PUBLICATION NO. FHWA-NHI- FHWA-NHI- Technical Report Documentation Page 1. Report No. 2. Government Accession No. 3. Recipient’s Catalog No.

FHWA-NHI-11-023

4. Title and Subtitle 5. Report Date

NATIONAL HIGHWAY INSTITUTE COURSE NO.130096 February 2012 FHWA - DESIGN GUIDELINES FOR ARCH AND CABLE-SUPPORTED 6. Performing Organization Code SIGNATURE BRIDGES

7. Author(s) 8. Performing Organization Report No.

Vijay Chandra, P.E. and Joseph Tse, P.E. See Acknowledgements for Recognition of Additional Authors

9. Performing Organization Name and Address 10. Work Unit No. (TRAIS)

Parsons Brinckerhoff, Inc. 11. Contract or Grant No. One Penn Plaza, New York, NY 10119 DTFH-61-D-00011/T-07-003

12. Sponsoring Agency Name and Address 13. Type of Report and Period Covered

National Highway Institute U.S. Department of Transportation 14. Sponsoring Agency Code

Federal Highway Administration, Washington, D.C. 20590

15. Supplementary Notes

FHWA COTR: Louisa Ward FHWA Task Manager: Firas I. Sheikh Ibrahim, Ph.D., P.E. See Acknowledgements for List of Additional Technical Reviewers

16. Abstract

These Design Guidelines have been developed to promote a uniform understanding of the state of practice in designing Arch, Cable-Stayed and Suspension Bridges. The provisions have been developed and formatted to supplement the AASHTO LRFD Bridge Design Specifications. The guidelines are, thus, presented in specifications-like language with supporting commentary to facilitate adaptation into project specific criteria where bridge owners deem applicable. Amongst cited references from the literature, national and international specifications, are reports on several investigations that were conducted as a part of developing these Guidelines including - review of HL-93 live load effects for designing long span structures, initial assessment of aerodynamic stability and bridge users’ comfort criteria, adaptation of the FHWA-TS-80-205 interaction equations for slender plate buckling as an extension of the AASHTO LRFD provisions, and other issues. Appendices to the Guidelines contain selected design topics deemed of immediate application to the readers covering the arch, cable-stayed, and suspension bridge types; general design process; -structure interaction; and bridge health monitoring.

17. Key Words 18. Distribution Statement

Arch, Cable-Stayed, Suspension, Long Span Bridges, Design, Construction, Loads, Load Factors, Load Combinations, No restrictions. Analysis, Aerodynamics, Seismic, Detailing

19. Security Classif. (of this report) 20. Security Classif. (of this page) 21. No. of Pages 22. Price

UNCLASSIFIED UNCLASSIFIED Approx. 350 Form DOT F 1700.7(8-72) Reproduction of completed page authorized FOREWORD

These design guidelines have been developed to promote a uniform understanding of the state of practice in designing arch, cable-stayed and suspension signature bridges. The word “signature” recognizes that these highly visible bridges are often considered iconic and a source of pride for the communities they serve. From the engineering perspective, these complex structures are generally of unique design, in specific configurations and proportions, for traversing challenging sites when the conventional structures will not deliver the required functionality at reasonable cost.

These Guidelines have been formatted to supplement the AASHTO LRFD Bridge Design Specifications. Built on the solid base of the AASHTO Specifications, provisions in this document highlight the need for gathering site-specific information; conducting appropriate levels of analysis for safe and optimized solutions; detailing for longer service life; and planning for future inspection and maintenance. It is our hope that these Guidelines will provide a framework for the preparation of project specific design and construction criteria, as as inspection, operation and maintenance manuals that would contribute to the making of safe and long lasting signature bridges.

Firas I Sheikh Ibrahim, Ph.D., P.E. M. Myint Lwin, P.E., S.E., Director Team Leader for Infrastructure Management Office of Bridge Technology

FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges

PREFACE

Design of complex bridges such as arch and cable-supported bridges is typically based on a set of project specific design criteria. These Design Guidelines have been prepared to promote basic uniformity in fundamental considerations, and may serve as a single-source reference when developing the project specific design criteria. The provisions are presented in specifications-like language with supporting commentary to facilitate adaptation into project specific criteria where bridge owners deem applicable.

Design topics in this document are presented in the same general sequence as the AASHTO LRFD Bridge Design Specifications (AASHTO LRFD), except that this document uses alphabetic designations for its various sections. Accordingly, these Guidelines are organized as presented below:

Section A is an introductory section that supplements Section 1 of AASHTO LRFD. It defines the scope of these Guidelines.

Section B supplements Section 2 of AASHTO LRFD. Project specific studies pertinent to signature bridges are outlined. Design life and serviceability issues are covered, highlighting strategies for extended design life, including the need to delineate between replaceable and non-replaceable elements in detail design.

Section C supplements Section 3 of AASHTO LRFD, covering loads, load factors and load combinations. Force effects that are especially pronounced on long span bridges are highlighted.

Section D expands upon the provisions of Section 4 of AASHTO LRFD. The treatment of nonlinearity and the limitation on superposition are addressed. The state of practice in considering soil-structure interaction and effective cross section are covered. Guidance on dynamic analyses and recommended structural damping are provided.

Section E supplements Section 5 of AASHTO LRFD. Limitations on tensile stresses in concrete elements and minimum requirements for reinforcement detailing for ductility are recommended.

Section F addresses the design of steel components. Resistance factors of the cable elements for suspension and cable-stayed bridges are presented in the context of the AASHTO LRFD specifications. The associated testing and quality control requirements are also presented. Provisions for checking slender steel panels are proposed.

FHWA-NHI-11-023 P-1 Preface February 2012 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges

Section G supplements Section 10 of AASHTO LRFD. This section covers geotechnical investigations and type selection for large structures. Additionally, provisions for design, construction, and integrity testing are recommended.

Appendix A presents information on selected suspension bridge topics that are most likely to be considered by the readers of these Guidelines. Topics include the sag of the main cables, sizing of structural elements, erection considerations, and corrosion protection of the cable systems.

Appendix B presents information on selected cable-stayed bridge topics. The special characteristics of the cable-stayed deck are presented, along with typical ratios between the main span and tower(s), cable anchorage details, construction considerations during design, and general discussions on deck replacement of cable-stayed bridges.

Appendix C presents information on selected arch supported bridge topics. Global geometry and load path are amongst the discussions. Other topics include the stability of the arch rib, bracing systems, and redundancy of the tie in a tied arch bridge. Distortion induced fatigue is also addressed in this Appendix.

Appendix D presents flow charts on the overall design of arch and cable-supported bridges, and wind engineering studies.

Appendix E provides a summary of the latest developments in the areas of bridge health monitoring. Materials include examples of bridge health monitoring systems that are being applied on major bridges.

Appendix F provides basic discussions and references on foundation modeling and consideration of soil-structure interaction.

FHWA-NHI-11-023 P-2 Preface February 2012 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges

ACKNOWLEDGEMENTS

The development of this document has been funded by the National Highway Institute, and supported by Parsons Brinckerhoff (PB), as well as numerous authors and reviewers acknowledged hereafter:

Guidelines Development Team Specialist Authors Professor Andrzej Nowak - Live Load and Load Combinations and Firas Ibrahim Stoyan Stoyanoff - Aerodynamics and Wind Load Karl Frank - Slender Steel Panel and Fatigue Specialists – Input and Review Akio Kasuga - Cable-Stayed Bridges and Stay Resistance Professor Yaojun Ge - Aerodynamics and Stability Robert Kimmerling - Geotechnical / Foundation Bala Sivakumar - WIM Data and Live Load Reid Castrodale - Concrete and LRFD Conformance PB Investigators John Bryson - Analysis Joe Wang and John Bryson - Seismic Design Chin Lien and Justin Lennon - Hydraulics and Waves Essam Bader - Steel Panel and Cable Resistance Ray Castelli and Sanjeev Malhotra - Geotechnical / Foundation Ruchu Hsu - Appendices A through C Roger Haight - Bridge Health Monitoring

FHWA Review Team / Technical Working Group

Myint Lwin (FHWA) Firas Ibrahim (FHWA) – Task Manager Tom Saad (FHWA) Susan Hida (Caltrans) George Christian (New York State DOT – Retired) Hossein Ghara (Louisiana DOTD) Gregor Wollmann and Ted Zoli (HNTB) Marcos Loizes (Jacobs) Khaled Mahmoud (Bridge Technology Consultants)

FHWA-NHI-11-023 P-3 Acknowledgments February 2012 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges

ACKNOWLEDGEMENTS (Cont.)

AASHTO LRFD Liaison on Loads and Load Factors John Kulicki (Modjeski and Masters) Professor Dennis Mertz (University of Delaware) Kevin Western (Minnesota DOT)

Last but not least, the Principal Investigators would like to extend our gratitude to Jeremy Hung for the overall guidance he had provided. Also, we would like to recognize Cassandra Leong for providing graphic support, and Isabel Lopes for assisting in editing the documents.

FHWA-NHI-11-023 P-4 Acknowledgments February 2012 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges

TABLE OF CONTENTS

1. Section A – Introduction 2. Section B – General Design & Location Features 3. Section C – Loads & Load Factors 4. Section D – Structural Analysis & Evaluation 5. Section E – Concrete Structures 6. Section F – Steel Structures 7. Section G – Foundations 8. Section A Commentary – Introduction 9. Section B Commentary – General Design & Location Features 10. Section C Commentary – Loads & Load Factors 11. Section D Commentary – Structural Analysis & Evaluation 12. Section E Commentary – Concrete Structures 13. Section F Commentary – Steel Structures 14. Section G Commentary – Foundations 15. Appendix A – Suspension Bridges 16. Appendix B – Cable-Stayed Bridges 17. Appendix C – Arch Bridges 18. Appendix D – General Design Process for Long Span and Complex Bridges 19. Appendix E – Bridge Health Monitoring 20. Appendix F – Foundation Modeling and Soil-Structure Interaction

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section A

TABLE OF CONTENTS

1.0 SCOPE...... 1 1.1 General ...... 1

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section A

SECTION A – INTRODUCTION

1.0 SCOPE

1.1 General

Unless otherwise stated or supplemented by these design guidelines for arch and cable- supported bridges (Guidelines), the provisions of the Fifth Edition of the AASHTO LRFD Bridge Design Specifications, with 2010 Interim Revisions (AASHTO LRFD) are applicable to the design of arch and cable-supported bridges.

The provisions of these Guidelines address the design of the following bridge types: • Suspension Bridges • Cable-Stayed Bridges • Arch Bridges

These Guidelines have been developed for bridges that support predominantly highway loading. Additional provisions shall be developed on a project specific basis where the bridges being designed are intended to support light rail and/or rail loading in addition to highway loading upon completion or in the future.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section B

TABLE OF CONTENTS

2.0 PROJECT SPECIFIC STUDIES ...... 1 2.1 Design Vessel & Vessel Impact Mitigation ...... 1 2.2 Stream Flow & Scour ...... 1 2.3 Geotechnical Considerations ...... 1 2.4 Effects ...... 1 2.5 Wind Effects ...... 2 2.6 Unusual Traffic Loading ...... 2 2.7 Aggressive Environment ...... 2 2.8 Blast and Deliberate Attacks ...... 2 3.0 DESIGN LIFE AND SERVICEABILITY ...... 2 3.1 Deformations ...... 2 3.1.1 Aerodynamic Stability ...... 2 3.1.2 Comfort and Deflection ...... 3 3.2 Durability ...... 3 3.2.1 Corrosion Protection Plan...... 3 3.2.2 Health Monitoring ...... 3 3.2.3 Replaceable Bridge Components...... 4 3.2.4 Maintainability ...... 4 4.0 REDUNDANCY ...... 5 5.0 CONSTRUCTABILITY ...... 5

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section B

SECTION B – GENERAL DESIGN & LOCATION FEATURES

2.0 PROJECT SPECIFIC STUDIES

A planned out design program shall be developed for arch and cable-supported bridges to incorporate as much site specific data as possible.

Design issues that warrant site specific studies include: • Design Vessel & Collision Mitigation • Stream Flow & Scour • Geotechnical Considerations • Earthquake Effects • Wind Effects • Unusual Traffic Loading • Aggressive Environment • Blast and Deliberate Attacks

2.1 Design Vessel & Vessel Impact Mitigation

Essential site specific data required to address vessel collision and mitigation shall be as provided in Article 3.4 of the AASHTO Guide Specifications and Commentary for Vessel Collision Design of Highway Bridges (2nd Edition) (B1).

2.2 Stream Flow & Scour

Unless supplemented by these Guidelines, the provisions of AASHTO LRFD - Article 2.6 shall apply for the design of long span bridges (B2).

2.3 Geotechnical Considerations

A subsurface investigation program shall be developed in conjunction with bridge design. Such a program shall reflect provisions in Section G, Article 26.1 of these Guidelines.

2.4 Earthquake Effects

Except as supplemented by these Guidelines, seismic design of long span bridges shall be in accordance with the following:

• AASHTO LRFD Bridge Design Specifications, 5th Edition, 2010, with 2010 Interim Revisions (B2)

• AASHTO Guide Specifications for LRFD Seismic Bridge Design, 1st Edition, 2009, with 2010 Interim Revisions (B9).

Where required, site specific analysis shall be conducted in accordance with applicable provisions of Section C, Article 12.0 of these Guidelines.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section B

2.5 Wind Effects

Wind climate studies and site analyses shall take into account - • Historical Data • On-site Wind Speed Measurements • Topographical Model Studies • Local Studies • Numerical Simulations

2.6 Unusual Traffic Loading

Site specific weigh-in-motion studies should be conducted for bridges planned in the vicinity of ports and goods movement hubs where extraordinary live loading and truck patterns could pose governing design conditions for critical structural elements.

2.7 Aggressive Environment

Bridges located in an aggressive environment shall be designed with multiple levels of corrosion protection. Protective measures shall be as indicated in AASHTO LRFD – Articles 2.5.2.1 and 5.12.

2.8 Blast and Deliberate Attacks

Blast loading and the effects of failure of one or more structural elements due to deliberate attacks shall be considered and assessed on a bridge-by-bridge basis.

3.0 DESIGN LIFE AND SERVICEABILITY

The minimum design life for long span bridges should be 100 years.

3.1 Deformations

Deflections, especially in long span bridges, shall be quantified by analyses in order that structural detailing, and the design of bearings, expansion joints and ancillary structures, can accommodate movements with adequate reserve.

3.1.1 Aerodynamic Stability

Structural integrity or serviceability may be compromised by the following phenomena and effects: • Vortex shedding • Flutter • Galloping • Torsional divergence • Cable • Interference effects

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section B

Structures shall be designed so that critical wind speeds for flutter and galloping are higher than the design wind speeds.

Peak dynamic rotation of the bridge deck around its longitudinal axis being greater than 1.5 degrees shall be considered the threshold for the onset of flutter instability.

3.1.2 Comfort and Deflection

Design shall make provisions to suppress noticeable oscillations/vibrations. In lieu of limits set forth in LRFD Article 2.5.2.6.2, design shall limit global movement at any portion of the bridge accessible by the public to the following: • Vertical accelerations:

amax ≤ 5% of g for winds below 30 mph;

amax ≤ 5%+0.25%(v-30) of g for wind speeds v up to 50 mph, following ASCE (B5).

• Lateral accelerations

where f is the bridge frequency in Hz; derived from ISO10137 (B6) for f ≤10 Hz.

The recommended acceleration limits can be converted into deflections by multiplying the 2 accelerations, and take the form of amax /(2πf) .

3.2 Durability

In addition to the provisions of AASHTO LRFD Article 2.5.2.1, the design of long span bridges shall address durability by explicitly distinguishing between non-replaceable and replaceable components of the bridge.

3.2.1 Corrosion Protection Plan

A corrosion protection plan should be in place as part of the design program to document the strategy for achieving the design life of the bridge.

Corrosion protection plans shall indicate predicted service lives of both replaceable and non- replaceable structural components. Materials specifications; special construction requirements; and inspection, testing, monitoring and maintenance requirements needed to achieve their predicted service lives shall also be included.

3.2.2 Health Monitoring

Where design life is predicated on minimum condition and performance of specific bridge components, such components shall be considered for instrumentation.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section B

3.2.3 Replaceable Bridge Components

In addition to material and construction specifications of replaceable components, the design document shall also include detailed performance requirements for these components.

Provisions for replacement of bridge components shall be incorporated in the design and indicated in the design plans.

3.2.4 Maintainability

Inspection access shall be an integral part of design. In addition to the standard ladders and platforms, consideration shall be given to incorporate access facilities including: • Inspection traveler(s) • Tower elevators • Gondola that may be needed for lifting maintenance equipment to high points of the bridge.

3.2.4.1 As-Built Records

As-built records shall be kept to sufficient detail to document the full stress history of the structure. Elements of the records may include: • As-Built Plans • Shop Drawings • Technical Fact Sheets • Test Reports • Material Certifications • Fabrication Data and Camber • Construction Engineering Analysis • Construction Manuals With Detailed Erection Sequence • Cable / Hanger Installation and Adjustment Records • Survey Records and Summaries • Retrofit and Rehabilitation Records • Rating Reports • Maintenance and Inspection Manuals • Inspection Records

3.2.4.2 Maintenance Manual

Contract documents shall specify the need for a maintenance manual. As a minimum, the manual shall – • Reference the Design Documents (inclusive of the Corrosion Protection Plan) • Reference the As-Built Records • Provide the specific time table and procedures for inspection, testing, maintaining and record keeping for the non-replaceable bridge components • Provide the specific time table and procedures for inspection, testing, maintaining, replacing and record keeping for the replaceable bridge components • Indicate the qualification of inspection and maintenance forces required for each activity

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section B

• Provide operational instructions and specifications of all mechanical and electrical equipment

4.0 REDUNDANCY

Long span bridges shall be designed such that controlled or sudden loss of a stay cable, individual hanger, or other structural element, will not cause collapse of the structure.

Tie girders for tied arch bridges shall be designed and detailed for redundancy. This may be accomplished through internal redundancy, external redundancy, or some combination of the two.

5.0 CONSTRUCTABILITY

Designer shall conduct sufficient studies and analysis to account for one constructible scheme where all locked in stresses associated with that construction scheme shown in the plans can be accommodated by the permanent design.

A feasible construction scheme that is consistent with the design intent should be shown in the design plans.

