support geometry remains as assumed in the design and, • the face members act as a spatial (3D) truss. To ensure this diaphragms are required to be detailed as rigid frames, i.e. be triangulated (Figure 4.04).
Figure 4.04 : Triangulation philosophy
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Experience has also shown that diaphragms play an important role to provide structural rigidity to the support against wind and wind induced vibration. The recent issue of Standard EN 50341-1 has introduced a hypothetical load case representing wind forces on the structure to design the diaphragm. Romania Departmental Rule PE 105 has included load cases representing these wind forces.
Diaphragms are also, in some countries, subject to other load forces such as snow and ice loads. These may be significant. JEC-127 identifies loading arising from the flow of snow which, if deep enough, will affect the design of a diaphragm.
4.1.4 Maintenance Loading
Loading due to the presence of maintenance or erection personnel is recognised by most standards, guidelines and Codes of Practice as a necessity. These forces are typically represented by vertical forces applied at the centre of structure members defined as “upon which a man can stand”. The Values of this mass typically range between 1 kN and 1.5 kN. This mass is normally considered to be applied without other stresses and is typically un-factored.
The Working Group suggests this requirement needs to be reinforced to properly take personnel safety into account and recommends a maintenance load condition be specified taking the following into consideration:
• a maximum wind velocity under which erection or maintenance can or will be performed • the selected mass of the vertical load applied is truly representative of the number of maintenance or erection personnel present. Evidence shows there is normally more than one individual required to perform most aerial tasks. • most guidelines addressing personnel safety requires a factor of 2 or 2.5 to be provided. This should be applied to the actual member loads while personnel are present on the structure.
4.2 ERECTION CONSIDERATIONS
Most members of the Working Group advised of previous experience with erection problems which could have been avoided if attention had been given to the provision of, or proper placement of, diaphragms.
During erection, unless the members are properly supported they will be affected by gravity, wind and erection (personnel) loads. The provision of temporary or permanent diaphragms is necessary to provide support stability at this stage. Numerous examples of support deformation, or even
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collapse, have been reported that can be directly attributable to the lack of proper support being provided during the erection process. Typically failure occurs at the tower connection with the foundation (see Figures 4.05, 4.06 and 4.07).
Figure 4.05 : Collapse of inadequately supported leg extension
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Figure 4.06 : Bending failure of main leg.
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Figure 4.07 : Legs unsupported.
Most designers have recognised the need for supports, whether temporary (guys) or permanent (diaphragms) during the erection process to avoid support instability (Figure 4.08).
Figure 4.08 : Temporary erection supports.
Actual applications are demonstrated in the following Figures.
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Figure 4.09 : Temporary guys for initial stability.
Figure 4.10 : Legs faces installed with temporary guys.
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The location of the diaphragm relative to the main leg joints has a significant influence on the ease of support erection (Figure 4.11). If the leg joints are detailed below the level of the diaphragm, the diaphragm itself cannot provide adequate support and temporary guys will be required to support the legs and bracings.
Figure 4.11 : Location of diaphragms for erection purposes.
The Working Group recommends leg joints always be provided above the level of the diaphragm as shown in Figure 4.12
Figure 4.12 : Main member joint located above diaphragm level.
Diaphragms also play a vital role in verifying the constructional quality control. If, after assembly, a diaphragm can be easily installed it usually infers all preceding erection and foundation setting is adequate. In such a
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way diaphragms emulate the role of construction templates as demonstrated in the following Figures.
Figure 4.13 : Diaphragm providing stability to partially erected support .
Figure 4.14 : Stability provided by two diaphragms.
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The provision of temporary supports or diaphragms is not restricted to the lower panels of the support. The designer should consider what erection methods are to be employed and take suitable precautions during the design and detailing stages.
Figure 4.15 : Unsupported main members above bottom panel.
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Figure 4.16 : Two diaphragm levels.
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4.3 OTHER BENEFITS OF DIAPHRAGMS
4.3.1 Support Upgrading
Diaphragms have proved a valuable tool for the purpose of upgrading or repair of existing transmission line supports. Diaphragms may be used to increase support resistance by providing additional rigidity and overall stability to the support.
In CIGRÉ paper 22-102, “Assessment and Upgrading of Transmission Supports”, the author conducted a non linear analysis to determine whether the structural capacity of a existing support could be upgraded to resist the loads arising from the application of a larger and heavier conductor. The results of the analysis indicated the existing support could not support the new upgraded ultimate loads in four of the seven considered loading conditions. However, based on the results obtained from the analysis and after introducing a horizontal diaphragm (Figure 4.17) the increase in structural capacity was adequate to permit the upgrading to continue.
Figure 4.17 : Upgrading by adding a diaphragm.
4.3.2 Support Repairs
It is recognized the internal member geometry (hyper-static geometry) of a support can change over time. These changes can arise from excess initial erection tolerances, overloading, theft of steel members, foundation movement, etc. These geometry changes do not always result in the collapse of the support as, typically, there remains adequate resistance to the normal “every-day” loading regimes. However, these geometry changes are, in the majority of cases, unseen and remain undetected. In cases where these changes are identified, normally from observation of deformation of members, and remedial works are performed or the loading
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causing the deformation is removed, the supports do not return to their original design geometry.
To quantify the residual strength of an existing support requires a very large number of dimensional measurements to be taken to determine its in-situ geometry and, in most cases, the analysis will indicate modifications to the support are necessary to achieve the required resistance.
Figure 4.18 shows a hyper-static structure where the support “waist” deformed due to the theft of some of the elements of the horizontal diaphragm. Further deformation of the connecting gusset plates of the diagonals resulted. Removing and replacing the existing diagonals to obtain the original design geometry was considered too dangerous for personnel and the overall security of the system.
Figure 4.18 : 400 kV Suspension support with deformed structure geometry.
The solution adopted was to include additional angle members to the horizontal diaphragm thereby making it fully triangulated and capable of carrying the loads imposed by the deformations and, in turn, stabilize the structure in its new revised geometry (Figure 4.19).
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Figure 4.19 : 400 kV Support with deformed geometry. Input data for analysis.
