CHAPTER 1 Failures Due to Long-Term Behaviour of Heavy

CHAPTER 1

Failures due to long-term behaviour of heavy structures

L. Binda, A. Anzani & A. Saisi Department of Structural Engineering, Politecnico di Milano, Milan, Italy.

1.1 Introduction

The authors’ interest towards the long-term behaviour of heavy masonry structures started after the collapse of the Civic Tower of Pavia in 1989, when L. Binda was involved in the Committee of experts supporting the Prosecutor in the trial, which involved the Municipality and the Cultural Heritage Superintendent after four people died under the debris of the tower. The response required by the Committee concerned the cause of the failure; therefore an extensive experimental investigation on site, in the laboratory and in the archives was carried out and the answer was given within the time of nine months. Several hypotheses were formulated and studied before ﬁ nalizing the most probable one, from the effect of a bomb to the settlement of the soil caused by a sudden rise of the water-table, to the effect of air pollution, to the trafﬁ c vibration and so on. Several documents were collected concerning the sudden collapse of other towers even before the San Marco tower failure and the results of the investigation were interesting. In fact, the failure of some towers apparently happened a few years after a relatively low intensity shock took place. In other cases, the collapse took place after the development of signs of damage, such as some crack patterns, for a long time. This suggests that some phenomena developing over time had probably to be involved in the causes of the failure, combined in a complex synergetic way with other factors. As for the experimental investigation carried out on some prisms cut out from the large blocks of the collapsed walls of the Pavia tower found on the site, the

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) doi:10.2495/978-1-84564-057-6/01 2 Learning from Failure attention was more and more concentrated on the dilatancy of the masonry under compressive monotonic and creep tests and on the fatigue behaviour of masonry under cycling loads. This chapter discusses the investigation carried out on the materials of the Civic Tower of Pavia and the conclusion reached by the previously mentioned Commit- tee. Furthermore, the phenomena of early and retarded deformations of historic masonry structures will be described together with the results of an investigation carried out on other damaged structures. Finally the research campaign carried out on site and in laboratory on the bell- tower of the Cathedral of Monza and the bell-tower of the Cathedral of Cremona. The investigation shows that the damaged state of the structures or of structural elements can be precociously detected by the recognition of the typical crack patterns, based on simple visual investigation. Collapses may be prevented by detecting the symptoms of structural decay, particularly the crack patterns, through on-site survey, monitoring the structure movements for long enough periods of time, choosing appropriate analytical models and appropriate techniques for repair and strengthening the structures at recog- nized risk of failure.

1.2 The collapse of the Civic Tower of Pavia: search for the cause

The Civic Tower of Pavia, an eleventh-century tower apparently made of brickwork masonry, suddenly collapsed on 17 March 1989 (Fig. 1.1). Several hypotheses were

Figure 1.1: The ruins after the collapse, seen from the arcade opposite to the Cathedral.

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) Failures due to Long-Term Behaviour of Heavy Structures 3 made about the causes of that sudden failure, from soil settlements to the presence of a bomb, from vibrations caused by trafﬁ c to the passage of super sonic jets. For a thorough understanding of the real causes of the collapse, an experimental investigation was carried out on site and in the laboratory, on the large amount of material coming from the remains of the tower.

1.2.1 Description and historic evolution of the tower

The tower, about 60 m high with a square base measuring 12.3 × 12.3 m was located close to the north-west corner of the Cathedral. Each of the four facades was divided horizontally into six orders (Fig. 1.2a and b). The fi rst four from the bottom were divided into fi ve parts by four pilaster strips topped by two small arches. The third and fourth orders had no pilaster strips, but were topped by similar hanging arches. The fi fth order terminated in a cornice. A large mullioned win- dow with two apertures opened out on each side of the sixteenth-century belfry. Inside the tower two timber fl oors were situated at a height of approximately 11 and 23 m.

(a) (b)

Figure 1.2: (a) The Civic Tower and Cathedral of Pavia, Italy. (b) Geometry of the Civic Tower and Cathedral of Pavia, Italy.

According to the few historical documents found, the ﬁ rst order and half of the second order can be dated between 1060 and 1100 AD [1, 2], the part from the middle of the second order and the ﬁ fth perhaps were built between the twelfth and thirteenth centuries; the tower was surmounted by a brick belfry and a timber roof. Between 1583 and 1598 the granite belfry weighing 3,000 tons, designed by the famous architect Pellegrino Tibaldi was set on top of the tower. A staircase built into the wall ran along all four walls from the south-west corner up to the belfry. The staircase was covered by a small barrel vault apparently made of conglomerate.

1.2.2 First experimental results and interpretation of the failure causes

The few documents available at the time of the collapse [3] were insufﬁ cient to give an accurate geometric conﬁ guration of the tower. Consequently, in order to draw prospects and sections of the tower the following operations, described in detail in [4, 5], were carried out:

• topographic survey of the remains of the tower (Fig. 1.3), and partial rectiﬁ cation of existing photographs to deﬁ ne the precise plan and the thickness and morphological features of the cross-section of the masonry; • reconstruction of the geometry of the belfry from a survey of the granite parts, practically all recovered from the internal portion of the remaining part of the tower; • assessment of the overall height of the tower from an existing aerial photogrammetric survey; • perspective plotting from existing photographs to reconstruct the geometry of the staircase and the arrangement of the architectural elements.

1.2.2.1 Structure and morphology of the walls The medieval walls, built according to the techniques normal at that time for towers, were characterized by two external brick cladding ranging from 120 to 400 mm

Figure 1.3: Photogrammetric survey of the remains of the tower.

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) Failures due to Long-Term Behaviour of Heavy Structures 5 with an average of 150 mm, with the intermediate portion of the walls consisting of irregular courses of large pebbles of brick and stones alternated with mortar, constituting a sort of conglomerate (Fig. 1.4). The walls of the second building phase were characterized by a much more irregular ﬁ lling and by thinner external facings. Figure 1.5 shows one of the large blocks among the remains revealing part of the section of the wall with the external cladding. Figure 1.6 shows a complete cross-section of the wall of the present remains of the tower (south side), the ratio between the thickness of the external leaf of the wall and the internal one was approximately 1:16. The section of the wall near the staircase was composed by an external wall similar to the one described above, but 1400 mm thick, a stairwell 800 mm wide, and an internal wall 600 mm thick. The latter wall was of the rubble type and was particularly heterogeneous.