Where pertinent, the construction scheme shown in the plans should indicate the magnitudes and locations of special construction equipment, as well as special construction procedures that are of significance to the design.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section B

REFERENCES

B1. AASHTO Guide Specifications and Commentary for Vessel Collision Design of Highway Bridges, 2nd Edition, 2009 with 2010 Interim Revisions.

B2. AASHTO LRFD Bridge Design Specifications, Fifth Edition, American Association of State Highway and Transportation Officials (AASHTO), with Interim Revisions, 2010.

B3. PTI Guide Specification. Recommendations for stay cable design, testing and installation. Post-Tensioning Institute Committee on Cable Stayed Bridges, 5th edition, October, 2007.

B4. Irwin, P. , Stoyanoff, S., Xie, J., and Hunter, M. Tacoma Narrows 50 years later - wind engineering investigations for parallel bridges, Bridge Structures, pp. pp. 3-17, 2005.

B5. ASCE Committee on Loads and Forces on Bridges, Recommended Design Loads for Bridges. ASCE Proc., Vol. 107, No. ST7. December 1981.

B6. ISO/DIS10137, Bases for Design of Structures - Serviceability of Buildings against Vibrations, Int. Standard Organization, Geneva, Switzerland, 1992.

B7. British Standard BS5400, Steel Concrete and Composite Bridges: Specifications for Loads, Part 2, 1978.

B8. ENV1995-2, Design timber structures, Part 2: Bridges, European Committee for Standardization, Brussels, Belgium, 1995.

B9. AASHTO Guide Specifications for LRFD Seismic Bridge Design, First Edition, American Association of State Highway and Transportation Officials (AASHTO), 2009 with 2010 Interim Revisions.

B10. Stoyanoff, S., and M. Hunter, “Wind Engineering Design of Bridges – Shared Experience”, 9th UK Conference on Wind Engineering, 2010, University of Bristol.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section C

TABLE OF CONTENTS

6.0 SCOPE...... 1 7.0 NOTATION ...... 1 8.0 LOAD FACTORS AND COMBINATIONS ...... 1 8.1 Load Factors and Combinations in Service Conditions ...... 1 8.1.1 Strength IV Combinations ...... 1 8.1.2 Special Strength Combinations ...... 1 8.1.3 Extreme Event Combinations ...... 2 8.1.4 Seismic Load Combinations ...... 2 8.1.5 Combination of Seismic Force Effects ...... 2 8.1.6 Combinations Involving Superimposed Deformations ...... 3 8.1.7 Fatigue ...... 3 8.2 Load Factors and Combinations for Construction Loads ...... 4 8.2.1 Construction Loads and Sequence ...... 4 8.2.2 Overturning ...... 4 8.2.3 Strength Limit State ...... 4 8.3 Locked-in Stresses ...... 4 8.4 Stay and Suspender Adjustment Forces ...... 4 8.5 Loss of Structural Elements ...... 4 9.0 LIVE LOADS ...... 4 9.1 Multiple Presence of Live Load ...... 5 9.2 Dynamic Load Allowance ...... 5 9.3 Parade Loading ...... 5 9.4 Other Live Load ...... 5 10.0 WATER LOADS...... 5 10.1 Wave Load ...... 5 10.2 Hydraulic and Scour Analyses ...... 5 11.0 WIND EFFECTS...... 6 11.1 General ...... 6 11.2 Wind Climate Analysis and Local Turbulence Characteristics ...... 6 11.3 Testing Requirements...... 7 11.3.1 Section Model Tests ...... 7 11.3.2 Aeroelastic Model Tests ...... 7 11.3.3 Force-Balance Tests ...... 8 11.3.4 Field Tests ...... 8 11.4 Aerodynamic Analysis ...... 9 11.5 Wind Loads ...... 9 11.6 Component due to Wind Excitations ...... 10 11.7 Towers and Arch Ribs ...... 10 12.0 EARTHQUAKE EFFECTS ...... 10 12.1 Seismic Performance Zones and Performance Objectives ...... 10 12.2 Design Earthquake Hazard Levels ...... 11 12.3 Design Ground Motion Parameters ...... 11 12.3.1 Design Response Spectra Based on Site-Specific Procedure ...... 12 12.3.2 Design Response Spectra Based on General Procedure ...... 13 12.3.3 Acceleration and Displacement Time Histories ...... 15 12.4 Seismic Requirements for Temporary and Staged Construction ...... 16 13.0 FORCE EFFECTS DUE TO SUPERIMPOSED DEFORMATIONS ...... 17 13.1 Thermal Effects ...... 17 13.1.1 Temperature Difference Between Structural Elements ...... 17 13.2 Creep and Shrinkage ...... 17 13.3 Irreversible Displacement of Substructure(s) ...... 17 14.0 VESSEL COLLISION ...... 17

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section C

SECTION C – LOADS & LOAD FACTORS

6.0 SCOPE

Unless otherwise stated or supplemented by these Guidelines, the provisions of AASHTO LRFD (C01) Section 3 shall apply for the design of long span bridges.

7.0 NOTATION

Unless indicated otherwise, notations for loads and load factors are in accordance with AASHTO LRFD Article 3.3.

8.0 LOAD FACTORS AND COMBINATIONS

Unless circumstance justifies otherwise, long span bridges shall have the following load modifiers for ductility, redundancy and importance, ηD, ηR and ηI, respectively, per LRFD Articles 1.3.3, 1.3.4 and 1.3.5:

ηD = 1.0 ηR = 1.0 ηI = 1.05

8.1 Load Factors and Combinations in Service Conditions

8.1.1 Strength IV Combinations

In lieu of the load factor, γp, of 1.5 on the dead load DC, Strength IV shall be computed as follows:

1.4 DC + 1.4 (LL+IM) …………… (8.1-1)

where:

DC represents all applicable effects (DC, DD DW, EH, EV, ES, EL, PS, CR and SH) in LRFD Table 3.4.1-1.

LL+IM represent all applicable effects (LL, IM, CE, BR, PL and LS) in LRFD Table 3.4.1-1.

8.1.2 Special Strength Combinations

Stay Cable or Suspender Replacement:

1.2 DC + 1.4 DW + 1.5 (LL** + IM) + Cable or Suspender Exchange Forces …….. (8.1.2-1)

** At least one lane of live load is assumed shifted away from the element under replacement.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section C

8.1.3 Extreme Event Combinations

Combinations below shall be investigated in addition to those specified in the LRFD Table 3.4.1- 1.

8.1.3.1 Extreme Events and Corresponding Scour Conditions

Extreme events shall be investigated for appropriate scour conditions.

Extreme Wind and Scour condition:

1.0 x (DC + DW + EV + EL + WX) ….. (8.1.3-1)

8.1.3.2 Loss of Structural Element

Loss of an element within a component, or failure of a component without resulting in the collapse of the structure:

1.25 DC + 1.5 DW + 1.3 (LL+IM) ……. (8.1.3-2)

8.1.3.3 Loss of Stay Cable or Suspender

Stay Cable or Suspender Loss:

1.1 DC + 1.35 DW + 0.75 (LL* + IM) + 1.1 x (Cable Loss or Suspender Dynamic Forces) …. (8.1.3-3)

* Full live load placed in their actual striped lanes.

Resistance factor = 1.0 with vehicular loads taken to be in striped lanes.

8.1.4 Seismic Load Combinations

For the Extreme Event I load combination, use load factors of 1.0 for all permanent loads. Unless otherwise noted, resistance factors (Φ) shall all be taken as 1.0.

The load factor for live load applied simultaneously with seismic loads, γEQ, shall be determined on a project-specific basis or as directed by the owner.

8.1.5 Combination of Seismic Force Effects

Earthquake actions shall be determined in three orthogonal directions; the horizontal longitudinal and transverse axis of the structure and the vertical axis. The longitudinal axis of curved structures should generally be represented by a chord connecting the two abutments.

A combination of orthogonal forces shall be used to account for the directional uncertainty of earthquake motions and the simultaneous occurrences of earthquake forces in the three perpendicular directions. Elastic seismic forces, moments and displacements resulting from analyses in the three perpendicular directions shall be combined using the Square Root of the Sum-of-the-Squares (SRSS) method or the 30% combination rule. These two alternative combination rules may be summarized, as follows:

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SRSS Combination Rule (preferred):

Design elastic seismic forces, moments and displacements are computed as the square root of the sum of squares (SRSS) of the response quantity from the analyses in each of the orthogonal directions (longitudinal, transverse and vertical).

30% Combination Rule:

Design elastic seismic forces, moments and displacements resulting from analyses in the three perpendicular directions are combined to form three load cases:

• Seismic Load Case 1: Combine 100% of the absolute value of the forces, moments and displacements resulting from the longitudinal loading with 30% of the absolute value of the corresponding forces, moments and displacements resulting from the analyses in the transverse and vertical directions.

• Seismic Load Case 2: Combine 100% of the absolute value of the forces, moments and displacements resulting from the transverse loading with 30% of the absolute value of the corresponding forces, moments and displacements resulting from the analyses in the longitudinal and vertical directions.

• Seismic Load Case 3: Combine 100% of the absolute value of the forces, moments and displacements resulting from the vertical loading with 30% of the absolute value of the corresponding forces, moments and displacements resulting from the analyses in the longitudinal and transverse directions.

When non-linear time-history analysis is used, all three orthogonal components (longitudinal, transverse and vertical) of the design ground motion shall be input simultaneously. The design forces, moments and displacements shall be taken as the maximum response calculated for the three ground motions in each principal direction. If a minimum of seven time histories are used for each component of ground motion, the design actions may be taken as the mean response calculated for each principal direction.

When direct integration (step-by-step) time history analysis is used, components subject to multiple force components (e.g., axial force plus biaxial bending) can be checked to satisfy the design requirements using corresponding forces and moments at each time step.

8.1.6 Combinations Involving Superimposed Deformations

Load factors associated with superimposed deformations, γTU, γCR, and γSH shall be taken as 1.0 regardless of construction materials. The inelastic response of the structure to these load effects shall be quantified through structural analysis.

8.1.7 Fatigue

Cable elements and stay cables shall be designed for infinite life conforming to the AASHTO LRFD Fatigue I load combination and the modifications shown in Articles 8.1.7.1 and 8.1.7.2 of these Guidelines.

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8.1.7.1 Main Cable Elements and Suspenders

Load factor for fatigue in AASHTO LRFD Table 3.4.1-1 shall be modified as 1.0.

8.1.7.2 Stay Cables

Load factor for fatigue in AASHTO LRFD Table 3.4.1-1 shall be modified as 1.4.

8.2 Load Factors and Combinations for Construction Loads

8.2.1 Construction Loads and Sequence

Construction loads and sequence assumed in the design shall be shown in the contract plans.

8.2.2 Overturning

For balanced cantilever construction or similar methods, where stability against overturning is a primary concern, follow pertinent provisions in LRFD Articles 5.14.2.3.2 for loads and 5.14.2.3.4 for load combinations irrespective of construction material(s).

8.2.3 Strength Limit State

Additional construction load combinations shall be investigated at the strength limit state:

For foundation supporting elements:

1.25 x (DC + DW + CR + SH) + 1.50 x (CLE +IE+ CLL) + 1.25 x (WS + WE + WUP) + 1.0 (TU+TG) ………………..……. (8.2.3-1)

For cable-stays and suspenders:

1.2 x (DC + DW+ CE + CLE) + 1.2 x (WS + WE + WUP) + 1.4 CLL …………….. (8.2.3-2)

8.3 Locked-in Stresses

The effects of locked-in erection stresses shall be considered with a load factor of 1.0.

8.4 Stay and Suspender Adjustment Forces

The effects of adjusting cable stay and suspender forces shall be considered with a load factor of 1.0.

8.5 Loss of Structural Elements

The impact from a dynamic force may be determined by nonlinear dynamic analysis for a sudden rupture event, but in no case shall the dynamic impact factor be taken as less than 1.5.

9.0 LIVE LOADS

The HL-93 loading shall be used for the design of long span bridges.

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9.1 Multiple Presence of Live Load

Multiple presence of live load shall follow the AASHTO LRFD provisions, except that the multilane factors shall be applied separately for bridges carrying traffic in two directions.

9.2 Dynamic Load Allowance

Dynamic load allowance shall be in accordance with LRFD Article 3.6.2, unless a site specific increase in live load requirements warrants refinement for long span action.

9.3 Parade Loading

Parade loading shall be considered where it is reasonable to expect such a gathering, planned or spontaneous, at the bridge site.

Parade loading shall be established based on anticipated behavior of the occupants. Loading shall be placed either uniformly or in checker pattern, so as to produce the most unfavorable loading conditions.

Parade loading shall be substituted as LL+I and checked under Strength II and Service II of the LRFD Table 3.4.1-1.

9.4 Other Live Load

Effects of maintenance traveler, allowance for repair loads, and inspection vehicles shall be considered in design.

When these transient loads are combined with reduced vehicular/traffic loads, the assumed loads and loading pattern shall be specified in the construction plans, and inspection and maintenance manual.

10.0 WATER LOADS

10.1 Wave Load

Wave loads for long span bridge design shall consider advanced numeric modeling of wave fields.

Bridge superstructures shall include design considerations for wave effects. Wave considerations include either elevation above the maximum wave level with acceptable clearance or design of the superstructure for wave impact.

Bridge piers and abutments shall include design considerations for wave impact. Impact analyses shall include horizontal impact forces, vertical forces, and vertical slamming forces.

10.2 Hydraulic and Scour Analyses

Hydraulic analyses for long-span bridges shall include consideration of advanced two- dimensional hydrodynamic modeling for design storms.

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Bridges with foundations supported on shall have the erodibility/scour susceptibility of the bedrock evaluated using published methods. Foundation design shall consider removal of bedrock found to be susceptible to scour.

Coastal scour analyses shall include time-dependent considerations in analysis of potential scour depths. Time-dependent analyses shall be developed for both contraction and local scour predictions.

Geomorphic evaluations of long-term stream stability shall be performed for design of bridge foundations. Evaluations shall include investigation of long-term bed degradation/aggradation and channel lateral migration potential. Bridge piers and abutments shall be designed considering the channel evolution potential.

Scour depths used for structural design and stability check shall be indicated on the construction plans.

11.0 WIND EFFECTS

11.1 General

Aerodynamic stability and comfort criteria for long span bridges shall be confirmed by an appropriate combination of wind studies, physical model testing and numerical analyses.

Wind loadings for long span bridges shall be established following confirmation of the aerodynamic stability, and shall take into account the effects noted throughout the wind studies, testing, and numerical analyses outlined in these provisions.

11.2 Wind Climate Analysis and Local Turbulence Characteristics

A wind climate analysis of the site shall be conducted to document - • Sources of wind data • Methodology employed to determine design wind speeds • The wind specialist’s recommendations regarding –

o Mean-hourly speeds at 30 ft height in open terrain o Wind directionality effects o 10 minute mean speed including the effect of wind directionality o Terrain and turbulence properties at the bridge site o Design wind speeds, for aerodynamic stability as well as wind loads, given at deck height o Terrain roughness power law constant for wind profile(s) selected for design o Expected deviations of the mean angle of wind attack from the horizontal, if applicable

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Results from wind climate analysis shall be reported in sufficient detail, using tables and diagrams, where necessary:

• Wind speeds shall be given for return periods starting from 0.1 up to 10,000 years • A description of the bridge site shall be given with a general classification, i.e., open, suburban terrain etc. • When describing wind directionality, power law constants shall be attributed to various wind directions • In complex topography, its effect on the mean wind and gust speed profiles shall be described • If present, the effect of large structures in the vicinity shall be described • Special site effects shall be addressed including shallow waters that, depending on temperature, affect turbulence, mean profiles and wind inclination • Turbulence shall be described as a minimum by - o Referencing mean wind speed, elevation and terrain roughness o Turbulence intensities and length scales and/or exponential coefficients of coherence

11.3 Testing Requirements

Wind tests shall be conducted on properly scaled models including but not limited to:

• Sectional Model Tests • Aeroelastic Model Tests • Force-Balance Tests

When ice and snow effects are investigated, the probability of simultaneous occurrence of the worst case scenario must be investigated.

Scope, objectives and methods of testing shall be defined individually by wind engineering specialists for each project.

11.3.1 Section Model Tests

Sectional model tests shall be carried out on as large a scaled model as possible.

Sectional model tests shall be performed in both smooth and turbulent flows, covering wind speeds from the lowest practical range, up to and beyond the project specific stability criteria.

Structural damping ratios shall be set in the range of their expected full scale values.

Static force and moment coefficients shall be measured for wind angles ranging, as a minimum, between ±10° in increments of 2°.

11.3.2 Aeroelastic Model Tests

Aeroelastic model tests shall be performed on bridges with expected strong aeroelastic responses.

The scaling principle shall be followed when testing suspension bridges.

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The following shall be the basic requirements of an aeroelastic model test:

• All masses on the bridge attached to elastic parts of the structure shall be appropriately represented • Site turbulence shall be modeled as close as possible to the wind study recommendations • Tests shall be carried out for various wind directions to capture the potentially governing conditions • Tests shall be performed at small speed increments when aerodynamic instability and/or interference is a concern. When structural and site symmetry is present, the number of tested wind directions may be reduced. • Local topography and structures of significance in the vicinity shall be represented (in case of bridge replacement projects, both the effects on the existing and the replacement structure shall be considered) • Loads at the tower bases shall be measured directly • Structural deflections and accelerations shall be measured at key locations such as the top of the towers and the mid-point of the main span • Wind speed shall be varied, during testing, in fine increments starting from the lowest practical minimum, below the expected vortex shedding speed, to speeds higher than the required flutter criterion • When vortex shedding response is found, the wind speed shall be adjusted to establish the peak response • Effects of ice and snow accumulations shall be studied for both stability and loading • Traffic congested conditions and wind shall be investigated where applicable • Construction stage tests shall include all construction equipment such as temporary scaffolding, frameworks, and lifting equipment

Corrections to the full scale response predictions shall be applied when turbulence has to be scaled down in the wind tunnel simulations. Such corrections shall be derived from theoretical response estimates and calibrated in the model to represent turbulence in the full scale.

Relevant statistics such as mean, root-mean-square (variance) and true peak values shall be presented along with measurements taken during the test.

All test results shall be extrapolated and presented at full scale. When corrections to the desired full scale parameters are undertaken, the correction coefficients and/or formulae background shall be included in the test report.