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5 CONCLUSIONS
Traditionally, the inclusion of diaphragms in supports has been for the following reasons:
• to enable torsion forces arising from broken conductors or earth-wires to be distributed around the faces of the support; commonly known as torsion plans; • to avoid the “double compression” effect in face bracings arising at changes of support leg slope (normally only affects the design of the face members of diaphragms); • to reduce the overall mass of a support by providing additional out-of- plane reinforcement to main members; • for erection purposes.
Recently it has been recognised that diaphragms play a significantly more important role to the overall structural integrity of the support.
Supports are typically designed to resist wind and other forces represented as point loads applied to the nodes of the main members. This practice ignores the bending effect of the wind on the bracing members. Recent standards have partially addressed this matter in that diaphragms are required to be designed to resist these loads, but only in terms of providing a design condition with a “typical” load. The working Group suggests support designs should consider the actual loads imposed on diaphragms arising for the design wind forces (when greater than the “typical” load contained in the standards.
Also support diaphragms are an important “tool” for the upgrading and or repair of existing supports providing a relatively simply means whereby the support resistance may be restored or even increased.
5.1 REQUIREMENTS FOR DIAPHRAGMS
The International, national and industry standards referenced in this brochure collectively identify the following requirements for diaphragms:
5.1.1 REASONS FOR DIAPHRAGMS
a) to distribute torsion and shear forces; b) to maintain the shape of supports; c) provide structure stiffness and reduce partial instability to reduce distortion caused by torsion and or oblique wind loads; additional loading effects arise from ice and snow loads c) to reduce the slenderness ratio of long horizontal members;
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d) for erection considerations that affect splice locations and member lengths; e) to provide lateral restraint to long horizontal members and to distribute local loads from ancillaries;
Diaphragms have been shown to provide a useful tool for the upgrading and repair of damaged supports.
5.1.2 Necessary locations for diaphragms
a) where the leg slope changes; b) at intervals not exceeding 23 m (on towers with heights > 61 m) ASCE. The Working Group would however recommend a maximum spacing of 10 m be adopted based on the practical experiences of the respondents to the questionnaire. c) at locations where torsion is induced into the support or required to be distributed between the support faces d) at positions to enable the adopted erection methods can be performed without inducing undue stresses on the support members
5.1.3 Design conditions for diaphragms
a) Resistance against wind load. EN 50341. A force (in kN) of 1.5 L applied horizontally at centre of tower width L (meters) with limiting condition that horizontal deflection should be less than L/1000. (unless exceeded by the actual force imposed by the design wind pressure). b) Maximum Slenderness ratio The overall slenderness of the horizontal edge member on the transverse axis should be less than 250 (normative value) - the effective length of the member is taken between intersections; c) vertical bending due to self weight;
d) detail member connections to eliminate any face slope effect which promotes downward displacement of inner plan brace members; e) maintenance loads including any dynamic effects due to the maintenance activities of men and equipment. The resulting loads should be considered concurrent with a limiting “maintenance” wind velocity (10 m/s). A minimum factor of safety of 2 is recommended f) resistance against the self weight of ice formation on members g) resistance against forces imposed by deep snow movements (also affecting face members).
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5.1.4 Arrangement of Diaphragms.
The usual diaphragm arrangements have been indicated in Figure 1 of Annex 1. Other diaphragms arrangement were suggested by the participants (see Annex 2). Examples of calculation methods applicable to the economic selection of diaphragms are provided in Annex 3 (UK) and Annex 4 (Brazil).
5.2 RECOMMENDATIONS
The Working Group recommends designers consider the following matters when designing or checking support resistances in relation to the positioning and or application of diaphragms.
• In all cases the arrangement of diaphragms should be a fully triangulated arrangement capable of independent analysis. • dimensioning of diaphragm members should ensure the wind forces imposed on the face members of a support can be safely transferred to the load points assumed in the design. • the positioning of diaphragms should consider the effects of face member oscillation, typically from low velocity winds, to avoid fatigue and other effects. • The support erection method to be employed and the forces likely to arise. • The forces imposed on the members by the erection personnel. Special attention should be made to ensure the safety of erection personnel. This should mean considering an increase in the number of personnel normally assumed to be on the tower during erection (typically one with a mass of 100 kg) and the resulting maximum mass to be applied to the centre of a member. It is recommended this mass should be multiplied by a suitable factor of safety and include the possibility of simultaneous external (wind) forces during the erection stage. • A maximum spacing between diaphragms of 8 to 12 m (10 m being recommended by the Working Group). This separation may be modified providing an analysis of the structure stability is performed in which the wind loads are imposed as uniformly distributed loads on the members themselves (and not averaged between nodes). Care should be taken to consider the method of erection employed.
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ACKNOWLEDGEMENTS
The working group acknowledges the support received from Mr. Henning Obro (Denmark) and Mr. Luis Carlos Bertola (Argentina) for their review of this document. Thanks are also given to the following for their assistance in compiling references and country practices:
Mr. G.B. Hean of Asea Brown Boveri, Malaysia Mr. S. Pankhurst and Mr. A. King of Balfour Beatty Power Networks, UK Mr. B. Zadnik of IBE Consulting Engineers, Mr. P.J. Riisiö of Finnmast Oy, Finland Mr. N. Masoka of Tomoe Corporation, Japan Mr. D. Muresan of Electromontaj, Romania Mr. P. Dallèves of Energie Ouest Suisse EOS, Switzerland Mr. A. Gallego of RED Electrica de Espana SA, Spain Mr. G. Brown of O'Donnell Griffin, Australia Mr. C. O'Luain of ESBI Engineering, Ireland. Mr. B. Rabjohns and Mr. M. Tunstall of National Grid, UK.