Figure 1.4: Cross-section of the wall of the Civic Tower of Pavia.

Figure 1.6: View of the complete section of the bearing wall. Note Figure 1.5: Part of the section of the how thin the external facing is in bearing wall (2.8 m thick), showing the comparison with the total thickness external brick cladding. of the wall.

1.2.2.2 Geotechnical investigation The remains of the lower part of the tower left standing reached a height from 0.1 to 5 m still visible at the moment, since it was decided to leave the remains as they were without reconstructing the tower. The outside main wall is continuous and does not show any appreciable signs of dislocation or displacement from its original position. This, as well as the behaviour of the tower over time (there is no evidence of any specifi c surveys, but no appreciable settlement appears ever to have been reported), suggests that the collapse cannot be attributed to failure of the foundation soil. At most, possible differential settlements, caused by abnormal variations in the groundwater level in the previous years, could have worsened the stress–strain distribution within the structure, leading to the collapse. The settlements were, however, very limited and their effect is considered to be negligible. These qualitative consider- ations have been confi rmed by calculation [5]. The soil consists mainly of sandy deposits, sometimes silty or with lithoid elements, intercalated with highly impermeable strata of clayey, sandy silt. The most important clayey, silty strata are found between 7.5 and 10 m, 13.5 and 15 m and between 29.5 and 32 m below ground level. These are over consolidated materials with a medium to low degree of plasticity. On-site and laboratory geotechnical surveys were carried out to obtain the mechanical parameters of the soil [5]. The on-site survey consisted of two geognostic drillings in which undisturbed samples were taken and measurements made with a standard penetrometer (SPT); two seismic cone penetration tests (SCPT) and four cone penetration tests (CPT) were also carried out. The samples taken during the drillings were subjected to identifi cation, three-axial and oedometric compressibility tests. The penetrometric resistances give a similar picture of the pattern of the resistance of the soil. The values measured are as follows: from the base of the tower to a depth of approximately 14 m Nspt ranging from 8 to 30 blows/foot, Nscpt ranging 2 from 5 to 23 blows/foot, Qc ranging from 4.0 to 9.0 N/mm ; from 14 m down to the maximum depth reached, Nspt ranging from 34 to 65; Nscpt ranging from 26 to 2 56; Qc ranging from 14.5 to 28.0 N/mm . The shearing strengths were determined: (i) for sandy soils on the basis of the correlations presented in the literature between Nspt or Qc, effective vertical pres- ϕ sure sv and friction angle j ( was subsequently suitably reduced to take into account the presence of silt); (ii) for silty soils by means of laboratory tests (three- axial tests and direct shear tests). To calculate the bearing capacity, for safety’s sake, the soil from a depth of 4 to 14 m was considered. This depth range showed mean j values of 34° and 33°, respectively, depending upon the correlations adopted. To get at least an indicative value of the ultimate capacity, the foundation was initially considered in the two extreme situations of a continuous beam with width equal to 2.8 m (thickness of the foundation walls) and of a square foundation with a side equal to 12.3 m (base of the tower). By adopting the smallest shear resistance angle j = 33°, and prudently assuming the groundwater to be at the foundation level, an ultimate capacity of 2788 kN/m2

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) Failures due to Long-Term Behaviour of Heavy Structures 7 and the second case 4583 kN/m2 was calculated. The unit load on the soil was 1161 kN/m2 and the safety factor was therefore 2.4 and 3.95, respectively. The effective safety factor will lie between these and even the lower value can be considered sufﬁ cient to guarantee the stability of the foundation. In the period from January 1987 to February 1989, the maximum measured variation in level was 400 mm. Even though there are no reasons to believe that the variations around the tower were greater, the effect of an abnormal drop in level of 3 m was examined. The soil Δ was considered deformable down to the depth at which sv the variation in pore pressure stale is about 0.2 of the s¢v geostatic pressure. The average settlement calculated was 8 mm. Since the ground around the tower is relatively uniform, it must be assumed that the differential settlements are negligible. In order to evalu- ate the maximum theoretical distortion possible, penetrometric proﬁ les were calculated at opposite sides using all the minimum and maximum values recorded during the various tests at various depths. Maximum settlements of 11 mm and a minimum of 6 mm were obtained. The ultimate differential settlement would, therefore, be 5 mm and consequently of negligible effect on the stress–strain condition within the structure.

1.2.2.3 Physical, chemical and mechanical tests on the components To determine the effect of any possible chemical or physical degradation of the masonry, numerous samples of mortar were taken from the large blocks of masonry. The bricks and stones showed no signs of degradation except in the outermost area; in fact, even in the most deteriorated areas of the examined blocks, the degradation did not penetrate any deeper than 80–100 mm. Chemical and mineralogical/petrographic analyses were performed on 22 samples of mortars. The chemical analyses revealed that the binder used for the mortars during the fi rst building phase consisted chiefl y of lime putty (soluble silica 0.28–0.40%) and that the aggregate was mainly siliceous (unsoluble residue between 69.94 and 82.04%). The binder/aggregate ratio varied from 1:3 to 1:5. Similar values were obtained for the mortars of the second and third building phases. The porosity was around 12–13% and the bulk density about 18.5 kN/m3. In most cases, the sulphur trioxide content was negligible (around 0.06). Optical inspection of thin sections of the mortar revealed numerous porous areas which were sometimes covered by a layer of carbonates of relatively recent formation, thus making the surface of the mortar far more resistant. This could be the result of calcareous matter being deposited by fl ows of water. Similar deposits have been found in different areas of the masonry [6] and in each case the covering layer strengthened the surface of the mortar. Thin section mineralogical/petrographic analysis also confi rmed the total carbonation of the mortars and the siliceous nature of the aggregate and revealed corrosion along the surface of contact between certain aggregates (pebbles of stained quartz and fl intstone, etc.) and the binder. As it is quite common [7], however, the reaction products cause no fi ssures inside the mortars. The adhesion

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) 8 Learning from Failure between mortar, bricks and stone was also fairly good (except in cases where the building techniques had left large voids). The mortars were consistent, as the mechanical tests confi rmed, had a low content of sulphates and did not show heavy deterioration except for the outermost ones. The possibility of any signifi cant reduction in structural strength of the masonry due to the chemical or physical degradation of the mortars or other materials was, therefore, excluded. Since the collapse was not caused by the degradation of the building materials or sudden or differential settlement, attention was turned to how dead and live loads might have affected the mechanical behaviour of the materials over time. Compression tests were performed on small cubes of mortar [5] (with sides ranging from 2.7 to 3.5 mm) taken from the mortar joints of the inner conglomerate. The strength was 2.92–13.37 N/mm2, with a mean value of 6.45 N/mm2 and SD 49% (n = 11). Since the specimens are very small these results are merely indicative. Nevertheless it can be said that the results confi rm the chemical and physical analyses; in general, the mortar was consistent despite its heterogeneity and very hard and strong when sampled. The compressive strength of the bricks, on the other hand, as tested on cubes with sides of 40–50 mm was rather low: the mean value was 13.37 N/mm2, with SD 26% (over 50 specimens). The elastic modulus between 20 and 60% of the peak stress was 1973 N/mm2 for the bricks and 905 N/mm2 for the mortars [5]. Tests reported later show that the strength of the masonry was less than that of the mortar, suggesting that the low carrying capacity of the masonry was mainly due to the construction technique.