11.3.3 Force-Balance Tests

When used to identify external dynamic loads associated with on-coming and signature turbulence (induced by the section), the models shall be geometrically correct and constructed of high strength and light weight materials for rigidity and reduced inertia effects.

11.3.4 Field Tests

Field tests on elements of the completed structure shall be specified if deemed necessary.

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11.4 Aerodynamic Analysis

Aerodynamic derivatives used to determine flutter speed analytically shall be extracted from testing at wind speeds beyond the flutter speed criteria. At least 8 derivatives shall be identified for 2 degree-of-freedom (2-dof) tests, and 18 derivatives shall be identified from 3-dof tests.

If a free vibration method is used, aerodynamic derivatives shall be extracted in smooth flow.

See Section D, Article 17.2 of these Guidelines for additional provisions.

11.5 Wind Loads

Wind loads for long span bridges shall include –

• Mean loads • Direct gust loads • Motion-induced loads

Mean and direct gust loads shall be applied to all elements of the bridge that are exposed to wind.

Motion-induced loads, caused by the acceleration of the bridge mass, shall be applied at the center of mass for every vibrating bridge element.

The direct gust and motion-induced loads shall be derived from • Buffeting analysis • Aeroelastic model tests

Mean, direct and motion-induced loads shall be combined as equivalent static loads in various wind load combinations to achieve the worst loading effects on all structural elements.

When wind loads are simulated and applied directly in time domain, special attention shall be given to the introduction of aerodynamic damping into the analysis.

Buffeting analysis predictions shall be applied for the estimation of direct gust loads and motion- induced loads. Buffeting loads shall be reassessed when structural geometry, stiffness and/or mass are modified during the design process.

Wind loads in regions prone to severe ice and snow accumulation shall be investigated and covered by project specific criteria.

Long span bridges that are expected to experience heavy traffic congestions shall also be covered by project specific criteria. The Designer shall develop project specific traffic patterns based on expected vehicle sizes and densities.

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11.6 Component Vibration due to Wind Excitations

Flexible structural components (cable stays, suspenders, etc.) that are prone to vibration due to wind excitation shall be designed to accommodate dynamic and static wind pressure, buffeting, vortex shedding, wake galloping and rain-wind induced vibrations.

If an aerodynamic countermeasure is proposed to mitigate vibrations, the desired effect shall be confirmed by experimental studies.

Both the completed bridge and the bridge under construction shall be considered.

Parametric excitation and linear resonance shall be considered.

Effects that may alter the surface roughness and shape of flexible elements, such as accumulation of ice on cables, shall be considered in the design.

11.7 Towers and Arch Ribs

Wind load on towers and arch ribs shall take into account dynamic and static wind pressure, buffeting, and vortex shedding.

Shading (sheltering) effects shall not be considered unless verified by tests.

Both the completed bridge and the bridge under construction shall be considered.

12.0 EARTHQUAKE EFFECTS

12.1 Seismic Performance Zones and Performance Objectives

Each bridge shall be assigned to one of four seismic performance zones, 1 through 4, based on the 1-second period design spectral acceleration for the design earthquake (SD1) as shown in Table 12.1-1, below.

If there is a potential for liquefaction-induced lateral spreading or slope failure that may impact the stability of the bridge, the bridge shall be designed in accordance with Seismic Zone 4, regardless of the magnitude of SD1.

Table 12.1-1 Seismic Performance Zones

When considering provisions from the AASHTO Guide Specifications for LRFD Seismic Bridge Design (C28), the partition between SDC A and SDC B in the guide specifications shall be lowered from 0.15g to 0.10g.

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Long span bridges shall be designed to achieve the following performance objectives for the two levels of design ground motions defined in Section 12.2.

Safety Evaluation Earthquake

• Minimal to moderate (repairable) damage to primary members; • Moderate (repairable) damage to other members and components; • Emergency/defense vehicles should be able to access the bridge immediately or within a few hours, after inspection. Vehicles may have to travel at reduced speeds; • General traffic should be able to access the bridge within weeks to months.

Functional Evaluation Earthquake

• No to minimal damage to primary members; • Minimal to moderate (repairable) damage to other members and components; • All traffic should be able to access the bridge immediately or within a few hours (with no reduction in speed), after inspection.

Higher levels of performance (such as no to minimal damage to primary members under the Safety Evaluation Earthquake) or lower levels of performance (such as significant damage, no- collapse/life safety under the Safety Evaluation Earthquake) may be established and used with the authorization of the bridge owner.

Unless otherwise specified by provisions of this document, seismic performance shall be assured by verifying that displacements are limited to satisfy geometric, structural and foundation constraints on performance.

12.2 Design Earthquake Hazard Levels

Long span bridges shall be designed for two levels of earthquake ground motion hazards:

• Upper level Safety Evaluation Earthquake (SEE) with a 4% probability of exceedance in 100 years (≈2500-year return period), or 1.5 times median deterministic value

• Lower level Functional Evaluation Earthquake (FEE) with a 50% probability of exceedance in 100 years (≈150-year return period).

Higher or lower levels of design earthquake ground motion hazards for both SEE and FEE may be established and used with the authorization of the bridge owner.

In addition, a one-level design earthquake approach may also be adopted and its design hazard level and corresponding performance requirements be established with the authorization of the bridge owner.

12.3 Design Ground Motion Parameters

The ground shaking hazard prescribed in these Specifications is defined in terms of acceleration response spectra which shall be determined in accordance with the site-specific procedure of Article 12.3.1, especially when any of the following conditions exists:

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• A very high degree of confidence of meeting the seismic performance objectives of Section 12.1 is desired. • The site is located within 6 miles of a known active fault and its response could be significantly and adversely influenced by near-fault ground motion characteristics.

If appropriate maps or results corresponding to the design earthquake hazard levels are available and their use is approved by the owner, then the design response spectra may be derived in accordance with the general procedure of Section 12.3.2. The appropriate seismic hazard maps and results may include the latest USGS seismic hazard maps or hazard results produced by the U.S. Geological Survey depicting probabilistic ground motion and spectral response.

In addition, site-specific dynamic soil response analysis to study the effects of the local soil/site conditions (site effects) shall be performed for long span bridges unless the use of the general procedure described in Section 12.3.2.1 is specifically approved by the owner.

Design acceleration time histories shall be derived in accordance with the requirements of Section 12.3.3.

12.3.1 Design Response Spectra Based on Site-Specific Procedure

A site-specific probabilistic ground-motion analysis shall be conducted in a manner to generate a uniform-hazard acceleration response spectrum for spectral values over the entire period range of interest of the proposed long span bridge. This analysis shall establish the following:

• The contributing seismic sources, • An upper-bound earthquake magnitude for each source zone, • Median attenuation relations for acceleration response spectral values and their associated standard deviations, • A magnitude-recurrence relation for each source zone, and • A fault-rupture-length relation for each contributing fault.

Uncertainties in source modeling and parameter values shall be taken into consideration. Detailed documentation of ground-motion analysis shall be provided and shall be peer reviewed.

For sites located within 6 miles of an active surface or shallow fault, studies shall be considered to quantify near-fault effects on ground motions to determine if these could significantly influence the bridge response. The fault-normal component of near-field (D < 6 miles) motion may contain relatively long-duration velocity pulses which can cause severe nonlinear structural response, predictable only through nonlinear time-history analyses. For this case, the recorded near-field horizontal components of motion shall be transformed into principal components before they are modified to be response-spectrum-compatible.

Deterministic spectra may be utilized in regions having known active faults if the deterministic spectrum is no less than 2/3 of the probabilistic spectrum in the region of 0.5TF to 2TF of the spectrum, where TF is the bridge fundamental period. Where use of deterministic spectra is appropriate, the spectra shall be either:

• The envelope of median spectra calculated for characteristic maximum magnitude on known active faults; or

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• Deterministic spectra may be defined for each fault, and in the absence of clearly controlling spectra, the spectra for each fault should be used.

Site-specific dynamic soil response analysis to study the effects of the local soil/site conditions (site effects) shall be performed based on site-specific geotechnical investigations.

12.3.2 Design Response Spectra Based on General Procedure

If the general procedure is considered appropriate and selected for use, the design response spectra for the SEE and FEE shall be constructed using the latest USGS national ground motion maps or hazard results in accordance with the procedure described in Article 3.4.1 of the AASHTO Guide Specifications for LRFD Seismic Bridge Design (C35).

12.3.2.1 Site Effects

Site effects shall be evaluated for long span bridges by performing site-specific dynamic soil response analysis in accordance with the requirements in Section 12.3.1, unless the use of the general procedure described in this Section is specifically approved by the owner.

The site effects described in this section shall be used with the general procedure for constructing response spectra described in Section 12.3.2. If geological conditions at the abutments and intermediate piers result in different following the general procedure described herein, then the design response spectra may be determined based upon the site-specific procedure. In lieu of the site-specific procedure and under guidance from the geotechnical engineer, the design response spectra may be determined as the envelope of the individual response spectra at each support.

12.3.2.2 Site Class Definitions

The site shall be classified as one of the site classes defined in AASHTO LRFD Table 3.10.3.1- 1 (C01). Where the soil shear wave velocity is not known, the site class shall be determined, as permitted in Table 3.10.3.1-1, from standard penetration resistance, or from soil undrained . When the soil properties are not known in sufficient detail to determine the site class, Site Class D shall be used. Site Classes E or F need not be assumed unless the owner determines that Site Classes E or F may be present at the site or in the event that Site Classes E or F are established by geotechnical data.

The steps for classifying a site shall be as follows:

Step 1: Check for the three categories of Site Class F requiring site-specific evaluation. If the site corresponds to any of these categories, classify the site as Site Class F and conduct a site-specific evaluation.

Step 2: Check for the existence of a total thickness of soft > 10 ft where a soft clay layer is defined by: < 500 psf, w > 40 percent, and PI > 20. If these criteria are satisfied, classify the site as Site Class E.

Step 3: Categorize the site using one of the following three methods with , N , and computed in all cases as specified below.

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Method 1: for the top 100 ft method)

is the generalized shear wave velocity for the upper 100 ft of the soil profile defined as

(12.3.2.2-1) where

νsi is the shear wave velocity of layer i in ft per second

th di is the thickness of any i soil layer in ft

n is the total number of distinct soil layers in the upper 100 ft

Method 2: N for the top 100 ft (N method)

N is the generalized standard penetration resistance of all in the upper 100 ft of the soil profile defined as

(12.3.2.2-2)

Ni is the standard penetration resistance of layer i (ASTM D1586-84), not to exceed 100 blows/ft, as directly measured in the field without corrections.

Method 3: for cohesionless soil layers (PI<20) in the top 100 ft and average, , for cohesive soil layers (PI>20) in the top 100 ft ( method)

is the generalized standard penetration resistance for only the cohesionless soil layers of the upper 100 ft of the soil profile defined as

(12.3.2.2-3)

where ds is the total thickness of cohesionless soil layers in the top 100 ft.

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includes cohesionless soil layers only when calculating Nch

is the generalized undrained shear strength for only the cohesive soil layers of the upper 100 ft of the soil profile defined as

(12.3.2.2-4)

where dc is the total thickness (100-ds) of cohesive soil layers in the top 100 ft. sui is the undrained shear strength in psf, not to exceed 5,000 psf, as determined by ASTM D2166-91 or D2850-87.

includes cohesive soil layers only

If Method 3 is used (i.e., the method) and the and criteria differ, select the category

with the softer soils (for example, use site class D instead of C).

The shear wave velocity for rock, Site Class B, shall be either measured on site or estimated by a geotechnical engineer or engineering geologist/ seismologist for competent rock with moderate fracturing and weathering.

The hard rock, Site Class A, category shall be supported by shear wave velocity measurements either on site or on profiles of the same rock type in the same formation with an equal or greater amount of weathering and fracturing.

12.3.2.3 Site Coefficients and Adjusted Spectral Response Acceleration Parameters

Site Coefficients for the (Fpga), short-period range (Fa), and long period range (Fv) shall be taken as specified in Article 3.4.2.3 of the AASHTO Guide Specifications for LRFD Seismic Bridge Design (C28).

12.3.3 Acceleration and Displacement Time Histories

The time histories developed shall have characteristics that are representative of the seismic environment of the site and the local site conditions.

Response-spectrum-compatible time histories shall be used as developed from representative recorded motions. Analytical techniques used for spectrum matching shall be demonstrated to be capable of achieving seismologically realistic time series that are similar to the time series of

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the initial time histories selected for spectrum matching.

Where recorded time histories are used, they shall be scaled to the approximate level of the design response spectrum in the period range of significance for the bridge. Each time history shall be modified to be response-spectrum compatible using the appropriate procedure.

At least three response-spectrum-compatible time histories and one response-spectrum- compatible time history shall be used for each component of motion in representing the SEE and FEE design earthquakes, respectively. The issue of requiring all three orthogonal components (x, y, and z) of design motion to be input simultaneously shall be considered as a requirement when conducting a nonlinear time-history analysis. The design actions shall be taken as the maximum response calculated for the three ground motions in each principal direction. If a minimum of seven time histories are used for each component of motion, the design actions may be taken as the mean response calculated for each principal direction.

For near-field sites (D < 6 miles) the recorded horizontal components of motion selected should represent a near-field condition and should be transformed into principal components before making them response-spectrum-compatible. The major principal component should then be used to represent motion in the fault-normal direction and the minor principal component should be used to represent motion in the fault-parallel direction.

Long span bridges shall be evaluated for spatially varying ground motion effects. In deriving the spatially varying ground motion time histories, as a minimum the following three effects shall be taken into consideration:

• Soil site effect; • Wave traveling effect; and • Near-field effect.

12.4 Seismic Requirements for Temporary and Staged Construction

Any bridge or partially constructed bridge that is expected to be temporary for more than five years shall be designed using the same requirements for permanent bridges and shall not use the provisions of this Article.

An earthquake shall not cause collapse of all or part of a temporary bridge that is expected to carry traffic. This requirement shall also apply to bridges that are constructed in stages and expected to carry traffic and/or pass over routes that carry traffic. The elastic seismic response coefficient and the ground acceleration coefficient given in Article 12.3.2.3 may be reduced by a factor of not more than 2 in order to calculate the component elastic forces and displacements. Response and acceleration coefficients for construction sites that are close to active faults shall be the subject of special studies.

The response modification factors given in Section D may be increased by a factor of not more than 1.5 in order to calculate the design forces. This factor shall not be applied to connections.

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13.0 FORCE EFFECTS DUE TO SUPERIMPOSED DEFORMATIONS

13.1 Thermal Effects

In addition to the LRFD provisions, uniform temperature variations within a structural component shall be considered.

13.1.1 Temperature Difference Between Structural Elements

For suspension bridges, design temperature differential shall be specified between hangers and girder, between hangers and main cables, and between girder and main cable.

For cable-stayed bridges, minimum temperature differences shall be checked at the service limit state between the following:

• Stay cables and the rest of the structure • Stays left and right of pylons (as seen in longitudinal elevation)

For steel arch ribs and orthotropic steel box deck, differential gradients within the element shall be specified for checking stress at the service limit state.

For all long span types with a steel deck, a temperature difference shall be specified between the steel deck and the rest of the structure for checking at the service limit state.

A temperature difference shall be specified over the width of the deck for checking at the service limit state.

Temperature difference through the cross-section of the concrete tower shall be specified for checking at the service limit state.

13.2 Creep and Shrinkage

The effect of reinforcement shall be assessed when considering creep and shrinkage of concrete.

13.3 Irreversible Displacement of Substructure(s)

Load factors in load combinations associated with effects of irreversible displacement(s) shall be 1.0 at both strength and service limit states.

14.0 VESSEL COLLISION

Long span bridges over major shipping channels shall consider site specific studies to ensure a safe and cost effective solution.

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REFERENCES

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C04. Hida, S.E., Statical Significance of Less Common Load Combinations, Journal of Bridge Engineering, ASCE, May/June 2007.

C05. Zoli, T., and Woodward, R., “Design of Long Span Bridges for Cable Loss,” IABSE Symposium Lisbon 2005, Lisbon, 2005.

C06. Nowak, A. S., Lutomirska, M, and Ibrahim, F, “The Development of Live Load for Long Span Bridges,” New York City Bridge Conference, 2009.

C07. Nowak, A. S., “Calibration of LRFD Bridge Design Code, NCHRP Project 12-33,” University of Michigan, Ann Arbor, MI, 1992.

C08. Dallard, P., et al, The London Millenium Footbridge, September 2001.

C09. Parsons Brinkerhoff / T. Y. Lin International, A Joint Venture, Seismic Study and Retrofit Design of the Verrazano-Narrows Bridge, Task 9 Report – Marathon Evaluation, November 2005.

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C11. ASCE/SEI 7-05, Minimum Design Loads for Buildings and Other Structures, ASCE, American Society of Civil engineers, 2006.

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C13. Somerville, P. G., N. G. Smith, R. W. Graves, and N. A. Abrahamson, “Modification of Empirical Strong Ground Motion Attenuation Relations to Include the Amplitude and Duration Effects of Rupture Directivity”, Seismological Research Letters, 1997.

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C16. Abrahamson, N.A. “Generation of spatially incoherent strong ground motion time histories,” Proceedings: Earthquake Engineering, Tenth World Conference, Balkema, Rotterdam, 1992.

C17. Gasparini, D.A., Vanmarcke, E.H. (1976), Simulated earthquake motions compatible with prescribed response spectra, MIT Civil Eng. Research Report R76-4, Cambridge, Massachusetts, 1976.

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C20. BS 5400-4:1990, Steel, Concrete and Composite Bridges – Part 4: Code of Practice for Design of Concrete Bridge (Appendix C), British Standards Institute, 1990.

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C24. Singh, L., Jones, N., Scanlan, R., and Lorendeaux, O. Simultaneous identification of 3-dof aeroelastic parameters, In Proc. 9th International Conference on Wind Engineering, p. 972-981, New Delhi, India, 1995.

C25. Davenport, A.G., The response of Slender Line-Like Structures to a Gusty Wind, Institute Civil Eng. 23, 389-408, 1962.

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C29. Froehlich, D. C. (2002) Finite Element Surface Water Modeling System (FESWMS) Two-dimensional Depth-averaged Flow and Sediment Transport Model (FST2DH). U.S. Department of Transportation, Federal Highway Administration.

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C32. Annandale, G., and Smith, S. (2001) “Calculation of Bridge Pier Scour using the Erodibility Index Method”, Report No. CDOT-DTD-R-2000-9, Department of Transportation Research, U.S. Department of Transportation, Federal Highway Administration.

