Performance of a Geogrid-Reinforced and Pile-Supported Highway Embankment Over Soft Clay: Case Study

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http://www.paper.edu.cn Performance of a Geogrid-Reinforced and Pile-Supported Highway Embankment over Soft Clay: Case Study

H. L. Liu1; Charles W. W. Ng, M.ASCE2; and K. Fei3

Abstract: This paper describes a case history of a geogrid-reinforced and pile-supported ͑GRPS͒ highway embankment with a low area improvement ratio of 8.7%. Field monitored data from contact pressures acting on the pile and soil surfaces, pore-water pressures, settlements and lateral displacements are reported and discussed. The case history is backanalyzed by carrying out three-dimensional ͑3D͒ fully coupled finite-element analysis. The measured and computed results are compared and discussed. Based on the field observations of contact stresses and pore-water pressures and the numerical simulations of the embankment construction, it is clear that there was a significant load transfer from the soil to the piles due to soil arching. The measured contact pressure acting on the pile was about 14 times higher than that acting on the soil located between the piles. This transfer greatly reduced excess positive pore water pressures induced in ¯ the soft silty clay. The measured excess pore water pressure ratio Bmax in the soft silty clay was only about 0.3. For embankment higher than 2.5 m, predictions of stress reduction ratio based on two common existing design methods are consistent with the measured values and the 3D numerical simulations. During the construction of the piled embankment, the measured lateral displacement–settlement ratio was only about 0.2. This suggests that the use of GRPS system can reduce lateral displacements and enhance the stability of an embankment significantly. DOI: 10.1061/͑ASCE͒1090-0241͑2007͒133:12͑1483͒ CE Database subject headings: Geogrids; Embankments; Clays; Monitoring; Three-dimensional models; Finite element method; Case reports.

Introduction 1. Embankment construction can be completed within a short time period; When highway embankments are constructed over soft soils, the 2. Embankment support piles reduce total and differential structure imposes a significant load over a large area. The soft settlements significantly; and clays and other compressible soils often bear intolerably large 3. The technique is suitable for various geological conditions. settlements or fail due to insufficient bearing capacity. A variety Conventional piled embankment construction ͑i.e., piled em- of techniques can be used to solve such problems ͑Magnan 1994͒. bankments without geogrid reinforcements͒ requires closely These include techniques to modify the embankment load on the spaced piles or large pile caps to transfer most embankment loads ground ͑lightweight materials, change in embankment geometry͒, techniques to improve the ground ͑preloading, surcharging, to the piles through soil arching. In order to place the relatively staged construction, excavation and replacement, stone columns, expensive piles as far apart as possible, a relatively inexpensive lime columns͒, techniques to accelerate consolidation ͑vertical geogrid material is included at the base of the fill. This geogrid- ͑ ͒ drainage and vacuum consolidation͒, techniques to reinforce the reinforced and pile-supported GRPS system has been used in ͑ ͒ road embankment fill ͑reinforcement͒, and techniques to provide several applications. Maddison et al. 1996 described an innova- additional structural support to the embankment ͑embankment tive system of ground improvement comprising vibroconcrete support piles͒. Each alternative has its own advantages and dis- columns and a load transfer platform incorporating low-strength advantages. The benefits associated with the use of embankment geogrids. The system was used to support a 6.0 m high embank- support piles are as follows: ment constructed over highly compressible peat and clay soils. Lin and Wong ͑1999͒ illustrated the use of mixed soil and cement

1 columns in an embankment to smoothen the differential settle- Professor, Geotechnical Research Institute, Hohai Univ., Nanjing ͑ ͒ 210098, China. E-mail: [email protected] ments of a bridge’s approaches. Han and Akins 2002 reported 2Professor, Dept. of Civil Engineering, Hong Kong Univ. Science and that vibroconcrete columns and geogrids were used for widening Technology, Clear Water Bay, Kowloon, HKSAR. E-mail: cecwwng@ an existing roadway. Based on the performance investigation of ust.hk 13 pile-supported and geogrid-reinforced earth platforms, Han 3Associate Professor, Geotechnical Research Institute, Yangzhou and Gabr ͑2002͒ recommended that area ratio could be reduced to Univ., Yangzhou 225009, China. E-mail: [email protected] 10–20%, in comparison with the relative high area ratio of con- Note. Discussion open until May 1, 2008. Separate discussions must ventional piled embankments ͑50–70%͒. be submitted for individual papers. To extend the closing date by one This paper presents a case history of a GRPS highway em- month, a written request must be filed with the ASCE Managing Editor. bankment project in which a low improvement area ratio of 8.7% The manuscript for this paper was submitted for review and possible publication on December 21, 2005; approved on February 28, 2007. This was used. The study of the embankment performance is based on paper is part of the Journal of Geotechnical and Geoenvironmental field measurements of pressures acting on the piles and soil sur- Engineering, Vol. 133, No. 12, December 1, 2007. ©ASCE, ISSN 1090- faces between piles, pore-water pressures, settlements and lateral 0241/2007/12-1483–1493/$25.00. displacements. The field measurements are compared with com-