1.2.2.4 Compression tests on masonry prisms Compression tests were performed on prisms of masonry, cut from large blocks that had remained intact, in order to obtain the stress–strain curve up to and beyond failure [5]. Fatigue tests were then performed using a load value reproducing the stress induced by the dead load and applying a cyclic load, the amplitude of which repro- duced the stress variations due to the effects of the wind. Lastly, a survey of the effects of the dead load of the tower on the behaviour of the materials over time was carried out by means of constant load tests. Prisms measuring 4000 × 600 × 700 mm approximately were obtained from the recovered blocks. These dimensions were chosen so as to simulate the behaviour of the masonry, which was very thick (2.8 m) compared to height (60 m) and plan form (12.3 × 12.3 m) of the tower. The load-control or displacement-control compression tests were carried out with a 2250 kN, servo-controlled MTS hydraulic press, with programmed cycles. Monotonic compression tests to failure were conducted on seven masonry prisms from the ﬁ rst two building phases. The tests were carried out under displacement-control, at rates of 3.85 × 10–3 mm/s and 9.62 × 10–4 mm/s.

Figure 1.7: s–e curves of Figure 1.8: s–e curve obtained for a cyclic prisms subjected to monotonic compression test. compression tests.

Figure 1.7 shows the curves obtained for the seven prisms, two of which (102A and 102B) were tested by applying the load in the direction of the horizontal joints. The peak strengths and ultimate strain values vary quite considerably: low ultimate strains appear to correspond to higher strengths. Strength varies from 2.0 to 4.1 N/mm2, ultimate strains from 3.0 to 5.5 × 10–3 and the modulus of elasticity, deﬁ ned between 20 and 40% of the peak stress, varies from 719 to 1802 N/mm2. Five prisms were tested to failure by means of loading and unloading cycles applied every 0.5 N/mm2 under displacement-control conditions up to and beyond the peak stress. Strength varied from 1.8 to 3.3 N/mm2, the elastic modulus from 544 to 1455 N/mm2 and the ultimate strains from 3.6 to 8.5 × 10–3. A typical curve is shown in Fig. 1.8. Although on average the strength is lower than that of the seven prisms subjected to the monotonic tests, the cycles appear not to have any great inﬂ uence on the s–e curve, the peaks of which at each loading approximate well to points of the monotonic curve.

1.2.3 Long-term tests

The behaviour detected from cycling tests and particularly the evident increase in deformation while the stress was kept constant (Fig. 1.8) led to a study of the effects of fatigue and long-term tests at constant load. The experimental research is described in the following sections.

1.2.3.1 Fatigue tests It is well known that repeated load cycles cause damage to the material. The damage originates from imperfections in the material itself, such as small cracks which

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) 10 Learning from Failure get larger as the cycling load is applied. Generally, failure occurs at peak load value lower than that measured when the load is statically applied. In the case of a masonry structure, fatigue may be caused by the repeated action of horizontal loads such as wind or seismic loads. In the particular case of the tower, no appreciable seismic effects have been recorded, whereas the effects of the wind must certainly have been felt over the centuries, causing signiﬁ cant variations in the stress due to the static load of the tower. As mentioned above, the damage caused by repeated cycles of loading and unloading is not very high when the average load applied is low and cycles are not frequent. However, signiﬁ cant damage may be caused if the cycles are repeated at an average stress close to the ultimate capacity [8]. Three prisms were subjected to cyclic loads corresponding to calculated stress variations of 0.2 N/mm2 starting from very high compression values similar to those produced by the dead load as calculated at the most loaded points of the structure. Cyclic loads corresponding to repeated wind effects, simulated according to the Italian Code, produced no appreciable damage except for higher strains (Fig. 1.9), when the loads starting from stress values very close to failure were applied. The fact that the load history included one or more cycling phases did not reduce the failure values (which remained 1.7, 2.7 and 4.4 N/mm2) obtained during the monotonic tests.

Figure 1.9: s–e curve obtained during Figure 1.10: s–e and e–t curves the fatigue test performed to simulate obtained during a step-by-step wind effects. constant load test.

1.2.3.2 Constant load tests Displacements measured on the prisms during loading showed a tendency to increase at constant load suggesting that the behaviour of the material could be time-dependent. During the constant load tests, almost all the prisms were tested under load control up to 1.0–1.5 N/mm2. The stress was then increased in steps of 0.14 N/mm2, at intervals of at least 15 min. The increase in the effects of strain was on average 1.6 × 10–3 for each 15-min interval at constant load. At higher stresses close to the ultimate strength of the material, the time-dependent effects of the constant load evolved more rapidly. No further load increases were made until the increase in strain stopped. At the last step the strain rate con- tinued to increase rapidly until failure occurred suddenly after a period of time varying between 10 min and 2 or 3 h. It seems reasonable to assume that the time needed to reach collapse is a function of the ratio between the load applied and the maximum load the specimen is able to withstand. The curves of strain as a function of time, (Fig. 1.10) clearly show the type of behaviour described earlier. At loads over approximately 70% of the ultimate compression strength, a small number of cracks appeared on one of the sides of the tested specimens. The cracks were found chiefl y on the bricks and at the surfaces of contact between the mortar and the stone or bricks of the inner face of the wall. The cracks, which were always vertical, were hardly visible right up to the moment of collapse (see Fig. 1.11a and b). Although the tests were carried out under load control, it was almost always possible to keep the collapse under check and thus prevent the sudden spalling, which often characterizes the failure of prisms of solid bricks arranged in regular courses, and/which is often an indication of a brittle failure of the bricks. Structural analysis carried out by a FE (fi nite element) elastic model [5] revealed that some parts of the tower were subjected to severe stress, very often close to the failure limits found experimentally. This probably led to the gradual evolution of micro-fi ssures which, over the centuries, may have contributed to the sudden collapse of the material for no apparent immediate cause and without the appearance

(a) (b)

Figure 1.11: Width of cracks after (a) 22 min and (b) 60 min under a constant stress of 2.0 N/mm2 (peak stress).