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C35. AASHTO Guide Specifications for LRFD Seismic Bridge Design, First Edition, American Association of State Highway and Transportation Officials (AASHTO), 2009 with 2010 Interim Revisions.

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C37. LEE D.M., and PEIRIS N., “Modelling of Ship Impact on a Bridge Foundation”, IABSE Symposium Shanghai 2004: Metropolitan Habitats and Infrastructure, IABSE Report, Volume 88, 2004.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section D

TABLE OF CONTENTS

15.0 GENERAL ...... 1 15.1 Nonlinearities ...... 1 15.1.1 Treatment of Nonlinearities ...... 1 15.1.2 Superposition ...... 1 15.2 Construction Analysis During Design Stage ...... 1 15.3 Construction Engineering and As-Built Analysis ...... 2 15.4 Foundation Modeling and Soil-Structure Interaction ...... 2 15.4.1 Pile / Pile Group Analysis ...... 2 15.4.2 Lateral Stiffness of Pile Caps ...... 2 15.4.3 Design of Pile Foundations Subject To Seismic Ground Motions ...... 2 15.4.4 Effective Support Motions ...... 2 15.5 Effective (Partially Cracked) Section Properties ...... 2 16.0 STATIC ANALYSIS...... 3 16.1 Approximate Methods ...... 3 16.1.1 Effective Flange Widths for Concrete Deck Slab ...... 3 16.1.2 Effective Flange Widths for Orthotropic Deck ...... 3 16.2 Stability and Buckling ...... 3 16.2.1 Second Order Analysis ...... 3 16.2.2 Arch Rib and Towers ...... 4 16.2.3 Cable-Stayed Deck ...... 4 16.2.4 Orthotropic Girders ...... 4 17.0 DYNAMIC ANALYSIS ...... 4 17.1 Structural Damping ...... 4 17.2 Aerodynamic Analysis ...... 5 17.3 Seismic Analysis ...... 6 17.3.1 Multimode Response Spectrum Analysis ...... 6 17.3.2 Time History Analysis ...... 6 17.3.3 Inelastic Static (Pushover) Analysis ...... 7 17.3.4 Moment-Curvature Analysis ...... 8 17.3.5 Hydro-Dynamic Effects...... 8 17.3.6 Miscellaneous Seismic Modeling Details ...... 8 17.3.7 Design Displacements ...... 8

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section D

SECTION D – STRUCTURAL ANALYSIS & EVALUATION

15.0 GENERAL

The requirements of AASHTO LRFD Section 4 shall apply to the design of long span bridges.

Separate analyses shall be performed for each level of scour when designing for extreme events and wind involving scour conditions.

The interaction between structure and surrounding geotechnical material shall be considered in the modeling of each scour level.

15.1 Nonlinearities

The global analytical model(s) for long span bridges shall address all significant structural and material nonlinearities including: • Cable sag effects • Geometric nonlinearities (P-delta effects) • Material nonlinearities

15.1.1 Treatment of Nonlinearities

Dead loads shall be applied to the model in the order of actual construction. Construction loads that will cause locked-in stresses in the final structure shall also be applied according to the expected sequence of application.

Where combinations of factored loads are to be computed, the individual load effects shall each be factored and sequentially loaded until all required loadings for a given combination have been applied. See Section C of these Guidelines for load factors.

Large loadings, such as the unloading of a superstructure constructed on falsework onto a suspension system, shall be applied analytically in small enough increments to capture the nonlinear effects.

15.1.2 Superposition

Superposition shall not be used for the design of long span bridges unless it can be demonstrated that loadings involved in the design are linear or essentially linear in characteristic.

15.2 Construction Analysis During Design Stage

Adequate construction analysis shall be performed to ensure that the final structure can accommodate dead load stress history effects (i.e. the effects of locked-in stresses associated with staged erection), and that the structure has a reasonable reserve capacity that accounts for construction tolerances.

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15.3 Construction Engineering and As-Built Analysis

Full stage-by-stage analyses shall be performed during the detailed construction planning stage and updated for the documentation of the as-built structure, where all construction sequence and effects from construction equipment shall be incorporated. See Section B, Article 3.2.4 – Maintainability – of these Guidelines.

15.4 Foundation Modeling and Soil-Structure Interaction

Soil-structure interaction effects shall be considered in all analyses for the design of long span bridges. Global analytical models shall include an adequate representation of soil-structure interaction effects. The soil-structure interaction problem can be solved using either de-coupled or fully-coupled analyses.

15.4.1 Pile / Pile Group Analysis

For final design or detailed evaluations, stiffness coefficients shall be obtained by performing laterally loaded pile or pile group analyses, using non-linear springs (i.e., p-y and t-z curves) distributed along the pile length.

15.4.2 Lateral Stiffness of Pile Caps

The lateral stiffness of the pile cap shall incorporate the soil resistance on its front vertical face as part of the total stiffness. Because of potential interaction between the pile cap and the supporting piles, soil resistance at the bottom and on the two side surfaces shall be ignored.

15.4.3 Design of Pile Foundations Subject To Seismic Ground Motions

Pile design shall consider two loading effects associated with seismic events: 1. the inertial effect from the structure, and 2. the kinematic effect from the ground displacements.

15.4.4 Effective Support Motions

The soil-structure interaction analysis shall include an evaluation of the appropriate (effective) level for application of the ground motions.

15.5 Effective (Partially Cracked) Section Properties

The analysis of long span bridges shall reflect effective section properties that take into account cracks in traditional reinforced concrete and prestressed concrete elements, including the superstructure and substructure (pier and tower columns, drilled shafts, etc.).

Effective stiffness, EIeff, of concrete elements should be based on evaluation of the moment- curvature relationship for the particular section under consideration.

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16.0 STATIC ANALYSIS

16.1 Approximate Methods

16.1.1 Effective Flange Widths for Concrete Deck Slab

Section properties shall be modified progressively in analysis to reflect staged construction for dead load effects. The level of detail shall vary according to provisions indicated in Articles 15.2 and 15.3 of these Guidelines.

In lieu of more refined analyses, the superstructure shall be represented by beam elements with effective flange widths based on Article 4.6.2.6 of AASHTO LRFD, with the exception of decks for cable-stayed bridges.

In lieu of more refined analyses, cable-stayed bridge decks shall be analyzed with two different effective flange widths, one for axial force effects and another for bending moment effects. Effective widths may be approximated according to Article 4.6.2.6 of AASHTO LRFD and the following:

• Bending effects – li, the notional span length, to vary according to the bending effects • Axial effects – distribution of cable forces at the deck to follow Figure 4.6.2.6.2-4 of AASHTO LRFD

16.1.2 Effective Flange Widths for Orthotropic Deck

The effective flange width of deck plate for bending effects shall be as given in Article 4.6.2.6.4 of AASHTO LRFD.

The methodology in separating bending and axial effects for cable-stayed decks shall be as indicated in Article 16.1.1 of these Guidelines.

Stress distribution of concentrated axial load in orthotropic deck shall be determined by refined analyses where plates and ribs are explicitly modeled. See also Article 16.2.4 of these Guidelines.

16.2 Stability and Buckling

Global and local stability of long span bridges may be verified by approximate methods or second order analyses. Given the complexity associated with designing long span bridges, care shall be exercised to identify the limitations of approximate methods frequently used for compression elements.

16.2.1 Second Order Analysis

Second order analysis shall address: • Initial perturbation force or displacement (imperfection in fabrication and erection) • Geometric nonlinearities (P-delta effects) • Material nonlinearities (cracked sections and plasticity associated with residual stress in case of steel construction)

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16.2.2 Arch Rib and Towers

In performing second order analysis, construction tolerances shall be defined and used as a basis for initial perturbation force or displacement (imperfection in fabrication and erection).

Analyses shall also account for stress from concentrated loads that may initiate instability.

16.2.3 Cable-Stayed Deck

The stability of typical cable-stayed decks with closely spaced stay cables may be assessed by linear elastic methods. Where linear elastic analyses indicate a F.S.<4 against elastic buckling, the deck shall further be subjected to second order analyses.

Where linear elastic analyses indicate a F.S. of 4, the moment magnification method in AASHTO LRFD may be used.

Decks not supported by closely spaced cables, or with unusual configurations such as a self- anchored suspension bridge, shall be analyzed by the second order analyses.

16.2.4 Orthotropic Girders

See also Article 16.1.2 of these Guidelines for local and global stability issues regarding analysis of orthotropic deck.

For cable-stayed girders or sections subjected to axial loading, global buckling capacity shall take into account: • Initial perturbation force or displacement (imperfection in fabrication and erection) • Geometric nonlinearities (P-delta effects) • Material nonlinearities (partially cracked sections and plasticity associated with residual stress in case of steel construction)

Care shall be exercised to account for the combination of global and local stresses, which are detail specific. As a minimum, the Designer shall consult Articles 4.6.2.6.4, 4.6.3.2.3, 6.14.3, and 9.8.3.4 of AASHTO LRFD.

17.0 DYNAMIC ANALYSIS

17.1 Structural Damping

Equivalent viscous damping can be used to represent energy dissipation for extreme wind and seismic effects.

The following structural damping ratios (percentage of critical damping) are recommended for analysis of wind effects, including aerodynamic stability analyses: • steel bridges – 0.5% • composite steel/concrete bridges – 0.7% • concrete bridges – 0.9%

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For bridges under construction, reduce the above recommended values by 0.2%.

The following structural damping ratios (percentage of critical damping) are recommended for seismic analyses: • steel bridges – 2.0% • composite steel/concrete bridges – 3.5% to 5.0% • concrete bridges including substructures – 5.0%

17.2 Aerodynamic Analysis

Bridges and bridge components are deemed wind-sensitive when the mass damping parameter Sw < 1.0, where Sw = (Sc) (B/L) (B/D) (fB/V) and Sc = mζ/ρD2

m - mass per deck length ζ - structural damping ratio ρ - air density D - deck depth B - deck width edge to edge L - main span length f - lowest torsional frequency V – flutter speed criteria

Flutter analysis shall be carried out based on recommendations from the Designers and the wind consultant.

Responses to turbulent wind shall be estimated theoretically on long span and flexible bridges.

Flutter and buffeting analyses shall be carried out based on accepted theoretical models. These analyses shall be based on:

• A selected design speed, wind profiles and turbulence properties • Wind tunnel test data • Data on bridges with similar cross sections, and/or data extracted from numerical simulations • Bridge geometry defined on the design drawings • Mass distribution • Modal information from dynamic analysis based on 3-D finite element models.

The completed bridge and selected critical construction/rehabilitation stages must be analyzed.

Structural components (cable stays, suspenders, etc.) shall be designed to accommodate dynamic and static wind pressure, buffeting, vortex shedding, wake galloping and rain-wind induced vibrations. Both the completed bridge and the bridge under construction shall be considered.

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17.3 Seismic Analysis

Seismic analysis of long span bridges shall be performed using the multimode elastic method (MM) or the time history method (TH). Where the geometric and structural configuration of the substructure units and mass distribution of the bridge are such that they lend themselves to pushover analysis, the multimode elastic method (MM), used as the primary method of dynamic analysis, shall be supplemented by inelastic static (pushover) analysis. The pushover analysis shall be used to verify that the displacement capacity of the substructure is adequate, i.e. that the substructure has adequate ductility to accommodate the seismic displacements.

Also refer to Section C (Article 12.0 – Earthquake Effects) of these Guidelines for a discussion of issues related to seismic analysis and design, including: • seismic performance objectives • design earthquake hazard levels • design ground motions • combination of seismic force effects

17.3.1 Multimode Response Spectrum Analysis

The multimode elastic method (multimode response spectrum analysis) is suitable for seismic analysis and design of long span bridges that are located in areas of low seismicity, and whose response under seismic loading is expected to be essentially elastic.

Multimode response spectrum analysis shall be performed in accordance with the provisions of AASHTO LRFD Article 4.7.4.3.3 and as described herein.

The total structure response (i.e. member forces, displacements or relative displacements) shall be computed by combining the respective response quantities from the individual modes by the Complete Quadratic Combination (CQC) method. Responses in multiple orthogonal directions shall be combined according to Section C, Article 12.5 of these Guidelines. Ground motion shall be applied at a sufficient number of angles to capture the maximum deformation of all critical components; refer to Method 2, Section 2.1.2 of Caltrans Seismic Design Criteria (D09).

Displacements resulting from a multi-mode response spectrum analysis shall be amplified in accordance with Article 17.3.8 of these Guidelines. These amplified displacements are the displacement demands against which the inelastic static (pushover) analysis results shall be compared.

Response modification factors (R-factors) for members and connections shall be as specified for Critical Bridges in AASHTO LRFD Section 3.10.7 (D02).

17.3.2 Time History Analysis

Time history analysis should be used for the seismic analysis and design of long span bridges in areas of moderate to high seismicity.

Time history analysis can be elastic or inelastic (material nonlinearity). For long span bridges whose response under seismic loading will go into the inelastic range, a full inelastic time history analysis is recommended.

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Time history analysis shall be performed in accordance with the provisions of AASHTO LRFD Article 4.7.4.3.4 – Time-History Method and as described herein.

The structure shall be analyzed for multiple sets of ground motion time histories as specified in Section C, Article 12.3.3 of these Guidelines.

The inelastic time history analysis performed for the Safety Evaluation Earthquake shall include spatial variation time histories (multiple support excitations) to capture the spatial coherence effects of earthquake motions. Spatial variation time histories and multiple support excitation input are not required for the Functional Evaluation Earthquake.

17.3.3 Inelastic Static (Pushover) Analysis

When the global dynamic analysis of long span bridges is done using multimode response spectrum analysis, inelastic static (pushover) analysis shall be used to verify that the displacement capacity of the substructure is adequate, i.e. that the substructure has adequate ductility to accommodate the seismic displacements.

The basis of the performance assessment criteria for the Safety Evaluation Earthquake is the limitation of the structure displacement demand to an acceptable fraction of the structure’s displacement capacity, as follows:

• For operationally critical structures: ∆d ≤ 0.67 ∆c

• For non critical structures: ∆d ≤ 0.95 ∆c

where: ∆d = Displacement (or strain) demand, i.e. the elastic displacement demands determined from the elastic multi-mode spectral analysis, adjusted (increased) in accordance with Article 17.3.8 of these Guidelines.

∆c = Displacement (or strain) capacity, determined from the inelastic static (push-over) analysis. The displacement capacity shall be defined as the generalized, controlling structure displacement which occurs when any non-sacrificial member of the structure reaches its displacement capacity in the pushover analysis, but not to exceed the displacement when the lateral resistance degrades to a minimum of 80% of the peak resistance. The member displacement capacity is considered to be reached when the concrete or steel reaches the specified strain limit, but not to exceed the displacement when the lateral resistance degrades to a minimum of 80% of the peak resistance. The strain limit (ultimate strain capacity) shall be taken as 0.005 for unconfined concrete and 0.09 for main (longitudinal) reinforcing steel and structural steel.

The strain limit (ultimate strain capacity) for confined concrete shall be defined by the constitutive stress-strain model for confined concrete.

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17.3.4 Moment-Curvature Analysis

The plastic moment capacity of ductile concrete members shall be calculated by moment- curvature (M-ф) analysis on the basis of expected material properties.

17.3.5 Hydro-Dynamic Effects

Dynamic (seismic) analysis of long span bridges with submerged structural components (e.g. foundations or drilled shafts) shall consider hydro-dynamic inertia effects, i.e. the dynamic interaction between the submerged foundations (or drilled shafts) and the surrounding water.

17.3.6 Miscellaneous Seismic Modeling Details

Transverse shear keys shall be included in the analytical model only if they are designed to adequately resist the design force(s) for the level of earthquake under consideration (Functional Evaluation Earthquake or Safety Evaluation Earthquake).

17.3.7 Design Displacements

For short-period structures where linear-elastic response models tend to underestimate inelastic displacement amplitudes, horizontal displacements calculated from the elastic response spectrum analysis shall be multiplied by the factor Rd to obtain the design displacements (D05).

Rd = [1 – 1/R ]T*/T + 1/R ≥ 1.0

where: Rd = Amplification factor applied to elastic model spectral displacements to obtain design displacements R = Response modification factor. R shall be taken as the maximum value of R used in design of that frame (i.e. use the maximum value for longitudinal / transverse analysis). T = Fundamental period of vibration of the bridge/structure as a whole, in seconds. T* = Characteristic ground motion period (in seconds) corresponding to the peak of the energy input spectra. Values of T* are given in Table 17.3.7.1, adapted from ATC-32 (D05).

Table 17.3.7.1 Values of Characteristic Ground Motion Period, T*

Note: Mw is the design earthquake moment magnitude defined in Sections 4.2.2 and 4.2.3 (D05).

The amplified design displacements shall be used as the global displacement demand for inelastic static (pushover) analysis.

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REFERENCES

D01 AASHTO Guide Specifications for LRFD Seismic Bridge Design, First Edition, American Association of State Highway and Transportation Officials (AASHTO), 2009 with 2010 Interim Revisions.

D02 AASHTO LRFD Bridge Design Specifications, Fifth Edition, American Association of State Highway and Transportation Officials (AASHTO), with Interim Revisions, 2010.

D03 (AASHTO) Standard Specifications for Highway Bridges, Sixteenth Edition, American Association of State Highway and Transportation Officials, 1996, with Interim Revisions through 1998.

D04 AISC Specification for Structural Steel Buildings, ANSI/AISC 360-05, American Institute of Steel Construction, March 9, 2005.

D05 ATC-32, Improved Seismic Design Criteria for California Bridges: Provisional Recommendations, Applied Technology Council (ATC), 1996.

D06 Brown, D. and Bollmann, H. T., Pile Group Design for Lateral Loading Using COM624, Proceedings of The Design of Bridges for Extreme Events, Atlanta, Georgia, 1996.

D07 Buckle, I.G., Freidland, I.M., Mander, J., Martin, G., Nutt, R. and Power, M., Seismic Retrofitting Manual for Highway Structures, Part I – Bridges, Report MCEER-06- SP10, Multidisciplinary Center for Earthquake Engineering Research and FHWA , December 1, 2006.

D08 Byers and McCabe, “Evaluation of Effective Slab Width for Composite Cable-Stayed Bridges”, TRB Paper #03-4519 delivered at the 82nd Annual TRB Meeting, January 12-16, 2003.