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REFERENCES:
(1) CIGRÉ paper 22-102, Assessment and upgrading of transmission supports, S. Kitipornchai & F. Albermani, for 38th CIGRÉ Session in Paris 2000. (2) ASCE Manuals and reports on engineering practice n° 52. Guide for design of steel transmission towers. Second edition – 1988. (3) CENELC European committee for electrotechnical standardization. Overhead Lines exceeding AC 45 kV. (4) ECCS Publication No 39. “Recommendations for angles in lattice transmission towers”, January 1985. (5) EUROPEAN STANDARD EN 50341-1 : 2001 Overhead electrical lines exceeding AC 45kV, Part 1 General Requirements - Common Specifications. (6) CENELEC prENV1993-3-1 : 1997 Overhead electrical lines (7) CLC/TC (SEC) 50 (CENELEC) (8) Japanese Electrical Code JEC-127 : Overhead Transmission Lines (1979 and 1965) (9) ASTM A6/A6M-00a : Standard Specification for General Requirements for Rolled Structural Steel Bars, Plates, Shapes, and Sheet Piling (10) ANSI/ASCE 10-97 “Design of latticed steel transmission structures” (11) DIN 18 800 Part 2 : Steel Structures, stability, buckling of bars and skeletal structures (12) Romania Departmental Rule PE 105 : 1990. Methodology for OHTL Steel Tower Design (13) Finnish code of practise A4-93 (superseded) (14) DIN VDE 0210 /12.85 : Planning and Design of Overhead Lines with rated voltages above 1 kV (superseded by EN 50341 part 1 and 3) (15) Eurocode 3: Design of steel structures (16) European Standard: EN 50341-1:2001 – Overhead electrical lines exceeding AC 45kV (17) DIN 4114 (superseded by DIN 18800 in 1990) (18) CESP - Companhia Energética de São Paulo (19) ELETRONORTE - Centrais Elétricas do Norte do Brasil S.A. (20) CHESF - Companhia Hidro Elétrica do São Francisco (21) Ordonnance sur les lignes electriques (OLEI) du 30 Mars 1994, Switzerland (22) Eskom (South Africa) Code of Practices for construction (23) Australian Standard AS 3995 - 1994 : Design of steel lattice towers and masts (24) Italian Standard - CEI 11-4 V1/V2/V3 (Revision)
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(25) National Grid Transmission Standard NGTS 2.4 - Issue 2, September, 1999 (26) BS 8100 Lattice towers and masts; Part 3 : Code of practice for strength assessment of members of lattice towers and masts : 1999 (27) Australia : HB C(b)1 1999. Guidelines for Design and Maintenance of Overhead Distribution and Transmission Lines (28) Slovenian Regulation On Technical Norms Covering Construction Overhead Power Lines of 1kV to 400kV, U.l. SFRJ 65/88 (29) Companhia Energética de São Paulo Technical (CESP) Specification ET- EMTL-100/91 (30) Centrais Elétricas do Norte do Brasil S.A. (ELETRONORTE) Technical Specifications PEL-000-10001 (31) Companhia Hidro Elétrica do São Francisco (CHESF) Technical Specification ET/DET 132 -Rev 1 – 1995. (32) Tohoku Electric Power Co. Inc. Standard For Transmission Towers (33) “Investigation into damage of tower horizontal plan bracing during high intensity winds”. Alan King, Gordon Humphreys, Balfour Beatty Power Networks.
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ANNEX 1
“The Questionnaire”
CIGRE / SC 22 / WG08-Task Force 3-3.3 Page.1 HORIZONTAL BRACINGS OF THE TOWERS No.
TABLE 1. QUESTIONNAIRE FOR THE HORIZONTAL BRACINGS PHILOSOFY Comments to the Comments No Questions Questions Answers to the Answers 1 Who is in charge to establish Please comment the reasons. the Consulting the levels were is necessary Engineer to provide the horizontal the Tower bracings for the towers? designer both 2 There are special loading If the answer is YES, please cases, ignored by the actual describe the loading cases, YES standards, but necessary to or indicate the references. be considered in the horizontal bracings NO philosophy? 3 In the establishing the levels, If the answer is no, please YES is enough the describe the calculus to be Consultant/Designer provided, or indicate the NO experience? references. 4 What is the maximum If the standard exist, please According to your distance (in meters) between attach a copy of the relevant standard two horizontal bracings to be paragraph. According to your considered? experience Are special rules to check If the answer is YES, please 5 the permissible horizontal add the values of the YES tolerances (sags) of the permissible of the horizontal
bracing members? tolerances, or indicate the NO references. 6 Is necessary to adjust locally the angle legs of the YES horizontal members of the tower to assure a perfect
horizontal connection of the NO bracing members to the horizontal members 7 The forces for the designing Please comment if the forces calculated of the bracing members are: are calculated or assumed. assumed neglected 8 The bracings for the towers If yes, please comment. YES proposed to be tested, are different in case of non NO tested towers? 9 The bracings of the heavy If yes, please comment. loaded towers (for example YES Angle Tower Types) are different comparing to the
bracings of the normal NO towers (for example Suspension Tower Types)? 10 Are you using the geometry - If YES, please mark them. YES described in the Figure 1? - If NO, or OTHERS, please NO describe them (by drawing). OTHERS
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TABLE 2. QUESTIONNAIRE FOR USUAL GEOMETRIE’S OF BRACING Type A, B, You use the proposed The proposed dimensions Comments Dimensions geometry? are OK? YES YES NO, we use it for … m 1 < 2 m NO - YES, together to the others YES NO, we use it for … m YES YES NO, we use it for … m 2 2 … 2.5 m NO - YES, together to the others YES NO, we use it for … m YES YES NO, we use it for … m 3 2 … 4 m NO - YES, together to the others YES NO, we use it for … m YES X YES NO, we use it for … m 4 3 … 5 m NO -
YES, together to the others YES NO, we use it for … m YES YES NO, we use it for … m 5 4 … 8 m NO
YES, together to the others YES NO, we use it for … m YES X YES NO, we use it for … m 6 5 … 8 m NO
YES, together to the others YES NO, we use it for … m YES YES NO, we use it for … m 7 5 … 8 m NO - YES, together to the others YES NO, we use it for … m YES YES NO, we use it for … m 8 7 … 12 m NO - YES, together to the others YES NO, we use it for … m YES YES NO, we use it for … m 9 10 … 16 m NO - YES, together to the others YES NO, we use it for … m
Annex 1 : Page 2 of 2 CIGRE / SC 22 / WG08-Task Force 3-3.3 Page.3 HORIZONTAL BRACINGS OF THE TOWERS No.