Figure 1.12: Vertical cracks on the external wall of the tower (1968). of warning signs, such as large cracks or spalling, even during the days immedi- ately preceding the collapse. A careful examination of the photographs taken by archaeologists for the Civic Museums in 1968 reveals that even at this time there were thin vertical cracks largely diffused on the outer face of the wall. These cracks were extremely difﬁ cult to see from the Piazza and were similar to those that appeared on the specimens during the tests (see Fig. 1.12).

1.3 Long-term behaviour of masonry structures

Masonry is a composite material and its mechanical and physical behaviour strongly depends on that of its components (mortar and brick/stone). Mortar inﬂ uences mainly the deformability and bricks or stones inﬂ uence mainly the strength. Historic buildings are very often characterized by high values of deformations, which may have taken place in the past or may still be in progress and may lead the building even to unexpected failure. Early deformation due to delayed hardening of hydrated mortars based on carbonation is typical of ancient buildings pre- senting thick mortar joints; multiple leaf masonry is usually characterized by differential creep displacements induced by the different deformability of the leaves; persistent and cyclic loads can give rise to a creep–fatigue interaction and to greatly retarded strain.

1.3.1 Deformation during mortar hardening

It has been shown that early creep behaviour, due to the carbonation process of fresh mortar, can last for a long time particularly when mortar was made with

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) Failures due to Long-Term Behaviour of Heavy Structures 13 hydrated lime [9]. This can be the case, for instance, of very thick ancient walls or of masonry characterized by very thick mortar joints like those of St. Vitale in Ravenna shown in Fig. 1.13 [10]. Increasing deformation due to heavy dead or cyclic loads can vary the geometry of masonry walls in a visible way already during the construction. These modifi cations can occur locally, or involve a whole structural element. Large displacements and deformations frequently involve piers and columns like those of gothic cathedrals (Fig. 1.14) [11] due to the horizontal thrust exerted by vaults and arches or due to soil and structure settlements. Generally speaking, old or ancient structures are continuously subjected to modifi cations concerning their geometry and their state of stress and strain. J.L. Taupin [12] says that ‘time moulds the structure of towers, cathedrals, bridges etc. which we would like to consider immutable.’ Time plays a role both in the short and in the long run dispersing and returning energy in three ways: through deformations and settlements, through vibration, and through material modifi cation or deterioration. Figure 1.15 shows a detail of the well-known Hagia Sophia at Istanbul, where the rotation of a column and the deformation of an arch in the north gallery can be clearly observed [13]. Figure 1.16 shows a much less famous small Romanesque church, St. Maria la Rossa in Milan, dating from the tenth to thirteenth century. This single aisled brickwork masonry building is covered by a timber gable roof, with a chancel comprised by two small chapels besides the choir, terminating with a semicircular apse. The church was subjected to different transformations during centuries, and its present aspect is due to the restoration works done in the 1960s. In the picture the tilt of the lateral walls and the deformation of the central arch can be seen [14].

Figure 1.13: View of the Basilica of San Vitale in Ravenna (Sixth century AD).

Figure 1.14: Pillar of the Cathedral of Salisbury [11].

Figure 1.16: Church of St. Maria la Figure 1.15: Hagia Sophia, north Rossa, view of the central nave gallery looking west. looking east.

1.3.2 First, secondary and tertiary creep in rock and hardened masonry

The inﬂ uence of time on the mechanical behaviour of stiff clays, soft rocks, fresh cement mortar, concrete and hardened concrete becomes evident when both uni- axial–triaxial compressive test at different rate of loading and compressive test at vertical constant load are carried out. On the one side, when testing compression of soft porous materials a decrease of the rate of loading produces a decrease of the vertical peak stress and of the stiffness of the material (Fig. 1.17 v. Cook). On the other side, if a constant load is applied an increase of deformation develops which is commonly subdivided into three phases: the so-called primary, secondary and tertiary creep (Fig. 1.18 v. Cook) [15]. The appearance of one or more of these phases and the strain rate of the secondary creep phase depend on the stress level. Very poor research was done before the collapse of the Civic Tower in Pavia, on the creep behaviour of masonry structures, apart from the papers published by Lenczner [16].

Figure 1.17: Dependency of strain on the loading rate.

Figure 1.18: First, secondary and tertiary creep.

The infl uence of time on the mechanical behaviour of masonry structures under high states of stress became evident after the collapse of the Tower of Pavia, when the identifi cation of a time-dependent behaviour, probably coupled in a synergetic way to cyclic loads [17], was identifi ed as a possible explanation of the sudden collapse.

1.4 Collapse and damage of towers due to long-term heavy loads

The failure of monumental buildings is fortunately an exceptional event; nevertheless, when their safety assessment is required, any risk factor that may affect the integrity of the buildings has to be taken into account. Ancient buildings often show diffused crack patterns, which may be due to different causes in relation to their original function, to their construction technique and to their load history. In many cases it is simply the dead load, usually very high in massive monumental buildings, which plays a major role into the formation and propagation of the crack pattern. The only way to prevent the occurrence of these failures is continuous observa- tion and maintenance of these structures.

1.4.1 St. Marco bell-tower and St. Maria Magdalena tower in Goch

The fi rst well-studied example of the collapse of a tower in Italy was certainly the one of the bell-tower of St. Marco in Venice in 1902 (Fig. 1.19) [18]. The tower collapsed suddenly with no previous evident signs of heavy damage. In the long debate following the tower collapse and in the long report made by Luca Beltrami in [18], the settlement of foundation was excluded from the causes and it was clearly described that the damage was interesting considering the structure and the repairs made 45 years earlier by confi ning the bearing corners of the tower with steel reinforcements (Fig. 1.20). A similar situation occurred in June 1902 with the collapse of a bell-tower in Corbetta, near Milan, in the same year. The tower was under modifi cation, being elevated from the original 24 m to 41 m in 1860 and by adding a spire in 1900. In 1993, the bell-tower of Sancta Magdalena church in Goch collapsed suddenly during the night (Fig. 1.21); it had already been decided a few years earlier to start a repair intervention due to the extent of the damage detected. Probably the waiting time was too long, taking into account that the tower was badly cracked for a long time and also damaged during the last world war.