D09 Caltrans Seismic Design Criteria, Version 1.6, Caltrans, November 2010.

D10 Clough, R.W. and Penzien, J., Dynamics of Structures, Second Edition, McGraw-Hill Inc., 1993.

D11 Goyal, A. and Chopra, A.K., Earthquake Analysis and Response of Intake-Outlet Towers, Report No. UCB/EERC-89/04, July 1989.

D12 Liaw, C-Y and Chopra, A.K., Dynamics of Towers Surrounded by Water, Earthquake Engineering and Structural Dynamics, Volume 3, pages 33-49, 1974.

D13 Hannigan, P. J., Goble, G. G., Thendean, G., Likins, G. E. and Raushe, F, Design and Construction of Driven Pile Foundations, Workshop Manual, NHI Course Nos. 13221 and 13222, Publication Report No. FHWA HI-97-013, 1997.

D14 Ho, T., Donikian, R., Ingham, T. Seim, C. and Pan, A., “Seismic Retrofitting Guidelines for Complex Steel Truss Highway Bridges”, Report MCEER-06-SP05, August 1, 2006.

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D15 Lam, I.P., Kapuskar, M., and Chaudhuri, D., Modeling of Pile Footings and Drilled shafts for Seismic Design, Technical Report MCEER-98-0019, MCEER, 1998.

D16 Lam, I.A., Martin G.R. and Imbsen, Roy, Modeling Bridge Foundations for Seismic Design and Retrofitting, submitted to the Third Bridge Engineering Conference at Denver, Colorado, March 10-13, 1991.

D17 Lam, I. P. and Martin, G. R., Seismic Design of Highway Bridge Foundations – Vol. II, Design Procedures and Guidelines, Report No. FHWA-RD-86-102, 1986.

D18 Kutsuina, Kasuga and Morohashi, Non-linear Behavior of the Ibi River Bridge under Ultimate Loads, FIB 2002 Congress.

D19 Mander, J.B, Priestly, M.J.N, and Park, R., “Theoretical Stress-Strain Model for Confined Concrete”, Journal of Structural Engineering, ASCE, Volume 114, No. 8, pages 1804 to 1826, 1988.

D20 Martin, G. R. and Lam, I. P., Seismic Design of Pile Foundations: Structural and Geotechnical Issues, State-of-the-Art (SOA4), Proceedings 3rd International Conference on Recent Advances in Geotechnical Earthquake Engineering and Soil Dynamics, St. Louis, Missouri, Vol. 3, 1995.

D21 MCEER/ATC-49, Recommended LRFD Guidelines for the Seismic Design of Highway Bridges, Applied Technology Council and the Multidisciplinary Center for Earthquake Engineering Research, 2003.

D22 NCHRP Report 543, Effective Slab Width for Composite Steel Bridge Members, 2005.

D23 Park et al, “Evaluation of Cable Loss in Cable Stayed Bridges During Design”.

D24 Parsons Brinckerhoff Quade and Douglas, Inc. and GeoSyntec, Inc., Geotechnical Earthquake Engineering, Reference Manual, Training Course In Geotechnical and Foundation Engineering, NHI Course No. 13239 – Module 9, Publication No. FHWA HI-99-012, 1998.

D25 Parsons Brinckerhoff Quade & Douglas, Inc./Imbsen Consulting Engineer, Seismic Investigation of the Bronx-Whitestone Bridge, Task 7 Report – Development of Bridge Model and Calculation of Vibration Characteristics, February 1996.

D26 Priestly, M.J.N., Seible, F. and Calvi, G.M., Seismic Design and Retrofit of Bridges, John Wiley and Sons, 1996.

D27 Priestley, M.J.N., and F. Seible, “Seismic Assessment and Retrofit of Bridges”, Structural Systems Research Project, Report SSRP-91/03, University of California San Diego, July 1991

D28 Tang, Man-Chung, Buckling of Cable-Stayed Girder Bridges, ASCE Journal of the Structural Division, Vol. 102, No. ST9, September 1976.

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D29 Taylor, P., Kaneko, A., and Bergman, D., Buckling Stability Analysis and Secondary Stress Effects in the Deck Girders of Cable-Stayed Bridges, 1994.

D30 Ren, Wei-Xin, Ultimate Behavior of Long-Span Cable-Stayed Bridges, ASCE Journal of Bridge Engineering, Vol. 4, No. 1, February 1999.

D31 Zoli T. and Woodward R., “Design of Long Span Bridges for Cable Loss”, IABSE Symposium Lisbon 2005, Lisbon, 2005.

D32 Davenport, A. and Larose, G., The Structural Damping of Long Span Bridges: An Interpretation of Observations, Canada-Japan Workshop on Bridge Aerodynamics, Ottawa, Canada, September 25-27, 1989.

D33 ESDU 87035, Calculation Methods for Along-Wind Response Part 1: Response of Buildings and Line-like Structures to Atmospheric Turbulence, December issue of the Engineering Sciences Data Unit, 1987.

D34 Stoyanoff, S. and Irwin, P. Flutter Analysis of Lions’ Gate Bridge during Deck Replacement, 6th Asia-Pacific Conference on Wind Engineering (APCWE VI), Seoul, South Korea, 2005.

D35 Davenport, A.G., The response of Slender Line-Like Structures to a Gusty Wind, Institute Civil Eng. 23, 389-408, 1962.

D36 Irwin, P.A., Wind Tunnel and Analytical Investigations of the Response of Lions’ Gate Bridge to a Turbulent Wind, National Research Council of Canada, NAE Report LTR- LA-210, June 1977.

D37 SCDOT Seismic Design Specifications for Highway Bridges, Version 2.0, South Carolina Department of Transportation, July 2008.

D38 Parsons Brinckerhoff, Inc., Circular No. 3, Reference Manual, NHI Course No. 130094 – LRFD Seismic Analysis and Design of Transportation Geotechnical Features and Structural Foundations, Publication No. FHWA-NHI-11-032, Rev. 1, August 2011.

D39 Tseng, W-S., and Penzien, J., “Soil-Foundation-Structure Interaction”, Chapter 42, Bridge Engineering Handbook – Edited by W-F. Chen and L. Duan, CRC Press, 1999.

D40 Lam, I.P. and Law, H., “Soil Structure Interaction of Bridges for Seismic Analysis”, Technical Report MCEER-00-0008, Multidisciplinary Center for Earthquake Engineering Research, State University of New York at Buffalo, Buffalo, NY, September 25, 2000.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section E

TABLE OF CONTENTS

18.0 GENERAL ...... 1 18.1 Material Properties ...... 1 18.1.1 Creep and Shrinkage ...... 1 18.1.2 Concrete Cover ...... 1 19.0 LIMIT STATES ...... 1 19.1 Service Limit States ...... 1 19.1.1 Stress Limits for Concrete ...... 1 19.2 Strength Limit States ...... 2 20.0 DETAILING REINFORCEMENT ...... 2 20.1 Detailing for Ductility ...... 2 20.1.1 Hollow Columns...... 2

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section E

SECTION E – CONCRETE STRUCTURES

18.0 GENERAL

Unless otherwise stated or supplemented by these Guidelines, the provisions of AASHTO LRFD (E01) Section 5 shall apply for the design of long span bridges.

18.1 Material Properties

18.1.1 Creep and Shrinkage

Design of long span bridges constructed segmentally and/or in stages shall account for time dependent effects associated with creep and shrinkage. Acceptable models include –

• ACI 209 (E02) • CEB-FIP 1990 (E03) • CEB-FIP 1978 (E04) • AASHTO LRFD

The influence of reinforcement shall be considered where appropriate.

18.1.2 Concrete Cover

Concrete cover specified under Article 5.12 of the LRFD specifications shall be evaluated against the anticipated concrete properties and the specified design service life of the structure. Modifications shall be made as required to suit project specific criteria.

19.0 LIMIT STATES

19.1 Service Limit States

Service limit states shall be checked for creep and shrinkage coefficients varying from the mean within a specified percentage of confidence limits.

Concrete sections shall be checked for cracking under service loads, including imposed deformations. The effects of cracked sections on the stiffness of the structural system shall be assessed and, where necessary, appropriately represented in iterative analyses to provide a realistic estimate of the design forces and deformations.

19.1.1 Stress Limits for Concrete

Tensile stress limits in non-replaceable concrete superstructure elements at the service limit state in both the longitudinal and transverse directions -

Joints without minimum bonded auxiliary reinforcement through the joint • Zero (no tension) under permanent loads and transient loads

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Joints with minimum bonded auxiliary reinforcement through the joint • Zero (no tension) under permanent loads • 0.0948√f’c (ksi) under permanent loads and transient loads

19.2 Strength Limit States

Strength limit states shall be checked for mean creep and shrinkage coefficients.

20.0 DETAILING REINFORCEMENT

20.1 Detailing for Ductility

Seismic detailing of foundations and substructures shall meet the requirements of Section 8 of the AASHTO Guide Specifications for LRFD Seismic Bridge Design (E09), which shall take precedence over the seismic detailing requirements in the AASHTO LRFD specifications (E01).

All long span bridges, regardless of location or seismicity, shall be detailed to meet the requirements for Seismic Design Category (SDC) B in the AASHTO Guide Specifications, as a minimum.

20.1.1 Hollow Columns

Detailing of hollow columns shall be consistent with the requirements of Articles 5.7.4.7 and 5.10.12 in the LRFD specifications.

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REFERENCES

E01. AASHTO LRFD Bridge Design Specifications – Fifth Edition, American Association of State Highway and Transportation Officials (AASHTO), with Interim Revisions, 2010.

E02. ACI Committee 209, “Prediction of Creep, Shrinkage, and Temperature Effects in Concrete Structures (ACI 209R-92), “American Concrete Institute, Farmington Hills, Mich., 1992.

E03. CEB-FIP Model Code, Design Code, Thomas Telford, London, 1990.

E04. CEB-FIP, Model Code for Concrete Structures: CEB-FIP International Recommendations, 3rd ed., Comite Euro-International du Beton, Paris, 1978.

E05. BS 5400: Part 4: “Steel, concrete, and composite bridges, Code of practice for design of concrete bridge”, London, 1990.

E06. fib bulletin no. 34: “Model Code for Service Life Design”, fib secretariat, Case Postale 88, CH-1015 Lausanne, Switzerland, 2006.

E07. Mo, Y. L. and I. C. Nien, Seismic performance of hollow high-strength concrete bridge columns, Journal of Bridge Engineering, November / December 2002.

E08. Hines, E.M.; Seible, F.; and Priestley, M.J.N., Seismic Performance of Hollow Rectangular Reinforced Concrete Piers with Highly-Confined Boundary Elements Phase I: Flexural Tests Phase II: Shear Tests, Structural Systems Research Project Report No. SSRP-99/15, Department of Structural Engineering University of California, San Diego, February 2002.

E09. AASHTO Guide Specifications for LRFD Bridge Seismic Design – First Edition, American Association of State Highway and Transportation Officials (AASHTO), 2009 with 2010 Interim Revisions.

E10. Caltrans Seismic Design Criteria, Version 1.6, November 2010.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section F

TABLE OF CONTENTS

21.0 GENERAL...... 1 22.0 SUSPENSION BRIDGE CABLE ...... 1 22.1 Main Cable Wires ...... 1 22.1.1 Limit States ...... 1 22.1.2 Quality Control ...... 2 22.2 Cable Saddles ...... 2 22.2.1 Tower Saddles ...... 2 22.2.3 Anchorage Saddles ...... 2 23.0 SUSPENDERS ...... 2 23.1 Suspender Wires ...... 2 23.1.1 Limit States ...... 2 23.1.2 Quality Control ...... 3 23.2 Suspender Socket ...... 3 23.2.1 Limit States ...... 4 23.2.3 Quality Control ...... 4 23.3 Cable Bands ...... 5 23.3.1 Cable Band Bolts ...... 5 23.3.2 Limit States ...... 6 23.3.3 Quality Control ...... 7 24.0 STAY CABLES ...... 7 24.1 Stay Components...... 7 24.1.1 Wire ...... 7 24.1.2 Strand ...... 7 24.1.3 Stay Anchorage ...... 7 24.1.4 Stay Saddle ...... 7 24.2 Limit States ...... 8 24.2.1 General ...... 8 24.2.2 Service Limit State ...... 8 24.2.3 Fatigue Limit State ...... 8 24.2.4 Strength Limit State ...... 8 24.3 Resistance Factor ...... 8 24.4 Quality Control ...... 9 24.4.1 Wire ...... 9 24.4.2 Strand ...... 10 24.4.3 Stay Anchorage ...... 10 24.4.4 Stay Saddle ...... 10 25.0 STRUCTURAL STEEL ELEMENTS ...... 11 25.1 Compression Elements ...... 11 25.1.1 Slender Plate Buckling ...... 11 25.1.2 Stiffened Plate Buckling ...... 13 25.2 Deformation Induced Stresses ...... 13 25.2.1 Cable Connections ...... 13 25.2.2 Floor System in Arch Spans ...... 14

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section F

SECTION F – STEEL STRUCTURES

21.0 GENERAL

Unless otherwise stated or supplemented by these Guidelines, the provisions of AASHTO LRFD (F01) Section 6 shall apply for the design of long span bridges.

22.0 SUSPENSION BRIDGE CABLE

22.1 Main Cable Wires

Main cable wires shall be parallel zinc coated carbon steel. The zinc coating shall conform to ASTM A586.

The modulus of elasticity of main cable wires shall be assumed as 29,000 ksi for the net section, and 28,500 ksi for the nominal section, respectively.

22.1.1 Limit States

22.1.1.1 General

Load factors shall be as given in Section C of these Guidelines. All limit states shall be considered of equal importance.

22.1.1.2 Service Limit State

Suspension bridge cables shall be designed to remain elastic under all service limit states.

22.1.1.3 Fatigue Limit State

The main cable shall be designed for infinite life using the Fatigue I load combination.

The nominal fatigue resistance of the main cable, (∆F)n, shall be taken as 24 ksi.

22.1.1.4 Strength Limit State

(22.1.1.4-1) (22.1.1.4-2) (22.1.1.4-3)

Where Prc = factored resistance of main cable Pnw = nominal resistance of wire Pnc = nominal resistance of cable N = number of cable wires Aw = cable wire nominal area fy = specified minimum yield strength of the cable wire φ = 0.8 = resistance factor for the main cable.

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22.1.2 Quality Control

Bridge wire shall be defined by the ultimate strength, fu, proportional limit, and yield strength at 0.2% strain.

The nature and frequency of testing shall be specified to ensure that the quality of the wires meets design expectations.

22.2 Cable Saddles

AASHTO LRFD, Section 6 shall apply for the design of the saddle structural components and their connection to the top of the tower and anchorage deviation supports.

22.2.1 Tower Saddles

Tower saddles and their connection to the tower structure shall be designed for the unbalanced cable force at the saddle.

22.2.3 Anchorage Saddles

Anchorage saddles and their connection to the structure shall be designed to account for both fixed and sliding conditions as may be anticipated through their service life.

Anchorage saddle bents shall be designed as free head steel columns subject to the unbalanced saddle forces during construction stages and the final condition.

23.0 SUSPENDERS

The provisions specified herein shall apply to the suspenders for suspension bridges and the suspenders for arch bridges.

Suspenders using stay cable technology and details shall conform to the requirements under Article 24.0, Stay Cables, of these Guidelines.

23.1 Suspender Wires

Wire ropes used for suspenders shall conform to the requirements of ASTM Designation: A 603 with Class A galvanizing for inner wires and Class B galvanizing for outer wires.

Helical wire strands for shall conform to ASTM A 586.

23.1.1 Limit States

23.1.1.1 General

Load factors shall be as given in Section C of these Guidelines.

23.1.1.2 Service Limit State

Suspender ropes shall be designed to remain elastic under all service limit states.

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23.1.1.3 Fatigue Limit State

Suspenders shall be designed for infinite life using the Fatigue I load combination.

The nominal fatigue resistance of suspender ropes, (∆F)n, shall be taken as 16 ksi.

23.1.1.4 Strength Limit State

(23.1.1.4-1)

Where

Prs = factored resistance of suspender rope (breaking load) Pns = nominal resistance of suspender rope φ = 0.50 = resistance factor for suspender ropes

23.1.2 Quality Control

In the absence of project specific requirements, twisted wire ropes shall conform to the requirements of Article 22.4, except where specified otherwise below:

• Modulus of elasticity of the pre-stretched suspender rope shall be 20,000 ksi. • Minimum axial fatigue endurance limit – 16 ksi for galvanized suspender rope.

23.2 Suspender Socket

Sockets shall be either forged or made of cast steel conforming to ASTM A 148, 90-60 or higher, with Supplementary Requirement S9.

Sockets shall be galvanized in accordance with ASTM A 123.

Typical Suspender Socket

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23.2.1 Limit States

23.2.1.1 General

Load factors shall be as given in Section C of these Guidelines.

23.2.1.2 Service Limit State

Suspender sockets shall be designed to remain elastic under all service limit states.

23.2.1.3 Fatigue Limit State

Suspender sockets shall be designed for infinite life using the Fatigue I load combination.

Detailing practice shall conform with Detail Category A as defined under AASHTO LRFD Article 6.6.1.2.3. Threaded socket, if used, shall conform with Detail Category B.

23.2.1.4 Strength Limit State

(23.2.1.4-1)

Suspender Socket Design Forces

Where

Pnsk = nominal resistance of suspender rope socket Prsk = factored resistance of suspender rope socket (breaking load) Nf = radial bearing forces acting on the inner wall of the socket due to the pull-out force (Prsk). F = force due to the pull-out f = coefficient of friction between the socket filler material and the socket wall. φ = 0.9 = resistance factor for suspender sockets

23.2.3 Quality Control

The designer should specify the nature and frequency of testing to ensure that the quality of the sockets meets design expectations.

As a minimum, one socket per casting heat shall be radiographed in accordance with ASTM E 280. One specimen per heat shall be Charpy V-notch impact tested at -29°C (-20°F).

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23.3 Cable Bands

Cable bands shall be made of cast steel.

Cable Band Design Forces

Cable Band Clamping Forces

23.3.1 Cable Band Bolts

Cable band bolts shall conform to ASTM A354 and be designed as bolts in slip-critical connections.

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Typical Cable Band Bolt

Cable Bolt Extensometer Hole Detail

23.3.2 Limit States

23.3.2.1 General

Load factors shall be as given in Section C of these Guidelines. All limit states shall be considered of equal importance.