Type A, B, You use the proposed The proposed dimensions Comments Dimensions geometry? are OK? YES YES NO, we use it for … m 10 10 … 16 m NO -
YES, together to the others YES NO, we use it for … m YES YES NO, we use it for … m 11 12 … 20 m NO
YES, together to the others YES NO, we use it for … m YES X YES NO, we use it for … m 12 14 … 20 m NO
YES, together to the others YES NO, we use it for … m
TABLE 3. QUESTIONNAIRE FOR THE BRACINGS PHILOSOFY OF GUY TOWERS Comments to the Comments No Questions Questions Answers to the Answers 1 Who is in charge to establish Please comment the reasons. the Consulting the levels were is necessary Engineer to provide the horizontal the Tower bracings for guy towers? designer both 2 What is the maximum If the standard exist, please According to your distance (in meters) between attach a copy of the relevant standard two horizontal bracings to be paragraph. According to your considered? experience
Completed on: ______by: ______Company: ______Country: ______
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ANNEX 2
DETAILED RESPONSES TO QUESTIONNAIRE
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ANNEX 2
Detailed Responses to Questionnaire
Responses were received from the following individuals / organisations:
No Company Country 1 Tractebel Belgium 2 ABB Brazil 3 Fichtner Romelectro Engineering Romania 4 IVO Transmission Engineering Ltd Finland 5 C.C.G. Edelca Venezuela 6 Norconsult AS Norway 7 Balfour Kilpatrik Ltd England 8 PB Power Limited. (Kennedy and Donkin Division) England 9 PB Power Limited. (Merz and McLellan Division) England 10 Electromantaj Romania 11 ABB Brazil 12 ABB SAE Italy 13 ABB Australia 14 ESKOM South Africa 15 National Grid Company Plc England 16 STFA Enerkom A.S. Turkey 17 EDF – CNIR France 18 Fichtner Romelectro Engineering Romania 19 Powerlink Queensland Country Australia 20 Linuhonnun Iceland 21 Bonneville Power Administration USA
The following give the actual answers received from the respondents.
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QUESTION 1 : : Self Supporting Who is responsible for the choice of principles employed for diaphragm design? The Consulting Engineer, the Tower designer or both? Please comment the reasons.
Answers to Question 1 Comments to the Answers Num Consultant Designer Both 1 X In some cases, the Consulting Engineer can require supplementary levels of horizontal bracings to provide transverse stability 2 X In many cases, however, the levels where the horizontal bracings shall be installed are defined by specifications and/or Consultant Engineers. 3 X 4 X 5 X 6 X No standard. 7 X 8 X If a tower is successfully tested, I don’t believe that the Engineer can reasonably insist that the Tower Designer add additional plan bracings 9 X The designer selects and the consultant accepts (unless the Consultant is the designer ) 10 X Consulting Engineer: to specify the plane bracing in case of special supports (crossings, high antenna). The Support Designer: for common self-supporting supports. 11 X 12 X Usually the Support Designer is the one who chooses the levels for the horizontal bracings based on the good practice and design standard. 13 X
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Answers to Question 1 Comments to the Answers Num Consultant Designer Both 14 X The line specifications sent by the Consulting Engineer most of the times indicate the attachment heights, leg & body extensions leaving little choice to the support designer. 15 X The design needs to consider stability during tower erection and under extreme wind loading 16 X 17 X The person who looks at the results 18 X The Consulting Engineer should specify the levels of the supports to be reinforced with horizontal bracings as minimum requirements for supports stability and the maximum distance between the levels. The Designer should provide all necessary bracings to guarantee the support stability. 19 X 20 X Standards does not explain why horizontal is needed, therefore it is usually the tower designer who is in charge. The Consulting Engineer should check the result. 21 X Standards do not explain why horizontal bracings are needed therefore it is usually the support designer who is in charge. The Consulting Engineer should check the result.
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QUESTION 2 : Self Supporting Are there any special loading cases considered for diaphragm design not addressed by standards? Yes or No? If the answer is YES, please describe the loading cases, or indicate the references.
Answers to Question 2 Comments to the answers Num Yes No 1 X The horizontal angles of the square on the outside of the diaphragm are calculated in accordance to the “Recommendations for angles in Lattice Transmission Towers” issued by the European Convention for Structural Steelwork (ECCS) in January 1985 as well as to the European prEN 50341 The diaphragm is calculated with a hypothetical local horizontal load of 1.5 kN multiplied by the distance (in m) between the main legs, applied to the centre of the outside angles (practical rule), in order to stabilise the diaphragm against overall buckling for wind forces applied on the diaphragm Each angle of diaphragm is also calculated with a vertical load of 1 Kn (weight of a man) applied in its centre 2 X We agree exactly with the comments & arguments explained in ASCE Manual Pub. No. 52 3 X 4 X 5 X 6 X 7 X E.g. extreme wind loading 8 X Wind acting directly on horizontal members and attached redundant members. A gust factor appropriate to the largest dimension of the relevant area of the tower face should used. A case controlling the stiffness of the horizontal members may also be necessary. 9 X There should be 10 X Consideration of transverse brace force (ref. ECCS – International Colloquium on Stability) Ice / snow and temperature loading. Ice / snow give increase of bending and torsion moments on horizontal members (Japanese code / reports) 11 X
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Answers to Question 2 Comments to the answers Num Yes No 12 X 13 X 14 X 15 X Bracing needs to resist wind loading on the frame it stabilizes. It must also be designed to resist forces imposed by men whilst erecting or maintaining it. 16 X Torsion case. Broken wire case. 17 X Contract values depending on the internal force in the connect member 18 X Wind acting on individual angles of bracings. Ice or snow acting on bracings 19 X Capacity of all horizontals to support the weight of man of 1.4 kN 20 X We have calculated the forces based on the “good rule of practice”, i.e. use concentrated horizontal load of 1.5 kN * L placed in the middle of the horizontal member. L = length of the horizontal edge member in m. Deflection under this force shall not exceed L/1000. See ENV 1993-3-1 and draft for CLC/TC (SEC) 50 (CENELEC). 21 X
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QUESTION 3: Self Supporting Is the Consultant / Designer’s experience considered adequate to establish the levels / heights at which diaphragms should be employed? Yes or No? If the answer is no, please describe the calculus to be provided, or indicate the references.