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Fig. 1. Soil proﬁles and properties

puted results from three-dimensional fully coupled finite-element 1. A temporary double-wall casing is driven into the ground by backanalyses. a vertically vibrating driving machine. The double-wall casing consists of two concentric 8 mm thick steel pipes with different diameters. The outer and inner diameters of pipes Site Conditions are 1.016 and 0.76 m, respectively. This creates a 120 mm thick annulus between the outer and inner pipes for concret- The site is located in a northern suburb of Shanghai, China. The ing. The inner pipe is open-ended whereas the annulus is profile of the soil is as follows: there is a 1.5 m thick coarse- fitted with a temporary conical-shaped driving shoe, which grained fill overlying a 2.3 m thick deposit of silty clay; this can be detached by wet concreting pressure during concret- deposit overlies soft silty clay that is approximately 10.2 m thick. ing. During driving the casing, soil is displaced into the inner Underneath the soft silty clay is a medium silty clay layer that is pipe and outside the outer pipe. This creates a 120 mm about 2 m thick followed by a sandy silt layer. The ground water thick annulus between the outer and inner pipes for in situ level was at a depth of 1.5 m. Fig. 1 summarizes the available concreting. detailed test data, including the water content, the unit weight, 2. During in situ concreting the annulus, the casing is with- and the vane shear strength to a depth of about 24 m below drawn at a steady rate of 0.8–1.2 m/min. An appropriate ground level. The soft silty clay layer has a low to medium plas- concrete head varying from 0.3 to 0.5 m is always main- ticity, and a liquidity index IL of 1.2. Its water content ranges tained within the annulus to provide stability, whereas the between 40 and 50% and is generally close to the liquid limit. The casing is withdrawn. uppermost coarse-grained fill layer has a relatively high precon- 3. After withdrawing the double wall pipe pile, a concrete plug solidation pressure, in comparison with the underlying soft silty is constructed by replacing the top 0.5 m of soil column in- clay, which is normally consolidated or lightly overconsolidated. side the original inner pipe with concrete. This is to repair The undrained shear strength of the soft silty clay layer as mea- any possible damage to the top part of the annulus pile sured by the field vane has a minimum value of about 10 kPa at a depth of 3.8 m and increases approximately linearly with depth. caused by withdrawing the double-wall steel casing. By considering the previous installation procedures of the annulus concrete pile, it is believed that the influence of the pile installation to the adjacent ground should be similar to that of a driven Geogrid-Reinforced and Pile-Supported Embankment steel pipe pile since soil is displaced during the installation and the effects of in situ concreting on the adjacent ground should be The embankment was 5.6 m high and 120 m long with a crown minimal. This is because in situ concreting takes place inside the ͑ width of 35 m. The side slope was 1 V to 1.5 H. The fill material annulus of the double-wall steel casing i.e., concrete is not in ͒ consisted mainly of pulverized fuel ash with a cohesion of contact with the adjacent soil before the casing is withdrawn. 10 kPa, an angle of friction of 30°, and an average unit weight of The annulus concrete piles were placed in a square pattern at a 18.5 kN/m3. A cross-section view of the test embankment and the distance of three times the pile diameter ͑3m͒ from the center to locations of the instruments are shown in Fig. 2. the center of the adjacent piles. The area ratio, defined as the The embankment was supported by cast-in-place annulus con- percent coverage of the pile ͑caps͒ over the total foundation area, crete piles that were formed from a low-slump concrete with a was 8.7%, which was close to the lowest limit suggested by Han minimum of compressive strength of 15 N/mm2. The annulus and Gabr ͑2002͒. One layer of a biaxial polypropylene grid concrete piles were 16 m in length and were founded on a rela- ͑TGGS90-90͒ was sandwiched between two 0.25 m thick gravel tively stiffer sandy silt layer ͑see Fig. 2͒. The outer diameter of layers to form a 0.5 m thick composite-reinforced bearing layer. each pile was 1.008 m and the thickness of the concrete annulus The tensile strength in both directions ͑longitudinal direction and was 120 mm. The design capacity of the pile was 600 kN and the transverse direction͒ of the geogrid is 90 kN/m and the maximum key installation procedures of the pile are summarized as follows. allowable tensile strain is 8%.

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Fig. 2. Cross section of instrumented test embankment

Details of Embankment Construction construction history. The embankment was constructed to a height of 5.6 m over a period of about 55 days. In total, 740 piles were installed with tolerances of 150 mm in plane and 1% vertical alignment. Following the installation of the piles, a 0.25 m thick cushion layer of well-graded compacted Instrumentation gravel was laid over the piles to provide a working layer and to prevent the lower geogrid from mechanical damage above the pile To verify the design assumptions and to monitor the performance heads. This layer was compacted using a light weight road roller. of the embankment, various instruments were installed on site. Vibration was not used in the compacting to minimize the risk of Fig. 2 shows an instrumented cross section including two inserts damaging the heads of the unreinforced piles. One layer of the illustrating the detailed plan view and sectional view around Pile TGGS90-90 biaxial geogrid was placed as an interlock with the A, which was located near the center line of the embankment. The granular ﬁll. Another 0.25 m thick cushion layer was placed on installed instruments included the following. the top of the geogrid. Thus, the height of the whole geogrid- 1. Earth pressure cells measuring the vertical load shared by the reinforced bearing layer was 50 cm. At the edges of the embank- piles and the surrounding soil. Ten pressure cells were placed ment, the geogrid was wrapped up and anchored back into the around Pile A. Earth Pressure Cells E1–E8 ͑measuring range: embankment over a 5 m length. Fig. 3 shows the embankment 0–0.3 MPa͒ were installed on the surface of the surrounding soil; E9 and E10 ͑measuring range: 0–1.0 MPa͒ were installed at the head of the pile. 2. Surface settlement plates. Four settlement plates were installed. The settlement plates were installed on both the top of the pile heads ͑S1 and S4͒ and on the surrounding soil ͑S2 and S3͒. S1 and S2 were located at near the shoulder of the embankment. S3 and S4 were located near the center of the embankment. 3. Subsurface settlement gauges. The settlement gauges ͑SS͒ were installed to a depth of 24 m and at 2 m vertical inter- vals near the center line of the embankment. 4. Vertical inclinometer. The inclinometer casing ͑I1͒ was installed at a distance of 1.5 m from the embankment toe. The lengths of the inclinometer above and below the ground were 1 and 24 m, respectively. Fig. 3. Embankment height versus time 5. Pore-water pressure piezometers. The piezometers were in-