1.4.2 The bell-tower of Monza Cathedral and the Torrazzo of Cremona

The bell-tower of the Cathedral of Monza is a sixteenth-century building made of solid brick masonry, at present subjected to a repair intervention. Its walls were showing large vertical cracks crossing the whole transversal section of the walls on the west and east (Fig. 1.22a and b), which were continuously widening at a constant rate [19, 20]. These cracks were certainly present before 1927, when they were roughly monitored. Wide cracks were also present in the corners of the tower

Figure 1.19: The bell-tower Figure 1.20: Detail of the collapse with the of the St. Marco Basilica in reinforced pillar [18]. Venice. at a height of 30 m, together with a damaged zone at a height of 11–25 m with a multitude of very thin and diffused vertical cracks. A similar crack pattern is visible on the Torrazzo, a medieval brickwork tower adja- cent to the Cathedral of Cremona (Fig. 1.23a and b) [21]. The precise date of construction is not known but is assumed to be around the thirteenth century. It belongs to a group of monuments, including the Cathedral, the Baptistery, the Town Hall Palace, the Militia Loggia, which forms one of the most beautiful Italian squares. The external load-bearing walls of the tower, which is about 112 m tall, have been showing several cracks for many years [21]; since the crack pattern has expe- rienced an evolution, a time-dependent behaviour of the material may possibly be assumed to cause the phenomenon.

1.5 The role of investigation on the interpretation of the damage causes

On the basis of the previous experience the authors have developed investigation procedures for the safety of these structures; the idea came ﬁ rst when studying the collapse of the Civic Tower in Pavia.

Figure 1.21: The church of St. Magdalena in Goch (Germany) after the collapse of the bell-tower, 1993.

The procedure is based on the following steps: (i) historic research to know the evolution of the structure over time, (ii) geometrical and crack pattern surveys, which allow one to understand the evolution of the structure, to calculate weights and give a fi rst interpretation of the crack pattern, (iii) geognostic investigation and monitoring, to understand the soil–structure interaction, (iv) on-site mechanical and non-destructive testing (radar, sonic, etc.) to defi ne local states of stress and stress– strain behaviour of the material, (v) chemical, physical and mechanical tests on mortars, brick and stones to fi nd their composition and their characteristics, (vi) if necessary, passive and active dynamic tests on site to survey the overall structural behaviour and (vii) monitoring system applied to the structure when necessary.

1.5.1 The bell-tower of the Cathedral of Monza

The bell-tower of the Cathedral of Monza, is a masonry structure 70 m high, with a square plan (a side is 9.7 m long) with solid brick walls 140 cm thick. The tower construction started in 1592, probably following the design of Pellegrino Tibaldi, the architect of the Pavia tower belfry, and ended in 1605 [19, 22]. The only damage to the tower reported by the documents occurred in 1740 and was due to a ﬁ re which started in the bell-tower and caused the collapse of the belfry dome and roof and the fall of the bells with their supporting frame down to the vault of the

Figure 1.22: Survey of the crack Figure 1.23: Survey of the crack pattern pattern for the bell tower of the for the Torrazzo tower: (a) west and Cathedral of Monza: (a) west and (b) east sides. (b) east sides.

ﬁ rst ﬂ oor at 11 m. No damages were reported in other known calamities, such as lightning or thunderstorms through the centuries. Nevertheless cracks are present since 1927 or even before, as mentioned above. From 1978 the cracks have been surveyed with removable extensometers: they show a slow increase of their opening through time. From 1988 the rate of opening seems to be increasing faster. The trend of widening of the three main cracks was calculated as 30.6, 31.3 and 39.7 μm/year from 1978 to 1995. Actually if this trend is considered from 1988 to 1997 the values change, respectively, into 41.2, 35.2 and 56.2.

The fi rst step of the investigation procedure [19] was the geometrical survey [20]. A geodetic network set up in the square of the Cathedral in 1993, was used as support. No relevant leaning was measured due to the small subsidence which is taking place in the square. Two distinct products were obtained: (i) a detailed three-dimensional model from which the external and internal prospects and the vertical sections were obtained and (ii) a simplifi ed model for which only the essential aspects of the geometry were preserved for the structural analysis. The survey of the crack pattern showed that the tower walls have a dangerous distribution of passing-through cracks on the western and eastern load-bearing walls for more than 50 years, and of a net of thin vertical cracks from a level of 11 m up to 30 m (Fig. 1.22). Other cracks can be seen on the internal walls of the tower; they are very thin, vertical and diffused along the four sides of the tower and deeper at the sides of the entrance where the stresses are more concentrated. The thin diffused cracks run 450 mm deep inside the section, reducing its total working thickness from 1400 mm to no more than 900 mm. From laboratory tests it was found that the mortar is very weak and made with putty lime and siliceous aggregates; also the bricks were of poor strength (between 4 and 12 N/mm2 measured on 40 mm-side cubes). On-site single fl at-jack tests were carried out at different heights of the tower (5.4, 5.6, 13.0, 14.0, 31.5 and 38.0 m) and the stress values against the height are plotted in Fig. 1.24. The maximum compressive stress acting in the tower, measured on site by the fl at-jack test, is about 2.2 N/mm2. The most interesting infor- mation came from the double fl at-jack test results, where it was possible to see the real risky situation if compared with the local state of stress measured by the single fl at-jack (Fig. 1.25). Passive dynamic tests using the bell ringing were also carried out monitoring the dynamic excitation of the extensometers applied across the

Figure 1.24: Single ﬂ at-jack tests of Monza.

(a)3.00 (b) 3.00

2.50 2.50 ] 2.00 2 2.00 ] 2

1.50 1.50 of stress of stress Local state 1.00 Local state 1.00 Stress [N/mm Stress [N/mm 0.50 0.50 ε ε ε ε h v h v 0.00 0.00 -2.00 -1.00 0.00 1.00 2.00 3.00 4.00 5.00 -2.00 -1.00 0.00 1.00 2.00 3.00 4.00 5.00 Strain [μm/mm] Strain [μm/mm] Figure 1.25: Monza tower stress–strain plot at (a) 5 m and (b) 13 m height.

main cracks, giving under these cycling stresses a maximum peak to peak (opening to closing) of 28 μm that has to be compared with a daily variation of 100 μm due to the temperature effects. The diagnosis based on the experimental survey and on the FE modelling lead to the conclusion that the bell-tower was a high risk building and needed a quick intervention. In Chapter 8 the preservation and repair intervention which is still being carried out is illustrated.