Answers to Question 3 Comments to the answers Num Yes No 1 X The tower designer has also to consider (temporary) tower stability during erection 2 X As a general rule, the levels are established as a combination of design experience & erection / construction practice. 3 X 4 X 5 X 6 X 7 X 8 X Bracing are necessary at every bend-line, and at horizontals where the width is too great to allow a horizontal of reasonable size to span across the face. 9 X Previous testing experience has indicated a slim tower requires intermediate support every 15m 10 X Except in cases where non-linear analysis needs to be considered 11 X Also construction practice is desirable 12 X 13 X 14 X 15 X 16 X 17 X 18 X In case of special design (supports for big crossings) where nonlinear analysis needs to be considered
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Answers to Question 3 Comments to the answers Num Yes No 19 X 20 X It should be clearly stated in standards based on real criteria’s. Looking at towers from different designers it is obvious that they apply different criteria. Is the experience different? As a consultant/designer we have established the level of horizontal planes for quite many towers but we have limited knowledge of secondary forces in towers as a results of the level, i.e. distortional effects. Use of experience is enough only in the case of designing a tower after tower test of similar structure using same/similar standard. 21 X There is no in-house (BPA) standard for the requirement of horizontal bracing. For new designs BPA would consider the ASCE Lattice Tower Standard.
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QUESTION 4: Self Supporting What is the maximum distance allowed between two horizontal diaphragms? According to your standard or According to your experience? If the standard exists, please attach a copy of the relevant paragraph.
Answers to Question 4 Comments to the answers Num Standard Experience 1 X Experience instead of a standard. An horizontal bracing is installed: - On the attachment level of each cross-arm (against torsion loads due to broken conductors); - On each crank in the main leg (to resist to the horizontal load component); - Above each K-bracing (to stabilise the centre point of the edge members) 2 X We adopt 15 m as maximum distance between 2 subsequent diaphragms levels. As general rule we specify them in the following circumstances: - In the cross-arms levels - In the support waste - In any level of changing in the slope of the support legs - In the base level of the supports (just above the feet) - At maximum each 15 m between anyone described above. 3 X According to our standard: 12 m. According to our experience: 18 m 4 X 5 X 6 X 7 X ASCE 10-90 Cl 4C.3 recommends 23m for towers taller than 61m or heavy dead end structures. Approx 20m as practical maximum. 8 - - I do not believe a definite rule is appropriate.
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Answers to Question 4 Comments to the answers Num Standard Experience 9 X about 15m - maximum but required at bend lines on levels requiring out-of-plane support (i.e. top of hillside leg extensions) 10 X According to Romanian code (PE 105): 8 to 12 m. With consideration of requirements to maintain face alignment and check of general stiffness 18 to 25 m (ASCE 10/90 recommendations : 23 m). 11 X According to the standard: ASCE; 75 ft. According to our experience: 20 m. 12 X According to our experience: 10 … 12 m. 13 X According to our experience: 15m 14 X The bracings have been placed without any limitation to suit the line parameters. 15 X I would provide a plan bracing at no more than 12 metres vertical spacing 16 X 17 X No general rule, depending on the type of support 18 X According to our rules : 8 … 12 m; in ice areas: 15 … 18 m 19 X Governed by l/r of member selected 20 X Experience 15- 20 m. The only specification we are aware of is in the Manual 52, i.e. 22.5m. We have used this value as a maximum distance between planes. 21 X Reviewing existing designs it was determined that the maximum distance between horizontal bracing was 30 m (100 feet). This was used in a 150 meter (500 feet) river crossing support. Based on the new ASCE 10 Lattice Support Standard, BPA would try to use 22,5 meters (75 feet). The shortest distance is 12 ... 18 meters (40 ... 60 feet).
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QUESTION 5: Self Supporting Are any special design rules employed to check the permissible sag of the diaphragm members when erected? Yes or No? If the answer is YES, please add the values of the permissible of the horizontal tolerances, or indicate the references. Answers to Question 5 Comments to the answers Num Yes No 1 X The deflection of the edge member must be lower than 1/1000 of its length 2 X According to ASTM A6 3 X 4 X 5 X ASCE 10-90 Redundant members 6 X 7 X Unless specified. Bending stresses due to weight & external loads should be checked. 8 X In the UK it is normal to design for a vertical point load of 1.5 kN on all horizontals. This is likely to prevent the selection of members with excessive sag. 9 X By default the a maximum slenderness ratio of 250 is specified together with a capability of withstanding a center point load - review requires the plan bracing to be “locked off” – i.e. not a mechanism. 10 X Check calculations to be provided if dead load sag exceed “e” angle profile value 11 X ASTM A6 12 X Usually we use the limit on the slenderness (λ < 250) and a resistance check under a loading condition with a man walking on the horizontal member 13 X Support sags checked at prototype and testing stage 14 X 15 X Horizontal members should meet the “normal” stiffness checks for all other members. Limiting slenderness to 200 is usually sufficient 16 X 1/500, DIN 18800
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Answers to Question 5 Comments to the answers Num Yes No 17 X The bracing should resist to a contract value depending on the internal force in the connect member, connection type, support use 18 X 19 X 20 X Slenderness crit eria should be enough combined with man loading criteria. 21 X Only the limits placed on the L/r values: 200 for compression members in a horizontal bracing system
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QUESTION 6: Self Supporting During fabrication, is it necessary to adjust member flange angles to ensure a flush connection between the flanges of the diaphragm and the horizontal support face members? Yes or No?
Answers to Question 6 Comments to the answers Num Yes No 1 X The connection between the edge member and the horizontal bracing is assured by a bent gusset plate if the vertical angle exceeds 3° 2 X The horizontal bracings are for the erection/construction people to provide temporary stability during the erection of the lower part of the support (legs). As well as to check the exactor of the foundation construction/erection 3 X 4 X 5 X 6 X 7 X 8 X Yes, if the angle of the tower face exceeds about 2 degrees to vertical. 9 X During detailing it is expected the member connections are flush. Normally a 2º “error” is allowed before bending is required. 10 X For supports with double slope over 35-40% (correlated with the flange size of the main post angles) 11 X In some cases it is necessary to bend the angle leg to connect the bracing. 12 X 13 X 14 X 15 X It is easier to open the flange of the outside horizontal member than to close the flanges of all internal horizontal members 16 X 17 X
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Answers to Question 6 Comments to the answers Num Yes No 18 X 19 X 20 X 21 X Sloped washers are used to correct for the orientation of connected members
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QUESTION 7: Self Supporting What forces are employed for the member design of diaphragms? Are they calculated, assumed or neglected? Please comment if the forces are calculated or assumed.