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Fig. 4. Plate-loading tests on foundation soil Fig. 5. Results of full-scale loading tests stalled at a depth of 4 m ͑P1͒ and 8 m ͑P2͒ midway between two piles near the center of the embankment. may be determined as the maximum test load which was about All the instruments were installed after the construction of the 1,380 kN, whereas the maximum measured test load of the piled- piles but before the building of the embankment. The earth pres- raft was about 2,500 kN or 278 kPa. Judging from the load– sure cells, the inclinometer, the subsurface settlement gauges and settlement curve in Fig. 5, it is clear that this measured maximum the piezometers were monitored using portable readout boxes. test load is far from the ultimate capacity of the piled-raft. By The surface settlement plates were located and monitored using comparing with the ultimate bearing capacity of the foundation surveying techniques. The field monitoring program was termi- soil ͑i.e., 80–90 kPa͒, it is found that the ultimate bearing capac- nated 180 days after the commencement of the construction of ity of a pile-raft should be at least three times larger when a pile the embankment ͑or 125 days after the completion of the is constructed. embankment͒. An increase in the ultimate bearing capacity of a soft soil using different ground improvement techniques ͑stone columns, granular piles, lime or cement columns͒ have been investigated by Full-Scale Field Loading Tests many researchers. Bergado and Lam ͑1987͒ carried out full-scale load tests and reported that the use of granular piles in soft Four plate-loading tests ͑P1, P2, P3, and P4͒ were performed at Bangkok clay could increase the bearing capacity of the clay by various locations at the site to determine the ultimate bearing up to four times. Kitazume et al. ͑1999͒ performed a series of capacity of the foundation soil. The steel loading plate was centrifuge model tests to investigate the increase in the bearing 0.5 mϫ0.5 m. Each test was carried out in a pit of 1.5 m depth capacity of soft ground improved with deep mixing columns at on the silty clay. The plate-loading tests were conducted accord- relatively low area ratio. Chen et al. ͑2002͒ carried out field loading to the method prescribed in GB 50007-2002 ͑Ministry of ing tests on 9.0 m deep mixing columns with two different area Construction 2002͒. Fig. 4 shows the pressure versus settlement ratios. Malarvizhi and Ilamparuthi ͑2004͒ conducted load tests on relationships for all four plate-loading tests. The measured results a soft clay bed stabilized with stone columns in the laboratory. are fairly consistent and the ultimate bearing capacity of the clay These reported cases and the field measurement from this study may range from 80 to 90 kPa. By using Terzaghi’s bearing capac- are compared in Fig. 6. It can be seen that although the data are ity equation ͑Terzaghi 1943͒, the backanalyzed undrained shear fairly scattered, the measured capacity is generally improved with strength of the clay is about 15 kPa, which is consistent with the an increase in area ratio. Comparing with deep cement mixing measured average undrained shear strength of the silty clay from piles and columns, the use of the cast-in-place concrete annulus vane shear tests ͑as shown in Fig. 1͒. piles in this study seems to provide considerably larger enhance- In addition to the four plate-loading tests on the foundation clay, two loading tests on piles were carried out including a static loading test on a 16 m long single pile and a plate-loading test on two 3 mϫ3 m sandwiched steel plates ͑i.e., 30 mm thick each separated by 200 mm high Ibeams͒ overlying a layer of 0.5 m thick densely compacted gravel, which was in turn supported by a 16 m long pile installed in the center of the area ͑or may be called a “pile-raft” test͒. These two test piles were the same as those used for the construction of the embankment. In the loading test of the single pile, the pile was loaded by using a hydraulic jack against a Kentledge formed by steel beams and concrete blocks. The jacked force to the pile head and the displacement of the pile head were measured by means of a load cell and dial gauges, respectively. In the pile-raft test, the 3ϫ3 m steel plates were loaded by jacking against concrete blocks. This plate-loading test may be regarded as the determination of the equivalent capacity of the “piled raft.” Fig. 5 shows the loading test results of the single pile and piled-raft. The axial load capacity of the single pile Fig. 6. Bearing capacity increase factor versus area ratio