1.5.2 The ‘Torrazzo’ of Cremona

The bell-tower of the Cathedral of Cremona, an interesting historic town not far from Milan (Italy), is known by the nickname ‘il Torrazzo’ from long time ago. The tower is situated at the northern side of the Cathedral and it is connected to it by a Loggia called ‘Bertazzola’. The geometry of the tower is rather complex, being composed of a lower part (Romanesque tower) with a square plan of 13 m side and 70 m high, an upper part, the Ghirlandina, with an octagonal plan (2.5 m side), more than 40 m high. The Torrazzo is known as the tallest medieval bell- tower in Europe being 112 m high [23]. The lower part of the tower, with a square plan, is a massive construction with few openings localized on the western and eastern sides. The upper Ghir- landina appears as a light structure with arches and large openings on all the four sides. The staircase from the lowest level up to the Ghirlandina level was built within the thickness of the walls (approximately 3.3 m thick). Along the staircase, covered with a barrel vault, the thickness of the external wall is approximately 1 m, while the thickness of the internal wall is 0.7–1 m with the span of the staircase measuring 1.3–1.6 m. The staircase allows one to reach some internal vaulted rooms.

Archive research did not clarify the date of construction; nevertheless the highest number of reference data collected locates the date of construction between the eighth and the thirteenth centuries. In 1491, the porch of the Bertazzola was added connecting the Torrazzo with the Cathedral and in 1519 the Loggia was built resting on the arches of the porch. Maintenance works were carried out starting from the fi fteenth century. These works mainly concerned the highest part of the tower damaged by storms and lightening, especially the stone and brick columns which were sometimes substituted. The last intervention at the Ghirlandina was carried out in 1977. The works performed were the following: connection of structural and decorative elements, construction of a concrete frame sustaining the twin columns of the ‘Stanza delle Ore’ (at 85 m height) and surface treatments of stone and brick elements with an epoxy resin. The fi rst step of the investigation carried out in 1998 was the geometrical survey. A principal network defi ning fi xed points in the horizontal and vertical plan was set up having 21 nodes inside and around the tower made with fi xed nails. The co-ordinates of the nodes were determined with a T2000 WILD equipment. The vertical and horizontal profi les were determined by rays starting from the network nodes, using a TC1600 DIOR system and an auto scanning Laser System MDL. A photogrammetric survey of the external prospects was also carried out using TC1600-DIOR and T460* DISTO equipment. The prospects were obtained by a Rollei special software, MSR. The survey enabled the fi nding of some irregulari- ties of the structure: (i) a 21 cm horizontal displacement of the centre of the tower in direction north-east, calculated from the ground level to the top at 112 m, (ii) a non-symmetrical reduction of the plan dimensions from the ground level to the top at 31 cm for the north-east corner and 66 cm for the south-west corner, (iii) the Ghirlandina not being perfectly centred on the square part of the tower, but with a slight counter-clockwise rotation toward west. The presence of a diffused crack pattern particularly on the western and the eastern sides of the tower and on the Ghirlandina can indicate high states of stress due to the dead loads, the temperature variations and/or to a slight leaning. The survey was carried out on the outer surfaces by reaching the height of 60 m thanks to a special crane. The crack pattern is certainly also infl uenced by differential movements due to temperature variation between one side and the other of the tower. The highest variations certainly occur between the north and the south side. The west side has a diffused fi ssuration with passing-through cracks; the cracks are mostly vertical and start from approximately 20 m. Important cracks appear also between 48 and 60 m from the ground level (Fig. 1.23a). The north side is cracked in the centre between 27 and 40 m and at the north-east corner. The east side is cracked between 6 m and 20 m from the ground level and between 35 and 60 m (Fig. 1.23b). The south side has few cracks located between 14 m and 27 m. The Ghirlandina shows the most important cracks, on the buttress and on the brick columns particularly on the south-west corner. Also the internal part of the tower, along the staircase and inside the rooms shows a diffused crack pattern with some passing-through cracks. Three thresholds were established concerning the mea sure of the opening of the cracks: <3 mm, between 3 and 10 mm, >10 mm. The crack

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) Failures due to Long-Term Behaviour of Heavy Structures 23 pattern survey helped one to understand and interpret roughly the mechanical damage and to locate the position for the monitoring system. The inspection of the masonry surface and the inside of the walls leads to the following description: (i) the walls are made with solid bricks and no rubble was used for the inner part of the section; (ii) the bricks are regular with varying dimensions: 240–280 × 100–1200 × 55–70 mm; (iii) the external walls of the Ghir- landina are irregularly scaled and tooled; (iv) the colour of the bricks is variable red, dark red, yellow, orange etc.; (iv) the mortar joints are regular with thickness variable from 10 to 30 mm; (v) different techniques of jointing and pointing can be found and often the vertical joints seem to be void or recessed; (vi) the masonry texture is also regular with header and stretcher alternatively positioned; (vii) an external leaf one brick thick with a weak collar joint is certainly present along the staircases, in the internal rooms and at the level of the Bertazzola and more research is needed to test the real extension of this leaf along the tower; (viii) in the staircase walls a row composed of 4–22 headers is repeated at rather regular intervals as if it should represent a connection of the external leaf to the internal one; (ix) several scaffolding holes externally closed can be seen along the masonry walls; and (x) numerous restorations by brick substitution can also be seen externally and internally. Together with the geometric survey an accurate survey of the material decay was carried out. Concerning the reinforced concrete frame (85 m level) the columns between the north and north-east and the east and south-east side show washout of the binder, formation of carbonates near the stirrups with partial detachment of the reinforcement cover (no more than 1 cm thick) and reinforcement corrosion. Sixteen samples of bricks and mortars were collected from the masonry: fi ve from the facades, four inside from the walls of the internal rooms, four along the staircase and three from the external and internal walls of the Ghirlandina. The maximum depth of sampling was 300 mm. All sampling operations were docu- mented graphically or photographically. Laboratory tests were carried out on mortar and bricks. Chemical analyses showed that the mortar binder was hydrated lime (probably lime putty) and the aggregates were mainly siliceous. Two types of bricks were used, which differ in colour (red and brown) and also in properties. The red bricks have high absorption (21–28.8%), low strength 8–12.4 N/mm2 in compression and in tension (0.1–1.6 N/mm2) and low modulus of elasticity (1000–2175 N/mm2); the brown bricks have lower absorption (18.5–21.7%), higher compressive strength (9.4–25.43 N/ mm2) and tensile strength (2.2–2.6 N/mm2) and modulus of elasticity (1725–4417 N/mm2). The two types of bricks are present everywhere in the tower, so an average between the two bricks can be considered as reference. In order to detect the suspected existence of an external cladding in use during the Middle Age as a false curtain to hide the roughness of the real wall, bricks were sampled from the external wall of the Bertazzola at 6 m level and from the walls of the ‘Stanza dei Contrappesi’ at 13.6 m level. This external leaf was confi rmed and its thickness is around 12 cm. The sampling allowed one to fi nd large areas