Answers to Question 7 Comments to the Answers Num Calculated Assumed Neglected 1 X The diaphragm is calculated with a hypothetical local horizontal load of 1.5 kN multiplied by the distance (in m) between the main legs, applied to the centre of the outside angles (practical rule). Each angle of the diaphragm is also calculated with a vertical load of 1 kN (weight of a man) applied in its center 2 X X In the horizontal view of the crossarms they are calculated but those used to provide rigidity to the supports are assumed to be unloaded 3 X 4 X 5 X 6 X Based on slenderness and load requirements. 7 X E.g. extreme wind loading 8 X X See question 2. John Short’s paper recommends a horizontal load of .5kN per m width of the plan to be applied in the center of the plan for UK. 9 X Unless they come under the ASCE rule to carry a percentage of the load in the main member 10 X A result from: - statically calculation with consideration of the shear/torsion forces and horizontal components (for changes in main leg’s slope) - Transverse brace force - Slenderness ratio limitation and the weight of the worker + tools - Nonlinear analysis for special supports (big crossings, high antenna supports, etc). 11 X X Calculated at the crossarm level
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Answers to Question 7 Comments to the Answers Num Calculated Assumed Neglected 12 X X For the horizontal bracing level we calculate the forces, for the horizontal bracing on the body we assume the force in the member equal to 2 … 2.5 % the force in the main member. 13 X As % of vertical, and as man – load in center 14 X The bracing members are included in the analysis 15 X Horizontal frame must withstand wind loading on the face of all attached members 16 X 17 X 18 X X In case of snow the forces for the designing of the bracing members are calculated according to JEC-127 – Japan Standard . In case of weight of man, horizontal wind and percentage of the load in the main member, the forces for the designing of the bracing members are assumed. 19 X Where they are modeled in the analysis 20 X X X Forces should be calculated for inexperience designer, experienced designer soon knows appropriate dimensions and does not need minimum calculation. For smaller widths minimum profile sizes are usually sufficient. 21 X
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QUESTION 8: Self Supporting Are different design philosophies employed for the diaphragms in supports to be tested as opposed to those not to be tested? Yes or No? If yes, please comment.
Answers to Question 8 Comments to the answers Num Yes No 1 X No comment 2 X 3 X 4 X 5 X 6 X 7 X 8 X No. Tower designer should use same philosophy in either case. 9 X (unless the tower is a short one without body extension). 10 X 11 X 12 X 13 X 14 X 15 X 16 X 17 X Supports are always tested 18 X 19 X 20 X 21 X
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QUESTION 9: Self Supporting Are there any differences in the design and or type of diaphragm employed for heavy supports (angle, terminal) than for the relatively lighter suspension supports? Yes or No? If yes, please comment.
Answers to Question 9 Comments to the answers Num Yes No 1 X The stabilising members of an horizontal bracing are nominally unstressed, possibly with the exception of the edge members. If the edge members participate to the transmission of the forces, for instance from the tower top to the foundations, the other angles of the horizontal bracing, excluding the edge members have to constitute a rigid (isostatic or hyperstatic) horizontal lattice structure without any contribution from the edge members 2 X If the dimensions of the support cross sections are the same they are similar. Maybe the profiles for the heavy loaded supports are bigger. 3 X 4 X 5 X 6 X 7 X But wider body widths may require an alternative geometry. 8 X 9 X 10 X 11 X The bracing varies with support dimensions 12 X 13 X Generally due to the larger size of the supports – cannot use simple bracing 14 X 15 X 16 X
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Answers to Question 9 Comments to the answers Num Yes No 17 X Because internal forces in members are more important in tension supports than in suspension supports 18 X 19 X Mainly due to the plan dimension of the tower being larger 20 X X For towers required being more reliable than normal towers it is logical to make strong requirements to planes. Especially if design is based on that: a) compression-tension cross diagonals are assumed to be restrained at the centre of the cross in case of bars broken at the centre gusset. b) horizontal planes are not fully triangle. 21 X
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QUESTION 10 : Self Supporting Are you using the geometry described in Figure 1? Yes, No or others? (Figure 1 of the Questionnaire is reproduced on the following page). If YES, please mark them. If NO, or OTHERS, please describe them (by drawing).
Answers to Question 10 Other Comments to the Answers You use the shown geometry Y(es)/N(o) Num s O 1 2 3 4 5 6 7 8 9 10 11 12 1 X See following sketches Many types (see Figure 1). In all cases they shall be rigid, not deformable. We start with a single X for small dimensions and 2 Y X we use to create lattice on the contour of the support section. See following sketches 3 Y General “Yes” answer – no specifics given Type 7 is used for 3 to 10m widths as shown (see end of 4 Y Y Y Y N N N Y* N N N N N question) See following sketches No 7, No 5 Terminal towers 230 kV line, type 7 5 Y Y Y Y Y Y 400 kV line Type 7 Terminal tower type 5 6 Y Y Y Y Y Y Type 1 up to approx 3.5m 7 Y Y* Y Y N Y Y Y Y Y N Y Y Dimensions indicated for A & B are not treated as absolute max or min values.