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Fig. 7. ͑a͒ Plan view of the analysis zone; ͑b͒ ﬁnite-element mesh

ment of the ultimate capacity at similar area ratios. This is be- elements within each lift. Owing to the relatively high permeabil- cause the annulus concrete pile has larger stiffness than that of ity, the embankment fill, reinforced gravel layer and the surface soft or semi-hard columns ͑stone columns or cement columns͒. coarse-grained fill are assumed to behave in a drained manner. On Other field monitoring results are reported and discussed later in the other hand, the piles are considered to be fully impermeable. this paper. Hence, a 20-node quadratic brick with reduced integration elements and without pore pressure degrees of freedom were used for these materials. The other soils are modeled by 20-node qua- Three-Dimensional Numerical Analysis dratic displacement, linear pore pressure ͑with eight more excess pore pressure degrees of freedom at the corners͒, and reduced Finite-Element Discretization and Boundary Conditions integration elements. The geogrid is modeled by eight-node quad- rilateral, reduced integration membrane elements that have a ca- The piled embankment shown in Fig. 2 is truly three-dimensional pacity to resist only tensile force. as each pile is not continuous in the out-of-plane direction. How- ever, by taking planes of symmetry, it is possible to analyze a three-dimensional slice only. The numerical analyses were per- Material Model and Parameters formed using a finite-element software called ABAQUS ͑HKS The pile was modeled as an isotropic linear elastic material with 1997͒. Fig. 7 shows the finite-element mesh used in the analyses. a Young’s modulus of 20 GPa, and a Poisson’s ratio of 0.2. The In Fig. 7, the soft foundation is 25 m deep and with a rigid and embankment fill ͑i.e., PFA͒, gravel and the surface coarse-grained impermeable layer underneath it. The horizontal length of the fill were modeled using a linear elastic-perfectly plastic model finite-element mesh is 77.7 m, which is three times the width of with Mohr–Coulomb failure criterion. The Mohr–Coulomb model half the embankment base so that the boundary effect can be Ј ␸Ј minimized. At the bottom of the finite-element mesh ͑z=0 plane͒, requires five parameters: effective cohesion, c , friction angle, , ␺ the displacements are set to zero in the three directions, x, y, and dilatancy angle, , effective Young’s modulus, E, and Poisson’s ␯ z. The displacements in the x direction are set to zero on the ratio, . The parameters used in the analyses are summarized in center line of the embankment ͑x=0 plane͒ and the far field lat- Table 1. These parameters were taken from the typical measured eral boundary ͑x=77.7 plane͒. In addition, symmetrical condi- values in Shanghai. The angle of dilation was assumed to be 0°. tions imply zero displacement in the y direction for nodes on the The four foundation soils shown in Fig. 2 were modeled as two vertical planes ͑y=0 and y=1.5 plane͒. With regard to the modified cam clay materials. The modified cam clay model re- drainage boundary conditions, the water table is assumed to be at quires five material parameters for each soil type, slope of the a depth of 1.5 m below ground level ͑z=23.5 plane͒ and the ini- virgin consolidation line, ␭, slope of the swelling line, ␬, the void tial pore pressures prior to the embankment construction are taken ratio at unit pressure, e1, slope of the critical state line, M, and to be hydrostatic. The bottom of the finite-element mesh is de- Poisson’s ratio, ␯. Values for ␭ and ␬ were obtained from one- fined as impermeable and the lateral flow is not permitted across dimensional compression tests in oedometer. The values of e1 the x=0, x=77.7, y=0, and y=1.5 planes. were deduced, using the modified cam clay model, from the po- The embankment construction is simulated in nine lifts ͑in- sition in the compression plane of the one-dimensional line found cluding the composite reinforced bearing layer of 0.5 m͒ by acti- in with an oedometer test. The values of M were obtained from vating the elements and “turning on” the gravity forces for the undrained triaxial tests with pore pressure measurements. The