23.3.2.2 Service Limit State

Cable band components shall be designed to remain elastic under all service limit states.

23.3.2.3 Fatigue Limit State

Cable band components shall be designed for infinite life using the Fatigue I load combination.

The stress range shall be compatible with the suspender rope design.

23.3.2.4 Strength Limit State

(23.3.2.4-1)

Where

Psu = suspender rope factored load

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θ = main cable slope at cable band μ = coefficient of friction between cable band and main cable = 0.30 Nb = number of cable band bolts Tb = cable band bolt tension Φ = 0.4 = bolt tension resistance factor

23.3.3 Quality Control

Fabrication shall be in accordance with AASHTO Construction Specifications.

Field installation shall follow specified procedures.

The cable band bolt tension shall be verified frequently after all dead loads are carried by the main cable, and every 15 – 25 years thereafter. Tension shall be restored to the original design value.

24.0 STAY CABLES

These Guidelines cover materials for stay cable systems that utilize prestressing steel and parallel wire.

Criteria for helical strands or locked-coil strands shall be developed to meet project specific requirements if deemed to be economical alternatives.

24.1 Stay Components

24.1.1 Wire

Wire used in stay cables shall conform to ASTM A421/A421M, Type BA.

24.1.2 Strand

Strand used in stay cables shall conform to ASTM A416/A416M, and shall be weldless, low- relaxation grade.

24.1.3 Stay Anchorage

Materials used in anchorages, and/or performance criteria thereof, shall be specified in the contract special provisions.

Stay cable anchorages shall be designed or specified to preclude failure in the anchorage during construction and the service life of the stay.

24.1.4 Stay Saddle

Materials, saddle details, and transition details shall be fully developed by the Designer.

The saddle shall be designed to preclude failure within the saddle during construction and the service life of the stay, taking into account both primary and secondary forces that affect the long term performance of the cable elements and the corrosion protection system.

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Details shall provide for stay replacement without the need to remove more than one stay, and shall provide for cable loss in accordance with Section C, Article 8.1.3.3.

Saddles shall be subjected to testing in advance of construction by the Designer.

24.2 Limit States

24.2.1 General

Load factors shall be as given in Section C of these Guidelines.

24.2.2 Service Limit State

Stay cables shall be designed to remain elastic under all service limit states.

24.2.3 Fatigue Limit State

Stay cables shall be designed for infinite life using the Fatigue I load combination.

For fatigue stress ranges, refer to the PTI Recommendations (F03), Article 5.3.5.

24.2.4 Strength Limit State

Stay cables shall be designed for Strength Limit States for axial load, and combined axial and bending.

24.3 Resistance Factor

Resistance factors shall be as follows:

Strength A – Axial only φ = 0.65

Strength B – Axial and bending combined φ = Phi Factor from Figure 24.3-1 that varies depending on the total LL + wind load stress divided by MUTS. (F09)

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Figure 24.3-1

Extreme event φ = 0.95

Fatigue I φ = 1.0

24.4 Quality Control

Contract special provisions shall be developed to ensure quality control during manufacturing, fabrication, as well as installation of stay cables.

24.4.1 Wire

In the absence of project specific requirements, the following shall be specified for the design of the stay cables:

• Ultimate tensile stress, min. –

f’s = 240 ksi • Yield stress, min. –

f’y = 0.90 f’s • Design elastic modulus – 29,000 ksi +/- 5% • Reduction of area due to necking at rupture – 30% min. • No failure after three “to and fro” bends through 90-degree around a mandrel with a diameter of 5 wire diameters • Fatigue stress range and static strength –

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o Range of 41 ksi for 6 million loading cycles with an upper stress at no greater than 0.45 f’s

o 0.95 of Minimum Ultimate Tensile Strength (MUTS) after specimen passes the fatigue test

24.4.2 Strand

In the absence of project specific requirements, the following shall be specified for the design of the stay cables:

• Ultimate tensile stress, min. – f’s = 270 ksi • Yield stress, min. – f’y = 0.90 f’s • Design elastic modulus – 28,600 ksi +/- 5% • Ductility – sample(s) shall pass the “One Pin Test” • Fatigue stress range and static strength -

o Range of 31 ksi for 6 million loading cycles with an upper stress at no greater than 0.45 f’s

o 0.95 of MUTS min. after specimen passes the fatigue test

24.4.3 Stay Anchorage

Acceptance of stay anchorage design shall be based on fatigue, strength, and corrosion protection (leak) testing of representative specimens comprised of the proposed components. Results of testing shall demonstrate that failure will not occur in the stay cable anchorage during construction and the service life of the stay.

24.4.4 Stay Saddle

Stay design methods and assumptions shall be verified by testing or by reference to previous records of successful testing.

Acceptance of stay saddle design shall be based on fatigue, strength, and corrosion protection (leak) testing of representative specimens comprised of the proposed components. Results of testing shall demonstrate that failure will not occur in the stay saddle during construction and the service life of the stay.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section F

25.0 STRUCTURAL STEEL ELEMENTS

25.1 Compression Elements

25.1.1 Slender Plate Buckling

The provisions specified herein shall apply to the analysis of unstiffened plate elements incorporated in the design of arch ribs, I-shape and box-shape edge girders or floorbeams, and steel tower sub-panels. AASHTO LRFD Bridge Design Specifications and FHWA-TS-80-205 shall apply for the stiffened plate support elements and any other items not covered in the provisions herein.

25.1.1.1 Combined Axial Compression, Bending and Shear at Strength Limit State

At strength limit state, the section shall satisfy:

(25.1.1.1-1)

Where

o Φnc = resistance factor for bending combined with axial load. o Φnv = resistance factor for shear. o fc= fb+fa factored maximum compression stress in the panel due to bending combined with axial load.

o fb = factored maximum compressive bending stress. o fa = factored compressive axial stress. o fv = factored shear stress in the panel. o Fnc = nominal compression resistance in the case of compression stress acting alone. o Fnv = nominal shear resistance in the case of shear stress acting alone.

25.1.1.1.1 Buckling Coefficient (simply supported panel)

The buckling coefficient, k, shall be taken as:

if f1 is tension

(25.1.1.1.1-1)

k=24 for symmetrical bending –f1=fc

OR

if f1 is compression

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(25.1.1.1.1-2)

k =4 for uniform compression f1=fc Where

f1= fb-fa corresponding stress in opposite flange.

25.1.1.1.2 Nominal Compression Resistance

The nominal compression resistance, Fnc, shall be taken as:

25.1.1.1.3 Nominal Shear Resistance

The nominal shear resistance of the panel is calculated using the equations found in AASHTO LRFD 6.10.9.3.2.

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The value of ks should be taken as 5 unless transverse stiffeners are spaced at intervals less than or equal to 3D. The buckling coefficient for transversely stiffened panels can be calculated using AASHTO equation 6.10.9.3.2-7.

25.1.1.2 Resistance Factors

The resistance factor shall be taken as

Strength Limit State:

Φnc = 0.9 Φnv = 1.0

Extreme Event Limit State:

Φnc = 1.0 Φnv = 1.0

25.1.2 Stiffened Plate Buckling

Stiffened plates shall be designed in accordance with pertinent provisions of the AASHTO LRFD Bridge Design Specifications.

25.2 Deformation Induced Stresses

Due consideration shall be given towards the effects of local deformation when detail designing steel elements.

25.2.1 Cable Connections

Pertinent effects requiring consideration shall include:

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. Vibration of cable due to wind . In- and out-of-plane motion of the connection associated with live loading on the structure . In and out-of-plane motion of the connection associated with other loads, including thermal, seismic, and differential settlement effects . Out-of-plane forces associated with construction tolerances.

25.2.2 Floor System in Arch Spans

The connection between floor beam and tie girder, and/or stringer and floor beam, shall be designed to accommodate the elongation in the tie girder caused by superimposed dead loads, live and other transient loading on the bridge.

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REFERENCES

F01. AASHTO LRFD Bridge Design Specifications – Customary U.S. Units, Fourth Edition, American Association of State Highway and Transportation Officials (AASHTO), 2007.

F02. Mahmoud, K.M., “Degradation of Bridge Cable Wire due to Stress Corrosion Cracking and Hydrogen Embrittlement”, International Mesomechanics Conference, University of Tokyo, Japan, 2003.

F03. PTI, “Recommendations for Stay-Cable Design, Testing and Installation”, 5th Edition.

F04. “Cable Stays – Recommendations of the French International Commission on Prestressing”, SETRA, Bagneux, France, June 2002.

F05. “Recommendations for the Acceptance of Stay Cable Systems, using Prestressing Steels”, FIB Bulletin No. 30, 2005.

F06. FHWA-TS-80-205 (Proposed Design Specifications for Steel Box Girder Bridge, 1980

F07. BS-5400, Code of Practice for Design of Steel Bridges, 1982.

F08. AASHTO Standard Specifications for Highway Bridges, Seventeenth Edition, American Association of State Highway and Transportation Officials (AASHTO), 2002.

F09. PTI, “Recommendations for Stay-Cable Design, Testing and Installation”, Committee DC-45 Approved Ballot Item for the 6th Edition.

F10. Ogawa, A. and A. Kasuga. “Extradosed Bridges in Japan”. FIP Notes No. 2, 1998: 11-15.

F11. Wollmann, G.P., D.L. Yates, J.E. Breen, and M.E. Kreger. “Fretting Fatigue in Post-Tensioned Concrete”, Center for Transportation Research, University of Texas at Austin, Report #465-2F, November 1988.

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TABLE OF CONTENTS

26.0 GENERAL ...... 1 26.1 Subsurface Investigations ...... 1 26.2 Types of Foundations ...... 3 26.2.1 General ...... 3 26.2.2 Foundation Selection ...... 3 26.3 Design ...... 3 26.3.1 General ...... 3 26.3.2 Group Effects ...... 4 26.3.3 Magnification of Displacements ...... 4 26.3.4 Pile Installation ...... 4 26.3.5 Caisson Foundations / Cable Anchorages ...... 4 26.3.6 Scour ...... 5 26.3.7 Vessel Impact ...... 6 26.3.8 Constructability ...... 6 26.4 Foundation Displacements ...... 6 26.5 Abutments ...... 7 26.6 Load Tests and Integrity Testing ...... 7 26.6.1 Driven Piles ...... 7 26.6.2 Drilled Shafts ...... 8 26.6.3 Drilled Shaft Integrity Testing ...... 9 26.7 Cofferdams ...... 9 26.7.1 Design Considerations ...... 9 26.7.2 Design Methodology ...... 10 26.8 Rockfill Islands ...... 10

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section G

SECTION G – FOUNDATIONS

26.0 GENERAL

Subsurface investigations, type selection, design and testing shall follow procedures and provisions used for conventional bridge structures as presented in Section 10 of AASHTO LRFD and as specified herein.

26.1 Subsurface Investigations

The geotechnical investigations shall provide data concerning the ground and the groundwater conditions at and around the proposed foundation location necessary for proper identification of the type of soil and rock materials present, definition of the boundaries of these materials, assessment of the engineering properties to be used in design calculations, and identification of underground features that may impact foundation construction.

For long span bridge projects, consideration shall be given to performing the subsurface investigation program in stages, appropriate to the phasing of project development.

The minimum number and depth of the investigation borings shall be as indicated in Table 26.1- 1. Investigation borings shall extend through the geologic formations which may have an impact on the performance of the foundations and below which the ground will have no substantial influence on foundation behavior.

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Table 26.1-1 Recommended Minimum Number of Exploration Points and Depth of Exploration for Long Span Bridges

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26.2 Types of Foundations

26.2.1 General

Foundations shall be suitable for the applicable structure loading combinations, existing subsurface conditions, and serviceability requirements for the bridge structure.

26.2.2 Foundation Selection

Foundation selection and design shall consider the foundation loads related to the type of bridge being supported, and the suitability of the foundation bearing strata to support these loads.

Foundation selection shall consider at least the following issues:

• Load requirements – under construction, service and extreme load conditions, • Factored foundation (geotechnical) resistance, • Serviceability under service and extreme loads, • Soil conditions that will influence foundation installation and performance, • Impact on hydraulics and scour, • Water depth, wave action, tides, seasonal flooding, etc., • Weather conditions, • Site accessibility, • Environmental impacts, • Impact on existing structures, • Constructability considerations, • Driveability for driven piles, • Local experience and availability of materials and equipment, and • Costs of fabrication and installation.

Foundation selection shall consider the serviceability of the bridge structure during and after completion of the structure.

The selection of the foundations shall consider the construction stage loading conditions.

Spread footing foundations shall not be used at sites where the foundation bearing stratum may be subjected to scour.

Battered drilled shafts shall not be used unless full-length permanent casing is provided.

Waterline, submerged and mudline footing alternatives should be considered, where feasible, for foundations located offshore.

26.3 Design

26.3.1 General

Foundations shall be evaluated for all construction loading cases, and designed for those construction conditions.

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For permanent foundation elements, the resistance factors to be applied for construction loading shall be the same as those specified for strength limit state load combinations.

26.3.2 Group Effects

AASHTO LRFD Specifications (5th Edition), with 2010 Interim Revisions, shall be used for determining group reduction factors for piles and drilled shafts in soil.

For drilled shafts socketed into rock, the group reduction factors shall consider the strength and quality of the rock, and the influence of discontinuities in the rock mass. For drilled shafts deriving the bulk of their resistance from rock, group reduction factors for vertical and lateral loading conditions can be obtained from elastic solutions.

The center-to-center spacing between drilled shafts in rock shall be not less than 2.5 times the diameter of the drilled shafts.

26.3.3 Magnification of Displacements

Foundation evaluation shall consider both displacements and strengths under the extreme limit state. Permanent displacements shall also be evaluated under the extreme limit state. Analyses shall be performed to estimate displacement of the substructure and superstructure, taking into consideration the magnification of displacements related to the height of the bridge structure and from plastic hinging in the columns.

26.3.4 Pile Installation

The Dynamic Formula outlined in the AASHTO LRFD Specifications shall not be used for evaluating pile axial resistance for long span bridges.

Wave Equation Analyses shall be performed for driven pile foundations to evaluate the suitability of the proposed pile driving equipment, to develop preliminary driving acceptance criteria, and to predict pile driving stresses.

26.3.5 Caisson Foundations / Cable Anchorages

The following shall be considered in the design of caissons: • Bearing resistance • Sliding • Overturning or tilt • Vertical and horizontal displacement • Differential settlement • Overall (global) stability

Bearing resistance analyses shall be performed using drained and undrained soil parameters, as appropriate, and the relevant groundwater regime using procedures and resistance factors for spread footings outlined in AASHTO LRFD Specifications, Section 10.6 (5th Edition, 2010). Unless jetting or slurries or other means of facilitating caisson installation are prohibited, side friction shall not be considered in determining resistance to vertical loads.

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The caisson resistance to lateral loads shall consider the strain incompatibility between the sliding friction and passive resistance.

Caissons shall bear on competent soil or rock providing uniform bearing resistance.

Earthquake generated inertial forces on partially embedded caissons shall be accounted for using soil structure interaction techniques.

Anchorages for Suspension Bridges

Caissons or anchorage blocks used to support anchorage loads from suspension bridge cables shall be designed to prevent sliding, overturning and uplift. For anchorage caissons founded on competent rock, the required resistance to the applied load can be provided by permanent rock anchors. For other ground conditions, resistance shall be provided by the dead weight of the caisson and shear resistance along the base of the caisson; passive resistance from the soil or rock adjacent to the caisson shall not be relied upon unless such resistance can be assured to remain for the life of the structure, and the displacement necessary to develop this lateral resistance is considered in the design of the bridge and the anchorage foundation. Rock anchors, if used, shall be provided with corrosion protection measures consistent with the design life of the structure.

26.3.6 Scour

Scour analyses shall be performed on an iterative basis during the course of the design process, starting from conceptual design through final design. As foundation design progresses and foundation locations, types and dimensions are finalized, a final scour analysis shall be performed. These analyses shall also include the impact of scour on any temporary construction.

In the computations for side resistance and end bearing resistance to axial loads, any reduction in overburden stress due to global scour shall be incorporated as shown in Figure 26.3.6-1.

In lateral capacity analyses, any reduction in overburden stress due to local scour shall be incorporated as shown in Figure 26.3.6-1.

Full scour shall be used for foundation analyses under service and strength limit states. Partial scour shall be used for the extreme limit state.

Original Seabed

GLOBAL SCOUR DEPTH

LOCAL SCOUR Original Effective Overburden Stress

Reduced Effective Overburden Stress from Global Scour

Figure 26.3.6-1: Reduction in Overburden stress due to Global Scour

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26.3.7 Vessel Impact

Loads from vessel impact shall be estimated and addressed in the foundation design, or measures shall be provided to protect the pier and pier foundation from vessel impact.

26.3.8 Constructability

Foundation design shall consider at least the following constructability issues:

Spread Footing: Verification of bearing material; base preparation; excavation method; rock excavation; groundwater control and dewatering; support of excavation; mass concrete placement and curing; quality control measures and documentation procedures.

Driven Piles: Hammer availability; driveability; driving stresses in piles; sequence of pile installation in pile groups; vibration induced settlement; soil and pile heave; driving time; use of jetting, predrilling or spudding; environmental considerations including noise and vibrations, and impact on fish; quality control measures and documentation procedures; and pile testing.

Drilled Shafts: Equipment availability; slurry selection (water, bentonite, polymer); use of temporary and/or permanent casings; excavation methods; presence of obstructions; bottom cleaning methods; concrete mix design; concrete placement method; mass concrete issues for large diameter shafts; quality control measures and documentation procedures; integrity testing; and .

Caisson: Site preparation; sinking method; measures for reducing side friction; penetration of hard strata; leveling of sloping rock surface; material handling and disposal; placement of bottom seal concrete; cofferdam for footing and pier construction; quality control measures and documentation procedures.

26.4 Foundation Displacements

Unless otherwise noted, the provisions of the AASHTO LRFD Specifications, 5th Edition (2010), with 2010 Interim Revisions, shall apply for determination of foundation displacements of long span bridges.

The design of foundations shall consider displacements that occur during and after construction, due to all applicable AASHTO loading combinations.

Displacements under extreme event loading shall be evaluated using non-linear structural analyses.

Analyses shall be performed to estimate the magnitude of permanent ground and foundation displacement that may occur, and the effect of these permanent displacements on the serviceability of both the substructure and superstructure.