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Answers to Question 10 Other Comments to the Answers You use the shown geometry Y(es)/N(o) Num s O 1 2 3 4 5 6 7 8 9 10 11 12 Generally similar, but locking members in the corners are sometimes omitted. See John Short’s paper. General Comment. Practice is not consistant in UK. Left to Tower designers discretion. Type 1 : Prefer type 2 without locking bar as makes erection easier 8 Y Y* Y Y N Y Y Y* Y* Y Y* N Y* Type 2 : As above Type 3 : Common in UK, sometimes without locking bar Types 4 & 11 : Not usual in UK Types 5 & 9 : Common in UK Type 6 : Very common in UK, sometimes without locking bars in corners. Types 7, 8, 10, 12 : Sometimes used 9 Y Y Y Y Y Y* Y Y Type 1 not preferred, usual to use type 2 or 3 Type 1 : < 1.5 m Type 2 : < 2.0 m Type 4 : < 3.5 m Type 6 :Unsupported length of horizontal members is correlated 10 Y Y* Y* Y Y Y Y* Y Y* Y* with specific panel bracing/redundant member’s arrangements. Type 7 : < 7.0 m Type 9 : (not stated) Type 11 : > 16.0 m. For special tower /built up members Type 12 : > 16 m 11 Y Y Y Y Y Y Y Y Y Y X together to others (please see Annex 1)
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Answers to Question 10 Other Comments to the Answers You use the shown geometry Y(es)/N(o) Num s O 1 2 3 4 5 6 7 8 9 10 11 12 12 Y Y Y Y Y Y The typical configurations we use 13 Y Y Y Y Y Y Y 14 Y Y Y Y Y Y Y Y Y Y We use types 1, 2, 3, 6 and 7 Other types used are similar to types 9 and 11 15 Y Y Y Y Y Y Y Y X We use Type 1 for loaded (Torsion) bracings Type 6 used for up to 17 metres 16 Y Y Y Y Y Y Y Y X See following sketches 17 Y Y Y Y Y Y 18 Y General “Yes” answer – no specifics given Type 1 < 3.0 m 19 Y Y* Y* Y Y Y Y Y Y Y* N N N Type 2 for 4+ m Type 9 for 25 + m
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Answers to Question 10 Other Comments to the Answers You use the shown geometry Y(es)/N(o) Num s O 1 2 3 4 5 6 7 8 9 10 11 12 Dimensions of plans depend partly on the steel grade used since the criteria “man loading” will often govern the member dimensioning. We often only use S355 (St.52) steel and therefore it is expected to see greater member sizes for some types of plans. Regarding preferred type of plan it is always the question of number of elements compared to weight. As a simple rule we often accept the use of L65x5 or L70x5 as the biggest members. If sizes go beyond these values we try to find 20 Y Y Y* Y* N Y* Y N N N N N N more economic solutions. Type 2 : We use it up to ca. 3,5-4,0m with tension members Type 3 : Width can be longer than 4m. Often we have closed square in the centre. Type 4 : We would usually prefer plan 3 instead of this one. Type 5 : We usually have a closed square in the centre. Sometimes we use two X-braces in front of each side of the square. Type 6 : We have used it for 9,5m Nº 1 for 2 ... 4 m 21 Y Y*Y* X Nº 2 for 10 ... 15 m; for the others, special design. See following sketches
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Figure 1: Proposed Arrangement of horizontal bracings
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The following diaphragm arrangements were provided in response to Question 10
Respondent 4 (Finland) Respondent 1 (Belgium) Type 7 is used for 3 to 10m
widths Typical Diaphragms
Respondent 16 (Turkey)
Typical Diaphragms Respondent 2 (Brazil) Typical Diaphragms
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Respondent 21 (USA) Typical Diaphragms
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QUESTION 11 : Guyed Towers Who is in charge to establish the levels where it is necessary to provide the horizontal bracings for the guyed towers? Please comment the reasons.
Answers to Question 11 Comments to the Answers Num Consultant Designer Both 1 - - - Neither; Guyed towers are not allowed by law in our country 2 - - - No response 3 X 4 X Constructional reasons, tower weight 5 - - - Not applicable 6 X No standard 7 X Because of the effect on the design of the mast. 8 - - - Guyed Transmission Towers are not used in UK, except for poles and lower voltage structures. 9 X The designer selects and the consultant accepts (unless the Consultant is the designer ) 10 X 11 X 12 X Usually is the support designer the choose the levels for the horizontal bracings based on the good practice and the design standard 13 X 14 X 15 X 16 X 17 X 18 X 19 X 20 X 21 - - - Do not use guyed supports
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QUESTION 12 : Guyed Towers What is the maximum distance (in meters) between two horizontal bracings to be considered? If the standard exist, please attach a copy of the relevant paragraph.
Answers to Question 12 Num According to According to Comments to the answers your standard your experience 1 - - Guyed towers are not allowed by law in our country 2 3 X At every trunk jointing 4 X Every body part has at least one horizontal bracing 5 - - Not applicable 6 X 7 X To suit mast design 8 Guyed Transmission Towers are not used in UK, except for poles and lower voltage structures. 9 X Previous designs have required a cross brace (single or double - if single alternating) at every joint level 10 X 3-6 m according to specific constructional details (size of cross section, built-up sections, etc) 11 X 9 m 12 X 6 … 8 m The guy supports need a global stiffness higher than self supported supports 13 X Generally at each body level, about 6 … 7 meters 14 X 33 meters 15 X 16 X 17 X No general rule 18 X 19 X Generally use 5.0m
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Answers to Question 12 Num According to According to Comments to the answers your standard your experience 20 X Typically the leg width is 1-1.5m and in that case there should be assumed to be some reduction in strength of leg members when maximum length between planes exceed 10- 12m. The reason for strength reduction is secondary effects from section distortion. In conversation with a Russian designer he informed that in Russia they use 12m as a maximum distance. They made some tests on effects of horizontal bracing and according to his memory it showed something about 10% reduction in capacity when the distance was up to 20m.Reduction of leg capacity of 1-2% for each 1m of plane distance exceeding 10m is a rule of thumb. 21 - - Do not use guyed supports
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ADDITIONAL CONTRIBUTIONS
The following additional contributions have been received from various sources after the issue of the questionnaire which are included to ensure this report is as complete as possible.
Response No 23 Slovenia IBE Consulting Engineers In addition to the Slovenian technical regulations U.l. SFRJ 65/88 (see Section 3.1.12 of the main report) the following design conditions typically apply for the design of diaphragms.
• Man weight, taken as 150 daN (with complete equipment), applied in the most unfavourable position on an element. • Maximum slenderness ratio of an element is 220.
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ANNEX 3
CALCULATION EXAMPLE FOR PLAN BRACING
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PLAN BRACING IN LATTICE SUPPORTS. CALCULATION EXAMPLE (By John Short – Former Chief Support Designer - UK National Grid Company)
1. GENERAL When K bracing are used in support it is usual to provide a plan brace to stabilize the center point of the K (and possibly the ¼ points for wider supports).