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Table 1. Parameters for FEM analysis ␸ ␺ ϫ −4 Drainage cЈ Ј Ј E kw 10 ͑ ͒ ͑ ͒ ͑ ͒ ͑ ͒ ␯␭ ␬ ͑ ͒ Material Model condition kPa deg deg MPa Me1 m/day Embankment MC Drained 10 30 0 20 0.30 Gravel MC Drained 10 40 0 20 0.30 Coarse-grained fill MC Drained 15 28 0 7 0.30 Silty clay MCC Consolidation 0.35 0.06 0.012 1.20 0.87 8.64 Soft silty clay MCC Consolidation 0.40 0.15 0.030 0.95 1.79 4.32 Medium silty clay MCC Consolidation 0.35 0.05 0.010 1.10 0.88 4.32 Sandy silt MCC Consolidation 0.35 0.03 0.005 0.28 0.97 43.2 Note: MCϭMohr–Coulomb and MCCϭmodified cam clay. overconsolidation ratios for these four soil layers were taken to be measured pressures on the pile agree better with the computed unity. The values of the earth pressure coefficients at rest, K0, are values, the maximum error percentage is 7%. It can be seen that a related to ␸Ј by majority of embankment load was carried by piles. As E5 and E8 were damaged during the embankment construction, no data from K =1−sin␸Ј ͑1͒ 0 these two cells can be reported. A summary of the parameters adopted is given in Table 1. For the It is clear that there was load transfer from the soil to the pile soil parameters used, the corresponding theoretical values of und- as a result of soil arching in the embankment fill. The load trans- rained shear strength by the modified cam clay model as repre- fer mechanism may be quantified by using a stress concentration sented by the following equation, are plotted as the solid line in ratio, which is defined as the ratio of an average vertical stress on Fig. 1 for comparisons: top of a pile to an average vertical stress applied on the foundation soil. Typical reported values of stress concentration ratio for M ͑1+e ͒ − ͑␭ − ␬͒ln 2 1+e ͩ 1 0 ͪ ͑ ͒ piled embankments ͑without geogrid reinforcements͒ range from cu = exp − 2 2 ␭ ␭ 1to8͑Barksdale and Goughnour 1984; Greenwood 1991͒. How- where e ϭinitial void ratio. ever, the measured stress concentration ratio in this current study 0 is about 14, which is higher than those of the piled embankments The coefficients of permeability, kw, are also given in Table 1. These values were assumed to be typical values for the soils in without the use of geogrids. The larger stiffness difference Shanghai. between the concrete pile and foundation soil than the stiffness The geogrid was modeled as an isotropic linear elastic material difference between stone columns and foundation soil may also with a tensile stiffness of 1,180 kN/m and a Poisson’s ratio of result in the higher stress concentration ratio. 0.3. Interface elements were used to model the interaction behav- ior between the gravel and the geogrid. The interface yield stress was determined by the Mohr–Coulomb failure criterion with zero cohesion. The interface friction angle was assumed to be equal to the friction angle of the gravel, as given in Table 1.

Comparisons of Measured and Computed Results

Load Transfer from the Soil to the Pile

Pressure Acting on the Pile Head and the Soil Surface Fig. 8͑a͒ shows the pressure acting on the soil surface between the piles measured by the earth pressure cells. The embankment load is also included in Fig. 8 for reference. As shown in Fig. 8, there is a marked reduction in actual stress applied to the foundation soil. When the embankment height increased to 5.6 m, i.e., the embankment load was about 104 kPa, the measured pressures acting on the soil surface increased by about 31–58 kPa, which is only 30–60% of the embankment load. On the other hand, the pressure acting on the pile head increased to 674 kPa ͓see Fig. 8͑b͔͒, which is about 6.5 times larger than the embankment load. Based on the measured pressures, the load carried by a single pile was estimated as to be 530 kN. By using the measured maximum test load as the capacity of the pile ͑i.e., 1,380 kN as estimated from Fig. 5͒, a factor of safety of 2.6 was obtained. The measured and computed pressures at the end of the embankment construction are compared and summarized in Table 2. ͑ ͒ At the soil surface, the maximum difference between the mea- Fig. 8. Measured pressure a acting on the soil surface between ͑ ͒ sured and the computed contact pressures is 46%. However, the piles; b acting on the top of the pile

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Table 2. Comparison between Measured and Computed Pressures at the End of Embankment Construction E1 E2 E3 E4 E5 E6 E7 E8 E9 E10 Measured ͑kPa͒ 31.2 38.7 36.7 31.4 — 47.9 56.0 — 583.6 552.2 Computed ͑kPa͒ 53.6 48.9 54.2 58.1 58.8 53.6 48.9 54.2 592.6 592.6 Difference ͑%͒ 42 21 32 46 — 11 15 — 2 7