Foundation and structure displacements shall be estimated for the loading and support conditions corresponding to each stage of construction, and the computed displacements shall remain within the applicable serviceability limits.

Bearings joint widths and set width shall be designed to accommodate deflections associated with foundation movements.

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26.5 Abutments

Unless otherwise noted, the provisions of Section 11 of the AASHTO LRFD Specifications shall apply to the design of abutments for long span bridges.

The design life of the abutment shall be not less than the design life of the other substructure and foundation components of the bridge.

Spread footing foundations shall not be used for abutments at locations where the foundation could lose stability or resistance due to scour.

The Designer shall establish an acceptable limiting displacement for seismic conditions based on the acceptable displacements at the bridge bearings. For walls taller than 50 ft and in areas where the peak ground acceleration is greater than 0.3g, soil-structure interaction analyses shall be considered in the design of these walls.

26.6 Load Tests and Integrity Testing

Load testing shall be performed on foundation elements of long span bridges. The type and number of load tests at each site shall be determined based on site variability, the size of the foundation, redundancy in the number of foundation elements at each foundation unit, available existing load test data, and the judgment of the Designer.

The scope of the foundation testing program shall be consistent with the resistance factors used in the design of the bridge foundations.

The load test program for long span bridge projects shall include at least one load test at each main pier foundation, at each anchor pier, at each additional site. A site shall be defined as a project site, or a portion of it, where the subsurface conditions can be characterized as geologically similar in terms of subsurface stratification, i.e., sequence, thickness, and geologic history of strata, the engineering properties of the strata, and groundwater conditions. Note that a site as defined herein may be only a portion of the area in which the structure (or structures) is located. For projects where conditions are highly variable, a site can be limited to a single pier.

Load tests at each site shall be completed and the results evaluated by the Engineer before installing final foundation elements, unless otherwise authorized by the Engineer.

26.6.1 Driven Piles

Dynamic formulae shall not be relied upon to determine pile driving criteria or to estimate pile resistance for long span bridges.

26.6.1.1

One-dimensional Wave Equation Analysis (WEA) shall be used to predict pile driveability and hammer performance prior to driving of the test piles.

26.6.1.2 Dynamic Testing

Dynamic load tests shall be performed in accordance with the procedures specified in AASHTO LRFD Bridge Construction Specifications (2nd Edition, 2008 Interim) Section 4.4.4.3. Dynamic

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section G

pile testing using the Pile Driving Analyzer (PDA) shall be performed on all test piles and on a selected number of production piles in the foundations for long span bridges.

For piles greater than 48 inches in diameter, instrumentation for dynamic measurement of the pile shall include four strain gages and four accelerometers orthogonally mounted near the top of the pile.

Signal matching analysis shall be performed using the data obtained from the dynamic pile tests to obtain a better estimate of the axial pile resistance, and to better define the dynamic soil properties for the refined wave equation analyses.

26.6.1.3 Static Load Tests

Static load tests for axial compression shall be performed in accordance with the procedures specified in AASHTO LRFD Bridge Construction Specifications (2nd Edition, 2008) Section 4.4.4.2 and in accordance with ASTM D 1143.

Tension load tests shall be considered for foundation support elements that will be subject to uplift loads. Static load tests for axial tension shall be performed in accordance with ASTM D 3689.

26.6.1.4 Force Pulse (Rapid) Load Tests

Force pulse (rapid) load tests shall be performed in accordance with the procedures specified in ASTM D 7383.

26.6.1.5 Quality Control Testing

Additional dynamic pile testing shall be performed during production pile driving operations as part of the construction quality control program. The production stage dynamic testing program shall be used to: • confirm continued proper performance of the pile driving equipment with time • assess axial pile resistance and driving stresses at locations not evaluated during the test pile program • evaluate unanticipated or unusual conditions encountered during driving of production piles • assess suspected damaged piles

26.6.2 Drilled Shafts

The method used for performing load tests on drilled shafts shall be appropriate for the size and nominal resistance of the drilled shaft to be tested, and shall provide data sufficient to confirm the required nominal resistance of the production drilled shafts.

Unless otherwise authorized by the Engineer, test shafts shall be installed to the same dimensions, details and elevations shown on the Plans, and shall be installed using the same equipment and installation procedures proposed for installation of the foundation drilled shafts.

When it is impractical to conduct a test on a full-size drilled shaft due to the load requirements or other factors, consideration can be given to testing a reduced diameter test shaft. If a smaller

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section G

diameter test shaft is used, the test shaft shall have a diameter not less than one-half the diameter of the corresponding production shaft, but not less than 30 inches. In addition, the smaller diameter test shafts shall be fabricated and installed in the same way as the drilled shafts used for the foundation, and the test shaft shall be instrumented in such a manner that the base and shaft resistances can be derived separately from the measurements.

If the equipment or procedures are changed following the completion of load testing, the Contractor shall install additional load test shafts, and conduct additional load tests as directed by the Engineer at no additional cost to the Owner.

26.6.2.1 of Drilled Shafts

Static load tests shall be performed in accordance with the procedures specified in ASTM D 1143.

26.6.2.2 Bi-Directional Load Cell Testing

Bi-directional load cell tests shall be performed in accordance with the procedures specified in Chapter 18 of the Federal Highway Administration Manual on Drilled Shafts: Construction Procedures and LRFD Design Methods (2010) (G8).

26.6.2.3 Force Pulse (Rapid) Load Testing

Force pulse (rapid) load tests shall be performed in accordance with the procedures specified in ASTM D 7383.

26.6.3 Drilled Shaft Integrity Testing

Integrity testing for drilled shafts shall be performed in accordance with the procedures specified in Chapter 18 of the Federal Highway Administration Manual on Drilled Shafts: Construction Procedures and LRFD Design Methods (2010) (G8).

26.7 Cofferdams

26.7.1 Design Considerations

The design of cofferdams shall include the following primary considerations:

• The cofferdam structure shall be designed to withstand the various loads applied on it, including lateral water and earth pressures; wave, current and ice loading; loads from floating debris; loads from construction equipment; and accidental vessel impact loading during construction. In addition, in highly seismic areas, earthquake induced hydro-dynamic loads may also need to be considered. • Sizing of the cofferdam sheeting, including sizing of the sheeting section, determination of bracing levels, and determination of the minimum penetration of the sheeting below the final excavation level. • Design of cofferdam bracing, including wales, struts, secondary support bracing, and their connections. • Sizing the thickness of the tremie seal, and analysis of the uplift loads on foundation piles and drilled shafts if used to resist buoyancy load.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section G

• If a tremie seal is not used, verify bottom stability and estimate the quantity of water entering the cofferdam that must be discharged by pumping. Stability analyses shall also consider the potential for bottom heave due to high water pressures in granular soil layers below the bottom of the excavation. • In areas of sloping ground, verify the cofferdam stability under the unbalanced earth pressures. • Design for extreme conditions of wave loads, stream flow and anticipated scour depth appropriate for the anticipated construction period. • Verify that the cofferdam is adequately sized and bracing levels are established to facilitate construction within the cofferdam, including installation of piles or drilled shafts. • Evaluate drivability of sheeting, or identify measures to achieve required sheeting penetration. • Verify that there is no interference between the cofferdam sheeting and new or existing foundation piles. • Determine the sequencing of cofferdam construction, excavation, pile or drilled shaft installation, installation of the tremie seal, dewatering and substructure construction. • Plan the removal of the temporary cofferdams, if necessary.

26.7.2 Design Methodology

The design of cofferdams shall be in accordance with the requirements in AASHTO LRFD Specifications Section 11.8 (5th Edition, 2010) and the FHWA Manual on Earth Retaining Structures (G14). Load combinations and load factors for strength, serviceability and extreme event limit states shall be in accordance with Section 3 of the AASHTO LRFD Specifications.

The cofferdam sheeting and bracing elements shall be designed for strength limit state. In cases where structure deflection must be controlled to protect adjacent structures or facilities, the cofferdam system shall also be designed for serviceability using appropriate criteria.

Cofferdams constructed on land shall be designed for external surcharge load. The loading shall be based on the construction equipment and material surcharge loads anticipated to be applied near the cofferdam, but shall not be less than 500 psf.

Concrete Tremie Seal:

The bond between seal concrete and foundation piles shall be limited to 10 psi, maximum. The bond between the seal concrete and the sheet piling shall be neglected unless positive shear elements are attached to the sheet piles.

For purposes of uplift calculations, the specific gravity for sea water and fresh water shall be taken as 64 pcf and 62.4 pcf, respectively. For inland saltwater lakes, values may be higher; in such cases, site specific values shall be obtained and used in design.

Load and resistance factors for sizing the thickness of the concrete tremie seal shall be in accordance with the requirements in AASHTO LRFD Specifications.

26.8 Rockfill Islands

Where permitted by regulating agencies, rockfill islands may be considered for the main piers of long span bridges to protect the piers and pier foundation from damage due to vessel impact.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section G

Issues to address in the design of rockfill islands shall include:

• Dimensions of the island for protection of bridge piers and pier footings from direct impact from vessel collision. • Vessel energy absorption (based on island geometry and top elevation). • Vessel impact forces on the pier and pier foundation, though considerably reduced, must still be addressed in design. • Island layout to avoid encroachment in the navigable channel. • Impact on waterway hydraulics and related scour and flooding conditions. • Dredging and foundation preparation, including possible ground improvement. • Settlement of underlying soils under the weight of the rockfill island. • Settlement of existing nearby structures. • Downdrag loads on pier foundations. • Specification of rockfill and armoring material, and identification of potential material sources. • Static and seismic slope stability of the island, and interaction between the slope and the pier foundation. • Influence of bridge foundation loads on the stability of the rockfill island. • Protection of the island from erosion due to waves and currents. • Coordination of island construction with foundation construction. • Interference of construction with possible future parallel bridge structures. • Impact on overall construction schedule.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Section G

REFERENCES

G1. AASHTO Guide Specifications and Commentary for Vessel Collision Design of Highway Bridges, Final Report, February 1991.

G2. ASTM D1143-07, “Test Method for Piles under Static Axial Compressive Load”, Book of Standards Volume 04.09, ASTM International.

G3. ASTM D689-07, “Test Method for Individual Piles under Static Axial Tensile Load”, Book of Standards Volume 04.09, ASTM International.

G4. ASTM D4945-96, “Test Method for High Strain Dynamic Testing of Piles”, Book of Standards Volume 04.09, ASTM International.

G5. ASTM D5882-95, “Test Method for Low Strain Integrity Testing of Piles”, Book of Standards, Volume 04.09, ASTM International.

G6. ASTM D6760-02, “Test Method for Integrity Testing of Concrete Deep Foundations by Ultrasonic Crosshole Testing,” Volume 04.09, ASTM International.

G7. ASTM D7383-10, “Test Method for Axial Compressive Force Pulse (Rapid) Testing of Deep Foundations,” Volume 04.09, ASTM International.

G8. Brown, D. A., Turner, J. P., and Castelli, R. J., (2010), Drilled Shafts: Construction Procedures and LRFD Design Methods, U.S. Department of Transportation Federal Highway Administration report No. FHWA-NHI-1—016 (May 2010).

G9. Carter, John, and Kulhawy, F., (1992), “Analyses of Laterally Loaded Shafts in Rock,” Journal of Geotechnical Engineering, ASCE, Vol. 118, No. 6, June 1992.

G10. Hannigan, P.J., Goble, G.G., Likens, G.E., and Rausche, F., “Design and Construction of Driven Pile Foundations”, Volumes I and II, for the National Highway Institute, Report No. FHWA-NHI-05-042, April 2006.

G11. Imbsen, R., (2006), AASHTO Guide Specifications for LRFD Seismic Bridge Design.

G12. Olson R.E., and Flaate, K.S. (1967), “Pile Driving Formulas for Friction Piles in ,” Journal of the and Foundations Division, Proceedings of the American Society of Civil Engineers, Vol. 93, No.6, November 1967.

G13. Paikowsky, S.G. (2004), Load and Resistance Factor Design (LRFD) for Deep Foundations,” NCHRP Report 507, Transportation Research Board, Washington, D.C.

G14. Tanyu, B.F., Sabatini, P. J., and Berg, R. R. (2007), Earth Retaining Structures, Federal Highway Administration report No. FHWA-NHI-071, December 2007.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary A

SECTION A COMMENTARY – INTRODUCTION

C 1.1

This document is intended to be a guide for design and design management purposes for major arch and cable-supported bridges that are generally flexible, long span and may not be adequately covered by the provisions of the AASHTO LRFD. For this reason, arch and cable- supported bridges are referred to synonymously as “long span bridge” in this document. The users are presumed to be familiar with the provisions of the AASHTO LRFD Bridge Design Specifications and the state-of-practice in designing short and medium span bridges.

Design topics in this document are presented in the same general sequence as the AASHTO LRFD Bridge Design Specifications. Figure 1 illustrates the alphabetic designations used herein vs. the numeric designations of the AASHTO LRFD Specifications.

These Guidelines do not specifically address the variety of cable-supported bridges such as the Cable-Stayed-Suspension Hybrid or Self-Anchored Suspension bridges. However, these Guidelines address structure types that exhibit flexibility that is uncommon in short and medium span bridges.

The flexibility and displacement of cable-supported bridges may not meet the more stringent requirements for proper rail operations.

Appendices Practical design considerations for suspension, cable-stayed, and arch bridge types are provided in Appendices A through C of this document, respectively.

Design flow charts, information on bridge health monitoring and soil-structure interaction are provided in Appendices D through F, respectively.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary A

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

SECTION B COMMENTARY – GENERAL DESIGN & LOCATION FEATURES

C2.0

Safe and economical design is generally dependent on the quality of site specific data. Historical data may be obtained through research of archives kept by local agencies. Data may also be acquired through a collection or exploration program.

When historical data are not readily available for the specific site, methods of extrapolation from existing databases are available to generate loads and effects of acceptable reliability. Wind and earthquake effects are two areas where this probabilistic approach has been accepted as the state of practice by many bridge owners.

The design process would often start with an investigation into and the collection of the data that are readily available. A conceptual study and evaluation of the alternatives will then be conducted based on the initial data collection. As design progresses into the preliminary and final stages, further data collection and studies are usually necessary for optimization purposes. The type of special studies needed would depend on the apparent preferred alternative(s). Nevertheless, arch and cable-supported bridges usually require aerodynamic assessment, constructability evaluation, and more elaborate geotechnical investigations. Given the greater investment associated with constructing complex bridges, corrosion protection of a post- tensioned deck or any non-replaceable structural elements, for example, would be another aspect warranting earlier considerations. The design process is iterative and tends to be longer in duration than required for the design of conventional bridges. Figure 2 (Design Flowchart) and accompanying discussions in Appendix D provide a summary of the relationship between main design activities.

C2.1

Commentary C3.4 of the AASHTO Guide Specifications and Commentary for Vessel Collision Design of Highway Bridges presents important sources of data regarding vessel traffic.

Data compilation, sorting and interviews of harbor pilots and operators could uncover critical information that may not otherwise be immediately apparent to the Designer. As the outcome of these studies may indicate a need for collision mitigation measures that could alter the span arrangement of the bridge, it is important that the studies be implemented in step with the overall design program.

C2.2

Due to the significant size of footings for long span bridges, stream flow and scour potential are important considerations. See Section G, Article 26.2 of these Guidelines for discussions on the selection of pile cap features.

C2.3

Section G, Commentary C26.1 of these Guidelines provides detailed discussions regarding the collection and application of subsurface data during the various stages of design development.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

C2.4

Seismic provisions of the AASHTO LRFD Bridge Design Specifications (B2) are considered to be “force-based”, wherein a bridge is designed to have adequate strength (capacity) to resist earthquake forces (demands). In recent years, there has been a trend away from “force-based” procedures to those that are “displacement-based”, wherein a bridge is designed to have adequate displacement capacity to accommodate earthquake demands. Displacement-based procedures are believed to more reliably identify the limit states that cause damage leading to collapse, and in some cases produce more efficient designs against collapse. For bridges designed in accordance with the AASHTO LRFD Bridge Design Specifications, it is recommended that the displacement capacity also be checked using the displacement-based procedures in the AASHTO Guide Specifications for LRFD Seismic Design (B9), particularly for bridges in high seismic zones.

Alternatively, the seismic design of the bridge may be completely based on the displacement- based procedures in the AASHTO Guide Specifications for LRFD Seismic Design (B9). This approach is preferred over force-based design procedures, particularly for bridges in high seismic zones where force-based procedures may result in uneconomical designs.

Generally, site specific studies can provide more accurate characterizations of seismic effects and ground motions. Results of these studies may include – • Probabilistic design ground motions where national hazard maps are not available for the desired design hazard levels. • Deterministic spectra in regions of known active faults. • Site-specific dynamic soil response based on results of geotechnical investigations. • Site class definitions and depth of motion determination based on in-situ shear-wave velocity measurements. • Site coefficients and adjustments to spectral response acceleration parameters. • Soil site effect, wave traveling effect, and near-field effect for spatially varying ground motion time histories.

C2.5

Wind climate studies and site analysis should commence early in the design program to address wind effects that often govern the foundation and substructure design of long span bridges.

Wind climate studies are typically based on historical data (records) of up to 50 years available from local meteorological stations. Statistical analysis can accurately predict wind speeds at the observation station location, and thereafter these speeds need to be scaled to the bridge site. The derivation of scaling factors is typically straightforward when the separation distances are not too great (order of 10-20 miles) and relatively open span the two sites. Sites that are further apart and in complex topography may introduce uncertainties in the final speed predictions, which require a more in-depth analysis.

The site analysis aims at derivation of mean wind profiles and scaling factors at the referenced stations as well as at the bridge site. Turbulence properties are also estimated depending on the local terrains.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

On-site wind speed measurements serve to increase the level of confidence in the derivation of wind speed scaling factors. As a minimum, 1-minute mean wind speeds and directions should be recorded. When deemed essential to a specific design, more reliable wind load estimates may be achieved by also measuring the on-site turbulence properties. Ideally, local data should be recorded, as a minimum duration, during the known seasons of higher winds, e.g., autumn and winter. To ensure complete statistical predictions, a full year of observations would be desirable. When local measurements are unavailable due to project and/or time constraints, alternative methods will have to be adopted.