Theoretically, the compression half of the K Brace is stabilized by the tension in the other half, but in practice it is necessary to cater for traverse horizontal loads arising from wind on the steelwork, and economy usually indicates the desirability of some sort of girder on plan to stabilize the center and intermediate points of the horizontal.
2. DATA Various types of plan bracing are possible and the basic requirements are 1). strength–sufficient to cater for the applied wind forces, and 2). stiffness–sufficient to stabilize the main horizontal against overall buckling. In addition it is usual to cater for the weight of a man with suitable safety factor and the self weight of the steelwork. Arbitrary slenderness ratios are often imposed but this may be intended primarily to ensure the above requirements are met.
A nominal horizontal load of 15kN (representing the wind force) is assumed to act at the center of the horizontals. The factored weight of a man to be used is taken as 350 lbs. or 152kg. and the weight of steelwork the actual computed figure. An arbitrary slenderness ratio of 200 ensures that most angles will carry this load satisfactorily although some of the smaller angles will need to be restricted in length slightly, particularly when designed on the rectangular axis with bracing on the weak axis.
3. RESULTS The attached list of 6 types of plan bracing (figures A3.1, A3.2) has been calculated on the above basis.
For small base dimensions type 1 is lightest providing that an intermediate support to the horizontal is not required. If such a support is required type 4 and 5 may be considered suitable. At span of 8.5m and over, type 2 becomes the lightest.
In general, type 3 appears to give the lightest weight possible for all spans up to 16.5m but because of the small size of angles used for spans below 12.5m there is some doubt about the overall strength and stiffness of such spans. For example, the load in member B for the nominal 15kN horizontal force would be 60kN and this could only be safely resisted by a single bolt if M20 in minimum 6mm thick material were used and a minimum size of angle would be 100.65 to resist the load.
Accordingly type 3 is recommended only for spans over 12.5m where M20 bolts and larger size angles can be used economically.
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Type 6 is sometimes used for the larger spans to provide increased horizontal stiffness, but care has to be taken that the vertical stiffness is not reduced too much. If a deflection limit of 1/360 span is imposed for all members loaded with dead weight only the base widths shown in the table for type 6 may be used. Any increase above these figures rapidly produces excessive deformations as the deflection varies as the fourth power of the span. This type of bracing is recommended as being the most economic for base dimensions over 16.5m.
Table A3.1 shows a summary of economic types of bracing for various spans.
Table A3.1 - Economic types of bracing for various spans.
Summary Economic types of Notes bracing for various spans less than 4.0m Consider design without plan Care should be taken bracing. 4.0 – 8.5m Type 1 for support only at Type 1 does not provide a full diaphragm to center. maintain squareness of support. Extra bars “A” on remaining 2 sides would provide this if required. Type 5 for support at ¼ Type 5 is certainly most economic for ¼ points, points supports and the extra weight although types 2 & may be involved used for greater stiffness. in strengthening the main corner members to take M20 bolts is considered worthwhile. Type 4 does not appear to offer any economic advantages but provides the stiffest plan bracing of all the types and if 70x70x5 angles were available, could be recommended up to7.8m as being no heavier than type 5.
8.5 – 12.5m Type 2 (with M16 bolts for Type 2. The extra strength given by using economy M20 bolts with 80x60 angles may be or M20 bolts for strength). considered worthwhile for 7-10% increase in weight.
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Summary Economic types of Notes bracing for various spans 12.5 – 16.5m Type 3. Type 3. This type is the most economical from 9.4m upwards but is considerably less stiff and strong than type 2. As there is only a marginal weight advantage over type 2 from 9.4 – 12.5m it is recommended that type 3 is only used over 12.5 m. For absolute economy type 3 should be used at smaller spans with smaller angles but care must be taken not to overstress members or bolts under high wind conditions.
16.5 – 20.9m Type 6. Type 6. The main angles in this type are twice the length of those used for type 3 with lengths of 12.6-16.5m being longer than any other members used in support work. Splices are not recommended due to vertical deflection considerations, but if necessary should be positioned at the point of minimum bending stress, .09L from the hip bracing support position leaving a maximum length of 0.5L (i.e. 8.3 – 10.5m)
These calculations were performed until 1976. Since that date various types of bracings have been considered and the requirements have been modified slightly. This part of the paper provides the latest recommendations.
The basic requirement of strength and stiffness still remains; consequently the arbitrary slenderness ratio of 200 still applies to all members.
A nominal horizontal (uniformly) distributed load of 1.5 kN/m width of plan is assumed to act at the center of the horizontals. Additionally the bracing is to be capable of supporting the factored weight of a man (taken as 150 kg) standing anywhere on the completed plan-bracing in addition to the self-weight of the steel-work. In some types of bracing this load may be shared by two continuous members joined together at their cross-over point. In these cases the individual members must be capable of supporting 120 kg without considering the support of the other to enable the erection connection to be carried out safely.
A similar case arises where members are supported by hangers or hip bracing. All such members must be capable of supporting 120 kg plus self weight without the assistance of the hip bracing.
Loads in members should be restricted to the capacities of single bolts, M16 for 50mm flanges and below, M20 for 60mm flanges and above.
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Figure A3.1 - Diaphragms Types 1, 2 and 3.
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Figure A3.2 - Diaphragm Types 4, 5 and 6.
A further 4 type of bracing (from type 7 to type 10) are shown and are recommended but may be adjusted to suit different requirements of strength and stiffness (Figure A3.3).
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Figure A3.3 - Different types of Diaphragms suggested as the latest recommendations.
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ANNEX 4
DIMENSIONING OF DIAPHRAGM MEMBERS (Brazilian experience)
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Annex 4
DIMENSIONING OF DIAPHRAGMS MEMBERS (Brazilian experience)
Secondary members of the diaphragm are required to provide support to the main horizontal stressed members (Figure A4.1). To do so they are calculated in accordance to the following criteria:
- kL/r limited to 250, - They shall support loads of 2.5% of the axial load of the main supported horizontal members, - A vertical load of 1kN applied in the middle of the length shall be supported by the secondary members.
A – Main member B, C, D – Secondary members D
B C
A
Figure A4.1 - Typical horizontal bracing.
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