Stress Reduction Ratio and Hewlett and Randolph ͑1988͒ are consistent with the mea- The degree of the soil arching may be quantified using the stress sured values and the computed results from the three-dimensional ͑ ͒ reduction ratio, S3d, which is defined as follows Low et al. 1994 : finite-element analysis. It should be noted that the influence of 0.5 m thick gravel layer p r ͑ ͒ S3d = 3 on soil arching is not included in the calculations by using the ␥H methods proposed by Russell and Pierpoint ͑1997͒ and Hewlett ϭ ␥ϭ and Randolph ͑1988͒. where pr pressure applied on foundation soil; unit weight of embankment fill; and Hϭheight of embankment. The stress reduction ratio lies between 0 and 1. When S3d is equal to 0, all the Variations of Pore-Water Pressure embankment load is transferred to piles. On the other hand, when S ϭ1, the pressure applied on soil surface is equal to the em- Fig. 10 shows the measured variations in the piezometric level 3d ͑ ͒ bankment load. Fig. 9 compares the measured and FE computed i.e., elevation head plus pore water pressure head with time in ͑ ͒ S as a function of the embankment height. With an increase in the soft silty clay at 4.0 m depth P1, z=21 m and 8.0 m depth 3d ͑ ͒ the height of the embankment, both the measured and computed P2, z=17 m beneath the embankment center. For calculating the value of S decreased gradually to 0.42 and 0.45, respectively. elevation head at each piezometer, the bottom of the sandy silt is 3d ͑ ͒ The measured and computed trends are in agreement with the taken as the reference datum i.e., at the z=0 plane . The com- findings by Han and Gabr ͑2002͒. puted results are included in the figure for comparisons and they In Fig. 9, predictions using two common methods developed are consistent with the general trends of the measurements at the by Russell and Pierpoint ͑1997͒ and Hewlett and Randolph two piezometers. As expected, both the measured and computed ͑1988͒ are also included for comparisons. It should be noted that results illustrate an increase in excess pore pressures during the the method of Russell and Pierpoint is based on Terzaghi’s soil period of construction due to an increase in the surcharge load of arching theory but extend it from the plane strain condition to 104 kPa. However, the increase of P1 was only about 11 kPa, three dimensions, whereas the method of Hewlett and Randolph whereas that of P2, the increase was only 14 kPa. The small in- ͑1988͒ considers soil arching as a series of domes of hemispheri- crease in pore water pressures was the result of load transfer from cal shape supported by piles. For low embankment heights ͑i.e., the foundation soft clay to adjacent piles due to soil arching and ͒ some dissipation of pore water pressures during construction. The less than 2.5 m in this study , S3d or soil arching is underpredicted by using the method proposed by Hewlett and Randolph ͑1988͒. piles carried the majority of the embankment load and so rela- This is because soil arching is captured by a series of hemispheri- tively the soft clay beneath the embankment was subjected to a cal domes between the pile heads in the method. When the fill much smaller compressive load. This was the main reason why thickness is less than the net spacing between the two near edges small excess pore water pressures were induced at P1 and P2 in of the piles ͑i.e., about 1.7 m in this case͒, there is no soil arching, the soft clay. By comparing the piezometric heads at these two locations, it i.e., S3d =1. On the other hand, predictions using the method proposed by Russell and Pierpoint ͑1997͒ are consistent with field is clear that there was an upward seepage developed as the piezo- measurements. In this method, soil arching is permitted by con- metric head at P2 was higher than that at P1 at a given time. ͑ sidering mobilized shear stresses arising from differential settle- About four months after the completion of the embankment i.e., ͒ ments between soil prism over the foundation soil and adjacent end of the monitoring period , about 70% degree of consolidation soil prisms over piles. was measured. On the other hand, the computed excess pore pres- For embankments higher than 2.5 m, predictions based on the sures dissipated more quickly. This may be because the reduction design methods developed by both Russell and Pierpoint ͑1997͒ of water permeability as a result of a decrease in void ratio during consolidation was not considered in the numerical simulation.

Fig. 9. Measured and computed stress reduction ration versus Fig. 10. Measured and computed variations of piezometric level at embankment height the center line of the embankment

Fig. 11. Computed excess pore pressure at the end of embankment construction

Usually it is not practical to construct an embankment at a slow rate that no excess pore pressures can develop in underlying soft clay. Rowe and Soderman ͑1985͒ warned that if the embank- ¯ ͑ ment is constructed at a rate such that Bmax which is the ratio of the maximum excess pore pressure to the change in total vertical stress͒ is substantially greater than 0.34, bearing capacity failure of an embankment may occur during construction. By comparing the measured pore pressure response with the estimated applied pressure of 44 kPa ͑which is taken as the average pressure of E1, ͒ ¯ 2, 3, 4, 6, 7 , the estimated Bmax is about 0.32, which is of the same order of magnitude as the recommended value. No failure was observed during the construction of the embankment. It can be seen from Fig. 10 that there was upward seepage in the foundation soil, because the piezometric level of P1 ͑installed at 4.0 m depth͒ was lower than that of P2 ͑installed at 8.0 m Fig. 12. Measured and computed surface settlements versus time ͑a͒ depth͒ during the whole measurement period. This suggests that at the pile head; ͑b͒ at the soil surface the excess pore pressures have an upward gradient. Fig. 11 shows the computed distributions of excess pore pressure together with 12. Although the rates of consolidation are slightly overpredicted the measured values indicated in the figure. As discussed earlier, by the 3D coupled analysis, the computed values agree reason- the piles carried a majority of embankment load and transferred it ably well with the field measurements. This gives us the confi- to greater depths. The soft silty clay was subjected to a smaller dence in the computed tensile forces in the geogrid to be compression load relatively. Due to the larger compressive discussed later. stresses and the longer drainage path, the maximum excess pore Due to the complex nature of the GRPS system, settlement pressures induced in the soils ͑i.e., medium silty clay and sandy analysis of GRPS embankments is usually carried out similarly as silt͒ near and at the pile base level. An upward gradient of excess that for an unreinforced case. In simplified methods, settlements pore pressure is clearly evident. of piles ͑or columns͒ and of the soil are calculated separately. From Fig. 8͑b͒, the average vertical stress acting on the pile was Settlements and Tensile Forces in Geogrid estimated to be 570 kPa. With the high compression modulus of the concrete annulus pile, the estimated pile compression was less Total and Differential Settlements than 0.5 mm. Thus, the measured pile settlement shown in Fig. Fig. 12 shows the measured settlements both on the pile heads 12͑a͒ was due to the compression of the sandy slit underneath the ͑measured by surface settlement markers S1 and S4͒ and on the pile base. soil surface between the piles ͑measured by S2 and S3͒. As ex- In this paper, differential settlement is defined as the difference pected, the maximum settlement occurs at the midpoint between in settlement at a pile head and at the soil surface between two the piles. At the end of the construction of the embankment, the piles ͑as illustrated in the insert of Fig. 12͒. Similar to the maximum measured settlements were 14 and 63 mm at the pile measured total settlements, the deduced differential settlement head and on soil surface, respectively. At the end of the monitor- also increased with an increase in the height of the embankment. ing period ͑i.e., 180 days since the commencement of the con- The measured maximum differential settlement was about 72 mm struction of the embankment or 125 days after the completion of and about 70% of it occurred during the construction of the the embankment͒, the measured maximum settlements increased embankment. to 19 and 87 mm due to soil consolidation, respectively. By using the hyperbolic method proposed by Tan et al. ͑1991͒, the final Distributions of Vertical Settlement Strain with Depth settlement is estimated to be 104 mm as indicated in Fig. 12. As Based on the measurements of the subsurface settlement ͑SS͒ the soil settlement between piles is considerably larger than the gauges ͑see Fig. 2͒, it is possible to calculate differential settle- pile settlement, downdrag of piles is inevitable and this may be ments between subsurface settlement gauges and hence the field considered in future designs. vertical strain distributions with depth can be obtained. Fig. 13 The computed settlements by the 3D coupled finite-element shows the deduced ͑or called measured͒ distribution of subsurface analysis are compared with field measurements and shown in Fig. strain with depth at the end of embankment construction. In Fig.