WIT Transactions on State of the Art in Science and Engineering, Vol 11, © 2007 WIT Press www.witpress.com, ISSN 1755-8336 (on-line) 24 Learning from Failure where the leaf seems to be detached from the rest of the wall; following these results the application of NDT technique was required in order to map the detached areas which represent structurally a reduction of the wall section to be taken into account when modelling. All the areas from where samples were taken were then repaired with similar bricks and mortars. Flat-jack tests: Single and double fl at-jack tests were carried out on the Torrazzo. The single fl at-jack test was also used to study the behaviour of the external leaf of the wall. Different types and dimensions of fl at-jacks were used: (i) 240 mm × 12 mm rectangular jacks where the detachment of the external leaf was suspected, (ii) 400 mm × 200 mm rectangular jacks and (iii) 350 mm × 240 mm semicircular jack where no detachment was suspected and for the double jack-test. Twenty-one tests were carried out, of which 19 were with single fl at-jack and 2 with double fl at-jack: 3 single fl at-jack tests at between 1 and 5 m from the ground, 7 single fl at-jack tests at 7 m, 10 single fl at-jack between 15 and 18 m and 1 single fl at-jack at 22 m. The double fl at-jack tests were carried out at 7.2 and 19 m from the ground. The results of the single jack tests are reported in Fig. 1.26 and show clearly two situations: a state of stress varying between 0.4 and 0.9 N/mm2 where the test found a detached leaf and a state of stress varying from 1.01 and 1.81 N/mm2 where no detachment was found. Also double fl at-jack tests were performed and Fig. 1.27 shows the stress–strain plots. It was impossible to carry out tests at higher levels due to the lack of scaffolding and of appropriate means for carrying the jack equipment. In future other tests will be carried out.

3.50 masonry section 3.3 m

masonry section 1 m 3.00 15.2-16.5 m presence of veneer

2.50 inner walls (rooms)

] 16-17.8 m 2

16.6-17.8 m 2.00 7.2-7.7 m 7.2-7.4 m 15.4-19.1 m 7-7.7 m 1.50 5 m Stress [N/mm 1.7 m

1.00

0.50

0.00 Figure 1.26: Single ﬂ at-jack tests of Cremona.

(a)4.00 (b) 4.00

3.50 3.50

3.00 3.00 ] ] 2 2.50 2 2.50

2.00 2.00 Stress [N/mm 1.50 Stress [N/mm 1.50 of stress of stress Local state Local state 1.00 1.00

0.50 0.50 ε ε h v ε ε 0.00 0.00 h v -3.00 -2.00 -1.000.00 1.00 2.00 3.00 -3.00 -2.00 -1.000.00 1.00 2.00 3.00 Strain [μm/mm] Strain [μm/mm]

Figure 1.27: Stress–strain plot at (a) 7.2 m and (b) 19 m height.

1.6 Comparison between the two towers

Since the bell-tower of Monza is considered a building with high risks of collapse, a comparison between the data collected on both towers seems to be useful to understand better the real situation of the Torrazzo. As mentioned above, the mortar composition of the two bell-towers does not differ much from one another, though the Torrazzo mortar seems to be more consistent. The bricks of the Monza tower are generally weaker than those of the Torrazzo (Fig. 1.28a and b) except for the brown type, which is mainly used on the outside surface of the bearing walls and very seldom used in the interior. On the contrary the brown and the red bricks are evenly distributed in the Torrazzo walls. It is also interesting to compare the results of single and double fl at-jack tests carried out on the two towers. The results of four tests, two for each tower are discussed. In Fig. 1.27a and b maximum stress reached with the double fl at-jack test on the Torrazzo together with the values obtained with the single one, respectively, at 7 and 19 m height are considered, showing an elastic linear behaviour up to, respectively, 2.45 and 2.7 N/ mm2. The maximum stress level when cracks clearly appear is, respectively, 3.77 and 3.77 N/mm2 and the state of stress measured is 1.5 and 1.5 N/mm2. So in these two cases the safety coeffi cient at collapse is certainly more than 3. In Fig. 1.25a and b the results of two tests at the height of 5 and 13 m, out of the four carried out on the walls of the Monza tower, are considered. Here the linear elastic behaviour stops at, respectively, 1.65 and 1.1 N/mm2 and the maximum stress reached before cracks propagated was 2.62 and 1.87 N/mm2. The measured local state of stress was, respectively, 1.67 and 0.98 N/mm2. In these two cases the safety

(a) 34 (b) σ 26 σ 32 red brick 24 brown brick 30 22 28 20 26 18 ]

24 2 16

] 22

2 14 20 12 18 10 16 14 Stress [N/mm 8

Stress [N/mm 12 6 10 4 8 red brick 2 ε ε brown brick h v 6 0 4 -20 -16 -12 -8 -4 0 4 8 12 16 20 2 ε ε Strain [μm/mm] h v 0 -20 -16 -12 -8 -4 0 4 8 12 16 20 Strain [μm/mm] Figure 1.28: Stress–strain plot for (a) Monza tower bricks and (b) Torrazzo bricks. coefﬁ cient at failure is much lower than in the ﬁ rst one and certainly less than 2. Furthermore in the case of the Torrazzo the modulus of elasticity is much higher and the Poisson ratio much lower than in the case of the Monza tower.