Fig. 13. Distribution of vertical soil strain with depth Fig. 14. Computed geogrid tensile force around Pile A at the end of embankment construction

13, the computed strain distribution from the 3D finite-element analysis is also included for comparisons. It can be seen that at ͑ ͒ the end of construction of the embankment, the maximum mea- soil settlement using Eq. 4 should be treated with caution if sured vertical strain reached about 1% at the upper silty clay, piles are present, especially they are closely spaced. Obviously, whereas the recorded vertical strain at the pile base level was the “hang-up” effects are automatically accounted for in the 3D 0.15%. It is very fortuitous that the measured and computed are FE simulations. consistent with each other. The two consistent results give us more confidence in the computed distribution by the 3D FE nu- Tensile Force in Geogrid merical analysis at the end of consolidation ͑i.e., marked as final͒ Fig. 14 shows the FE computed tensile force developed in the in Fig. 13. At the end of consolidation, the maximum computed geogrid around the Pile A at the end of the embankment construc- vertical strain occurs in the upper silty clay. The compression of tion. It should be noted that the diameter of each pile was 1 m and the soft silty clay counts for about 35% of the total settlement. On the pile spacing was 3 m center to center. As there was no mea- the other hand, there is a substantial increase in computed vertical surement of tensile force in the geogrid, no field data is available strains below the pile toe, where the maximum excess pore pres- for comparisons unfortunately. It can be seen that from Fig. 14 sure occurred at 17 m or deeper below the ground surface ͑see that the maximum computed tensile force developed in the geo- Fig. 11͒. As the field monitoring program was terminated 180 grid is about 16 kN/m, which is only 18% of the tensile strength ͑ ͒ days after the commencement of the construction of the embank- i.e., 90 kN/m of the geogrid. There is a sharp reduction of ten- ment ͑or 125 days after the completion of the embankment, there sion force in the geogrid from edge of pile toward the center. This is no measured strain distribution available for comparisons. is because of the sharpest change in the settlement occurred near ͑ ͒ On the other hand, it is useful and interesting to investigate the the edge of pile. Hence the maximum strain tensile force com- difference in computed vertical strain distribution by the 3D FE puted in the geogrid is also near the pile edge. numerical simulation and that from simple hand calculation of This small computed tensile force is in fact consistent with the vertical strain increment, ⌬␧ using the following one- relatively small measured differential settlement between a pile v ͑ ͒ dimensional equation: and the surface soil between piles see Fig. 12 . According to Han and Gabr ͑2002͒, geogrids have little influence on differential C settlement for 4 m high embankment supported on stiff concrete ⌬␧ = c ⌬͑log ␴Ј͒͑4͒ ഛ v 1+e v piles and differential settlement 100 mm. This implies that the 0 small tensile force in the geogrid is expected in this study as the ϭ ϭ where Cc compression index; e0 initial void ratio; and maximum measured differential settlement was only about ␴Јϭ v vertical effective stress. By using the measured pressure of 72 mm. 44 kPa acting on the soil surface due to the embankment, the hand-calculated vertical strains at various depths below the Lateral Soil Displacements and Lateral Displacement– ground surface is also shown in Fig. 13. As compared with the 3D Settlement Ratios FE numerical simulations, it is obvious that the vertical strains are grossly overestimated by using the simple one-dimensional equa- Fig. 15 shows the measured and computed lateral displacement tion, especially between the depths of 2 and 16 m where the piles profiles with depth at inclinometer I1 at 1.5 m from the toe of the were located. Based on this hand-calculated strain distribution, embankment ͑refer to Fig. 2͒. It can be seen that the measured the estimated surface settlement should be 350 mm, which is sub- maximum displacements at 2.8 and 5.6 m height of the embank- stantial larger than those measured in the field ͓see Fig. 12͑b͔͒. ment are about 4 and 10 mm measured at 4 m depth in the upper This significant overestimation of vertical strains in the soil be- silty clay, respectively. Although the shape of the lateral displace- tween the depths of 2 and 16 m is because the hand-calculations ment profiles is captured by the 3D FE simulation, the magnitude cannot take into account of the influence of piles on ground settle- of lateral displacements is overpredicted significantly. This may ments, i.e., the so-called “hang-up” effects on soft settling soil be attributed to the use of isotropic homogeneous soil models and inside a pile group provided by the surrounding piles ͑Ng et al. soil parameters in the FE analysis for simulating the anisotropic 2005, 2007͒. The hang-up effects provided by surrounding piles soils in the field. It has been widely studied and illustrated ͑Ng via shear stress lead to a significant reduction of vertical effective et al. 2004͒ that it is necessary to use an anisotropic model and its stress in the settling soil and hence reduce the magnitude of ver- corresponding anisotropic soil parameters to obtain consistent tical settlements and strains of the soil. Therefore, any calculation predictions in both soil settlements and lateral displacements.

are consistent with the measured values and the computed results from the three-dimensional finite-element analysis. For lower embankment, on the other hand, only the method proposed by Russell and Pierpoint ͑1997͒ gives consistent predictions with the field measurements. 4. As a result of the load transfer from the foundation soil to piles, the piles carried the majority of embankment load and transferred the load to greater depths. This transfer greatly reduced excess positive pore water pressures induced in the soft silty clay. The measured excessive pore water pressure ¯ ratio Bmax in the soft silty clay was only about 0.3. Fig. 15. Measured and computed lateral displacement profiles by the 5. The measured settlements both at the pile heads and at the inclinometer soil surface showed that the GRPS supporting system could reduce settlements greatly. The majority of the measured pile settlement was caused by the compression of the sandy slit According to Indraratna et al. ͑1992͒, the ratio of lateral beneath the pile base. Without considering the influence of displacement–settlement is a good indicator of embankment sta- piles ͑i.e., hang-up effects͒, the one-dimensional settlement bility. The ratio of lateral displacement–settlement is defined as analysis overpredicted vertical soil movements significantly. the ratio of the maximum lateral displacement at the toe to the At the end of the construction of the embankment, the maxi- maximum settlement at the center line of an embankment. By mum observed vertical strain reached about 1% in the upper studying the performance of a test embankment loaded to failure, silty clay, whereas the recorded vertical strain at the pile base Indraratna et al. ͑1992͒ concluded that a small value of the ratio is level was 0.15%. necessary to maintain the stability of the embankment over sub- 6. The measured relatively low differential settlement between soil. Chai et al. ͑2002͒ showed that when an embankment ap- the pile and the foundation soil indicates that the tensile force proached failure, the lateral displacement-settlement ratio generated in the geogrid was small. Based on the three- increased rapidly. At this embankment failure, the value was dimensional finite-element analysis, the maximum computed larger than 0.5. tensile force was about 16 kN/m, which is significantly In the current study ͑see Figs. 12 and 15͒, the measured lateral smaller than the tensile strength of the geogrid ͑i.e., displacement–settlement ratio in Shanghai was 0.17 at 50% em- 90 kN/m͒. bankment height ͑i.e., 2.8 m͒ and increased slightly to 0.21 at full 7. During the construction of the piled embankment, the mea- height ͑i.e., 5.6 m͒. The small increase in the lateral sured lateral displacement–settlement ratio was about 0.2. As displacement-settlement ratio during the embankment construc- compared with those cases where piles were not used, this tion indicates that the use of the GRPS system enhanced the sta- current measured value was relatively small. This suggests bility of the embankment significantly. that the use of GRPS system can decrease lateral displacements and enhance the stability of an embankment significantly. Conclusions

This paper presents a case history of a GRPS highway embank- Acknowledgments ment project at which a low improvement area ratio of 8.7% was used. Based on the results of field observations and three- The writers would like to acknowledge the Research Grants dimensional finite-element backanalysis, the following conclu- HKUST6025/01E and CA03/04.EG02 provided by the Research sions can be drawn: Grants Council of the Hong Kong Special Administrative Region 1. By comparing field loading tests on the foundation clay and and the support of the National Science Foundation of China ͑No. the 3 m by 3 m pile-draft at low area ratio, it is found that 50679017 and 50778063͒. the bearing capacity of the soft soil supported with piles can be improved by at least a factor of 3. 2. It is clear that there was a significant load transfer from soil to the piles as a result of soil aching. When the embankment Notation height increased to 5.6 m, i.e., the embankment load was about 104 kPa, the measured pressures acting on the soil The following symbols are used in this paper: ¯ ϭ surface increased by about 31–58 kPa, which is only 30– Bmax ratio of maximum excess pore pressure to change in 60% of the embankment load. On the other hand, the pres- total vertical stress; ϭ sure acting on the pile head increased to 674 kPa, which is Cc compression index; about 6.5 times larger than the embankment load. The mea- cЈ ϭ effective cohesion; ϭ sured contact pressure acting on the pile head was about 14 cu undrained shear strength; times higher than that acting on the soil located between the E ϭ Young’s modulus; ϭ piles. Both the measured and backanalyzed stress reduction e0 initial void ratio; ϭ ratios decreased gradually with an increase in embankment e1 void ratio at unit pressure; height. H ϭ height of embankment; ϭ 3. For embankment higher than 2.5 m, predictions of stress re- K0 coefficient of earth pressure at rest; ϭ duction ratio based on the design methods proposed by Rus- kw coefficient of permeability; sell and Pierpoint ͑1997͒ and Hewlett and Randolph ͑1988͒ M ϭ slope of critical state line;

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