1.7 Conclusions

The investigation carried out on the specimens cut from the walls of the Pavia tower after its collapse allowed formulating for the ﬁ rst time on an ancient masonry the hypothesis of a collapse due to the long-term behaviour of the material. Probably since the construction of the bell-tower in the sixteenth century the structure was under a high state of stress and the damage very slowly but continuously increasing until the collapse. The creep behaviour of the material was shown clearly during the experimental research which started in 1989 and is still developing, as will be shown in Chapter 2. Examples of other similar situations were found in the history of collapses of towers and damages or collapses of churches (Noto cathedral). The two experiences of investigation on tall towers allow some concluding remarks:

• the on-site and laboratory tests carried out following the methodology described in the ﬁ rst section allowed one to detect situations of danger and to characterize the materials and calculate input parameters for the structural analysis; • the laboratory tests were able to show the difference of properties of the masonry in the two buildings and that where the materials used are weaker, the damage is more;

• the ﬂ at-jack test is a powerful tool to calculate the actual state of stress in compression and to detect the mechanical behaviour of the masonry, so that two different situations (Torrazzo and Monza towers) can be compared; • the investigation allowed the authors to state that the situation of the Monza tower is very difﬁ cult and that a quick intervention has to be started; • for the Torrazzo, even if the state of damage is not considered dangerous, a monitoring system has been set up, and the tower will be under control for 4–5 years at least in order to study its further evolution; in the meantime some repairs are being done for the external surfaces.

FE numerical models were used for the static and dynamic analysis of the two towers [19, 23]. The results of the experimental research were used to calibrate the FE models.

References

[1] Panazza, G., Campanili Romanici a Pavia, Arte Lombardia, pp. 18–27, 1956. [2] Ward-Perkins, B., Scavi nella Torre Civica di Pavia, Sibrium, 12, pp. 177– 185, 1973–75. [3] Milano, F. & Toscani, X., Il fond di documenti relativi alla Torre Civica esist- ence nell’Archivio Comunale di Pavia, Sibrium, 12, pp. 467–493, 1973–75. [4] Anti, L. & Valsasnini, L., Indagini preliminari all’analisi strutturale ed alle prove sui materiali della Torre Civica di Pavia, TEMA J., L’Arsenale, Venezia, 1991. [5] Binda, L., Gatti, G., Mangano, G., Poggi, C. & Sacchi Landriani, G., The collapse of the Civic Tower of Pavia: a survey of the materials and structure. Masonry International, 6(1), pp. 11–20, 1992. [6] Knoffel, D.F.E. & Wisser, S.G., Microscopic investigation of some historic mortars, Proc. 10th Int. Conf. Cement Microscopic, S. Antonio, Texas, 1988. [7] Baronio, G. & Binda, L., Reazioni di aggregati in intonaci antichi, Conv. Intonaco: Storia, Cultura e Tecnologia, Bressanone, pp. 269–276, 1985. [8] Binda, L., Anzani, A. & Mirabella Roberti, G., The failure of ancient towers: problems for their safety assessment, Int. Conf. on Composite Construction - Conventional and Innovative, IABSE, Insbruck, pp. 699– 704, 1997. [9] Ferretti, D. & Bazant, Z.P., Stability of ancient masonry towers: moisture diffusion, carbonation and size effect, Cement and Concrete Research, 36, pp. 1379–1388, 2006. [10] Binda, L., Lombardini, N. & Guzzetti, F., St. Vitale in Ravenna: a survey on materials and structures, Int. Conf. Historical Buildings and Ensembles, invited lecture, Karlsruhe, pp. 113–124, 1996. [11] Gordon, J.E., Strutture, ovvero Perche’ le cose stanno in piedi, Edizioni Scientiﬁ che e Tecniche, Mondadori, 1979.

[12] Taupin, J.L., Réfl exions sur la cathedrale Saint-Pierre de Beauvais. ANAGKH, 12, pp. 86–100, 1995. [13] Mainstone, R.J. Haghia Sophia: Architecture, Structure and Liturgy of the Justinian’s Great Church, Thames and Hudson, 1985. [14] Binda, L., Mirabella Roberti, G. & Guzzetti, F., St. Vitale in Ravenna: a Survey on materials and structures, International Symp. on Bridging Large Spans (BLS) from Antiquity to the Present, Istanbul, Turkey, pp. 89–99, 2000, ISBN 975-93903-02. [15] Jaeger, J.C. & Cook N.G., Fundamentals of Rock Mechanics, 2nd edn, Chapman & Hall: London, 1976. [16] Lenczner, D. & Warren, D.J.N., In situ measurement of long-term movements in a brick masonry tower block. Proceedings of the 6th IBMaC, Rome, pp. 1467–1477, 1982. [17] Anzani, A., Binda, L. & Mirabella Roberti, G., The behaviour of ancient masonry towers under long-term and cyclic actions, in Proc. Computer Methods in Structural Masonry – 4, Computer & Geotechnics: Swansea, pp. 236–243, 1998. [18] Fradeletto, A., et al., Il campanile di S. Marco riedifi cato. Studi, ricerche, relazioni, ed. Comune di Venezia, Carlo Ferrari: Venezia, 1912. [19] Binda, L., Tiraboschi, C. & Tongini Folli, R., On site and laboratory investigation on materials and structure of a bell-tower in Monza. Int. Zeitschrift für Bauinstandsetzen und baudenkmalpfl ege, 6, Jahrgang, Aedifi cation Publishers, Heft 1, pp. 41–62, 2000. [20] Binda, L., Tongini Folli, R. & Mirabella Roberti, G., Survey and investigation for the diagnosis of damaged masonry structures: the ‘Torrazzo of Cremona’. 12th Int. Brick/Block Masonry Conf., Madrid, Spain, pp. 237–257, 2000. [21] Binda, L. & Poggi, C., Ricerca volta a stabilire le condizioni statiche ed il comportamento meccanico della muratura del campanile del Duomo di Cremona. Relazione Finale, Contratto Consiglio della Chiesa Cattedrale di Cremona, 1999. [22] Scotti, A., L’età dei Borromei in Monza. Il Duomo nella storia e nell’arte, Electa: Milano, 1989. [23] Binda, L., Falco, M., Poggi, C., Zasso, A., Mirabella Roberti, G., Corradi, R. & Tongini Folli, R., Static and dynamic studies on the Torrazzo in Cremona (Italy): the highest masonry bell tower in Europe. Int. Symp. on Bridging Large Spans from Antiquity to the Present, Istanbul, Turkey, pp. 100–110, 2000.