Topographical model studies using weather research and forecasting (WRF) numerical models could provide scaling factors at the site as well as important information about turbulence properties such as intensities of turbulence and scales, and mean angle of wind attack. These methods could be used to increase the quality of the predictions obtained from the on-site wind speed measurements or as their alternative.

The local terrain studies should take into account the effect of topography and any local large buildings and/or bridges. Project specific criteria must be developed where site conditions cause wind direction to deviate from the horizontal.

While the historic records (up to 50 years of data) acquired from local meteorological stations through the wind climate studies are considered high quality data, this length of time may still be insufficient for establishing the design speeds at sites of extreme winds such as hurricanes. Wind speed predictions for very long return periods are required, e.g., 100, 1000 or more years. Therefore, statistical extrapolations based on short records alone may not be sufficient. In this case, additional numerical simulations that would extend the time length of data are becoming the state-of-practice.

Section B, Article 3.1.1 and Section C, Article 11.0 of these Guidelines present more comprehensive coverage of wind data.

C2.6

See Section C, Article 9.0 of these Guidelines for provisions addressing site specific live load.

C2.7

Provisions regarding durability design can also be found in Section B, Article 3.1 of these Guidelines.

C2.8

Assessing risk scenarios, enhancing stand-off distance, and adopting structural systems that are resistant to progressive collapse are means of mitigating between risk and cost. Redundancy, robustness against extreme events, and cable loss load cases are some of the requirements of these Guidelines that would provide a basic level of safeguarding against progressive collapse.

Security considerations may influence the decisions on type, size, and location for the bridge, which in turn may affect construction cost. Owners should decide the extent to which a bridge is designed to sustain blast loading or other acts of deliberate attack.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

C3.0

Design life is defined, usually by the owner, as the period of time during which the bridge is intended to remain in service. See Article 3.2 for current design practice for extending design life.

C3.1

The design of long span bridges need not be limited by past practice for deflection control.

The design of long span bridges generally requires consideration of non-linear effects and a rigorous assessment of its elements for structural stability. Section D of these Guidelines provides guidance regarding the state-of-practice on structural analyses.

C3.1.1

Critical wind speed is the wind speed at which a particular destabilizing effect is likely to occur for a given structure.

Design wind speeds for long span bridges should be project specific. The method to determine design wind speed can be found in Section C, Article 11.2 of these Guidelines.

Vortex Shedding

Vortex shedding originates from alternating forces on bridge elements caused by regular vortices shed from each side of a given cross section. The shedding frequency of these vortices can be expressed as ns=V/DSt, where D is a characteristic dimension, such as the depth of a deck or a tower leg; V is the mean wind speed; and St is the non-dimensional constant called the Strouhal number which is constant for a given cross-section (e.g., for a circular cylinder 0.2 to 0.25, for a square section 0.12, and from 0.1 to 0.2 for typical bridge sections). On slender deck sections and especially on multiple box sections, St may be outside of these typical ranges. Multiple St values attributed to different excitation mechanisms may appear.

When the frequency of these forces is close to one of the structural natural frequencies, oscillations may build up, provided that the attributed mass m and structural damping ζ are low. The predisposition to vortex-induced oscillations is often evaluated with the Scruton number Sc = mζ/ρD2, where ρ is the air density. For example, the Scruton number for stay cables without any dampers is typically Sc<1, where it is known that Sc>2.5 is required to restrict vortex shedding. For bridge decks, this parameter varies in the range of 0.2 to 4. On bridge decks the product DB (=depth x width) is used instead of D2.

Once vortex-induced oscillations have been established at a given wind speed, the vortex shedding frequency may “lock” into the bridge natural frequency as long as the wind speed does not vary significantly. Therefore, vortex shedding often persists over a narrow range of wind speeds. Vortex shedding is most noticeable in smooth flows and tends to decrease as the turbulence of the wind increases.

Unlike flutter, vortex-shedding oscillations tend to be self-limiting in amplitude. Therefore, they are not of major concern provided the amplitude of the oscillations is not excessive. However, on lightweight decks with low damping, the oscillations may reach sufficient amplitude to cause

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

discomfort or alarm for bridge users, or result in fatigue issues over time. If these cannot be fully eliminated, motions due to vortex shedding must be limited to acceptable levels.

Flutter

Flutter is a self-excited aerodynamic instability, which may involve torsional motion only (torsional flutter), or coupled torsional and vertical motion (classic coupled flutter). It originates from forces caused by the relative motion between the bridge deck and the approaching wind. These motion-induced forces are sensitive to the deck cross-section. These forces may result in the wind either suppressing the bridge motion, in the stable case, or promoting it, in the unstable case. Generally, bridge decks that are narrow (i.e. where width/depth ratio is small) will tend to experience pure torsional flutter. Decks that are wide (i.e. streamlined cross- sections) will tend to experience classic flutter. Coupled motions tend to occur only when the cross-section of the structure exhibits similar mode shapes for both the torsional and vertical modes. On very long-span bridges, complex modal coupling may occur resulting in an unusual mode shape at the state of flutter. This problem must be addressed by performing advanced flutter analysis including all the modes of concern in the solution. The critical wind speed for flutter appears to be lower for decks with low torsional stiffness. Flutter should be avoided at all costs.

Galloping

This is a quasi-static type of instability that is found primarily on narrow bridge decks. Due to a negative rate of change in lift, the section may start to move vertically across-the-flow to very large amplitudes. The sectional test procedure for flutter, which takes vertical and torsional motions into account, can identify whether this type of instability exists. As in the case of flutter, galloping should be avoided at all costs.

Galloping tends to occur on narrower bridges. Typically, sections with slenderness ratios (width/depth) greater than 5 are less likely to gallop.

Stability Solutions

When flutter or galloping occurs, the oscillations can diverge to levels causing failure of the bridge. The more that the wind speed exceeds the critical value, the more quickly oscillations will grow to a level where the bridge would fail. Thus, the design flutter speed should always be much higher than the wind speeds expected at the site. In current design practice, where the 100-year return period wind is appropriate for structural strength design, the design speed for flutter is typically set at a return period in the order of 10,000 years.

The onset of instability speed tends to be linearly proportional to structural frequency and width of the bridge cross-section. Stiffer bridges with wider decks generally exhibit higher critical speeds.

The standard section model tests are typically adequate to estimate the critical wind speeds for flutter and galloping. If any unacceptable instability is found, modifications to the cross-section should be made until acceptable stability is attained.

Details at the edge of the deck can be very important for stability. Reductions to the cross- section bluffness (streamlining) generally improve performance against all types of instabilities. Typical measures include reduction of edge girder and traffic barrier heights; adding edge

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

fairings and soffits; corner cuts for towers, etc. Baffle plates on bridge decks are very effective in reducing vortex shedding. Sectional model testing has been the traditional tool for stability verifications. It is advisable for bridge designers to work closely with the wind consultant to develop a number of acceptable solutions which would then be evaluated during wind tunnel testing. Computational fluid dynamics tools can be implemented for preliminary evaluations.

Vortex shedding instabilities may be tolerated provided the comfort and fatigue guidelines of Section B, Articles 3.1.2 and 3.1.3 of these Guidelines are met.

Interpretation of Stability Speed Criteria

Permanent on-site factors such as unusual terrain and/or large buildings and bridges should be considered as indicated under Article 2.5.

In open terrain where there is an absence of permanent on-site factors that may affect the expected mean angle of wind attack on cross sections, reductions have been applied to the stability criteria on past projects (B4). The reductions applied on various angles of attack were: • 0.0 degree – no reduction; • ± 2.5 degrees – 20% reduction; • ± 5.0 degrees – 50% reduction,

where an angle of attack of 0 degrees signifies a wind with no mean horizontal inclinations over time intervals longer than 1 minute.

The above reduction factors were supported by full scale observations where it was found that strong winds on long span bridges in open terrains tend to be nearly horizontal. Where deviations from the horizontal do occur, the deviations tend to be localized, and/or too transient, to cause destabilizing effects.

Interference Effects

The possibility that bridges may vibrate due to the proximity of other structures should be identified as part of the local terrain studies indicated in Article 2.1. When such effects have been identified, detailed studies should be planned and carried out by wind engineering consultants.

Cable Vibrations

On long span bridges, the stay cables are typically the most flexible elements. Due to their low mass and damping, cable vibrations have been frequently observed. Although not as threatening as flutter and galloping, these vibrations should be controlled to avoid fatigue failure in the cable system. Information on cable vibrations can be found in the PTI Guidelines (B3). It is recommended that bridge designers address this issue during the design phase by working closely with wind engineering consultants.

C3.1.2

The acceleration criteria are based on human perception and shall be applied to user comfort evaluations.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

For wind induced vibrations, the proposed criteria apply for mean wind speeds < 50 mph at deck level.

Since ASCE criteria for vertical accelerations are not practical for deflection calculations on very 0.5 long span bridges at low frequencies, the criteria amax ≤ 0.051 f (%g), following the references BS5400 (B7) and ENV1995-2 (B8) may be applied.

Visual criteria for bridge users’ comfort is another topic currently under investigation that should be considered for long span bridges, where travelers may discern movement while traversing the spans. (B10)

C3.2

The design life of long span bridges is generally achieved by specifying durable components, detailing for easy inspection access, having regular inspection and maintenance performed, and replacing components that have reached their service lives.

Service and predicted service lives are, respectively, the actual and predicted periods of time during which a component of the bridge would perform its design function as long as maintenance and repairs are performed according to plan.

C3.2.1

Concrete Properties

The properties and quality of the concrete are important for concrete durability and protection of reinforcement from corrosion. In addition to provisions in AASHTO LRFD Article 5.12, a combination of physical barriers and mix design that limit permeability and cracking may be considered to ensure longevity of the elements. Concrete decks of several recent long span designs have called for a protective wearing surface and a program that would continually monitor the presence of corrosive agents.

As part of the concrete mix design, a durability study is recommended for non-replaceable concrete components such as towers, piers, and concrete deck for cable-stayed bridges. Factors to consider may include - • concrete permeability • chloride loading • chloride rates Such a concrete durability study may result in recommendations for - • concrete design mixes • concrete covers • possible use of stainless steel, galvanized or coated rebar

Environmental Control

Proper ventilation of interior spaces of large chambers within the bridge, including towers, arch ribs and cable anchorages of suspension bridges, should be considered as one means of corrosion protection. Ventilation may be adequate in some cases by simply creating a natural draft or chimney effect created by providing openings at the top of towers or arch ribs.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

Some recent designers of major bridges have found it economically feasible to provide mechanical / electrical systems for the control of corrosion. A typical example would be the protection of suspension bridge cables by placing them inside air-tight plastic sheathings, and having the interior of the sheathings filled with inert gas.

Limiting Stresses in Concrete

In lieu of or in combination with special coated reinforcement, tensile stress in non-replaceable concrete components may be limited or eliminated by post-tensioning, as may be required to meet design life requirements. See additional provisions in Section E, Article 19.1.1 of these Guidelines.

C3.2.2

Owners and Designers should establish the objectives of the instrumentation project based on the specific needs of the bridge. Some common objectives include:

• Verification of design assumptions used for wind, seismic, vessel impact, storm surges, and other special events • Collection of data to monitor longer-term and non-extreme events such as temperature gradient, and creep and shrinkage of concrete • Collection of data to aid in assessing structural deterioration and degradation of structural elements • Verification of design assumptions for truck live loads.

Some of the recent sensory and data acquisition systems in use on major bridges include – • Anemometers • Accelerometers • Temperature Sensors • Strain Gauges • Global Positioning Systems • Displacement Transducers • Buffer Sensors • Bearing Sensors • Tensile Magnetic Gauges • Barometers, Rainfall Gauges & Hygrometers • Corrosion Cells • Digital Video Cameras • Weigh-in-Motion Sensors

Refer to Appendix E for additional information on health monitoring of long span bridges.

C3.2.3

Bridge components for long span bridges are often proprietary, and it is common practice for designers in the US to prepare performance specifications for specialty products. Nevertheless, it is worth repeating as the engineering required to satisfy the performance requirements, if not already done during design, is complex and relatively costly for long span bridges. Performance

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

requirements in the design documents will allow consideration of alternative products that may be available in the future when replacement is required.

Replaceable bridge components may include – • Stay Cables and their Anchorages • Suspenders • Bearings • Lock-Up Devices • Tie-Downs • Expansion Joints • Wearing Course • Paint and Sealers • Cathodic Protection

C3.2.4

Generally, appropriately sized access openings and interior lighting or a power source should be provided.

The design should provide sufficient points of access and minimum opening sizes for inspection and maintenance.

C3.2.4.1

Tracing the stress history in long span bridges, especially those with cable-supported superstructures, can be complex and crucial for future maintenance of the structure. Good record keeping during new construction as well as subsequent reconstruction is vital to long term maintenance of these structures.

It is recommended that technical reports be prepared by the Designer, the Construction Engineer and the Construction Inspector to highlight essential aspects of the project and to document reportable issues encountered during the performance of their respective duties.

C3.2.4.2

A maintenance manual should be seen as an integral part of the design, as regular inspection and maintenance is required to achieve the recommended minimum design life of 100 years.

C4.0

An example of providing internal redundancy in a tie girder would be to detail the tie girder as a built up member comprised of plates and angles that are bolted together. The failure of an individual plate or an angle would not translate into the total failure of the tie girder. Other means for providing internal redundancy include providing post-tensioning tendons to resist the tension in the tie, or making the deck slab composite with the tie member in order to prevent abrupt failure.

An example of external redundancy would be a composite, post-tensioned deck system where the stringers and deck participate in the tie girder’s load path.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary B

C5.0

This would imply that the Designer is expected to perform at least a limited amount of stage-by- stage analysis.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

SECTION C COMMENTARY – LOADS & LOAD FACTORS

C8.0

These provisions assume that long span bridges are designed to meet or exceed conventional standards for ductility and redundancy according to the AASHTO LRFD Design Specifications (C01). The higher value for ηI reflects an expectation that long span bridges are uniformly considered as important bridges. Designers may elect to further adjust the values for individual bridge components.

C8.1.1

The Strength IV Combination, with a γp factor of 1.5 on component dead load (DC), will govern the design of substructures and foundations without necessarily increasing the reliability. A recent study (C02) suggests that the inclusion of the live load effects would have the desirable effect of maintaining the level of reliability.

At least one US jurisdiction has adopted a policy of applying a γp factor of 1.5 for the superstructure DC only and a reduced γp factor of 1.25 for substructure design. In terms of long span bridges, this policy is expected to impose a more stringent requirement on the design of the substructure for bridges that have uneven span arrangements.

C8.1.2

This provision has been adopted from the PTI “Recommendations for Stay Cable Design, Testing, and Installation” (C03).

C8.1.3.1

Scour depths for the various extreme load combinations should be project specific. Scour conditions should be evaluated based on storm frequencies and differentiated between long term degradation and transient scour (local and contraction). See Article 10.2 of these Guidelines for further discussions on scour. Accepted practice for considering scour conditions may be generalized as follows:

where: ave (LL) = LL per AASHTO LRFD ave (LL) = 0 live load for low volume ave (LL) = load factor for live load, γEQ, taken as 0.5 for high volume road

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

ave (CV) = drifting barge (vessel not under power) ext (CV) = CV per AASHTO LRFD Specifications ext (WX) = extreme wind loading on structure – see suggested return periods under Article C 11.5 WA = WA per AASHTO LRFD Specifications EQ = EQ per AASHTO LRFD Specifications

C8.1.3.2

Loss of an element within a component, or failure of a component without resulting in the collapse of the structure, refers to situations such as –

• One of the flanges or web plates in a tie girder of a tied arch bridge that would contain the spread of structural damage by means of special details, such as bolting

• A floorbeam in a deck system that has been designed composite with the deck slab and that has adequate capacity for an extreme load combination

The equation for loss of an element reflects the state of practice from recent projects, including the Lake Champlain Arch Bridge.

In the case of the tie-girder, designers often use an elastic analysis / inelastic section approach. In this approach, the acceptable strain when determining the section capacities will be specified, say at 1%. Then no advantage of the local inelastic rotation would be taken which would lead to moment redistribution in the structure, i.e. the section is designed for the forces obtained from a linear-elastic analysis.

C8.1.3.3

The provision for stay cable loss has been adopted from the PTI “Recommendations for Stay Cable Design, Testing, and Installation”. If time history analysis is used, “1.1 x (Cable Loss or Suspender Dynamic Forces)” may be simulated by increasing the force being discharged by 10% of the theoretical value.

C8.1.4

The possibility of partial live load with earthquakes, i.e., γEQ < 1.0, should be considered. LRFD Article C3.4.1 notes that application of Turkstra’s rule for combining uncorrelated loads indicates that γEQ = 0.50 is reasonable for a wide range of values of average daily truck traffic (ADTT). This was based on evaluations of “standard” bridges, and the use of a load factor γEQ = 0.50 would likely be conservative for most long span bridges.

Hida (C04) notes: “Vehicular traffic can be significant during seismic events, but the likelihood of fully loaded side-by-side occurrences is clearly remote; design for a fully-factored live load in addition to maximum earthquake would be inappropriate. Design for live load in combination with a more frequent seismic event is appropriate where the likelihood exceeds that of the maximum seismic event designed for – without live load. For example, if a 500-year event is designed for, i.e., the likelihood of 0.14, then a 150-year event should be checked with live load and a reduced live load factor since the likelihood is 0.17 to 0.78.”

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

Most long span bridges have high dead-to-live load ratios. For most of the country (outside of portions of the west coast), design for the upper level Safety Evaluation Earthquake (~2500 year return period) will control over design for the lower level Functional Evaluation earthquake (~500 year return period) at their respective performance criteria. Therefore, for most long span bridges, it is reasonable not to consider any, or full, live load applied simultaneously with seismic loads.

C8.1.5

The AASHTO LRFD Specifications (C01) as well as the AASHTO Guide Specifications for LRFD Seismic Bridge Design (C35) call for earthquake effects to be analyzed in two horizontal directions only. The addition of the vertical ground motions is deemed appropriate for long span bridges because the response of long span bridges to vertical excitations tends to be more sensitive than regular bridges, particularly for structures where P-delta effects may be significant.

The SRSS combination rule is preferred over the 30% rule. Section 7.4.2 of the FHWA Seismic Retrofitting Manual for Highway Structures (C36) recognizes the SRSS combination rule and a 40% rule, but notes that the SRSS method “is especially recommended if the vertical components of the ground motion are being used in conjunction with the horizontal components”.

Other combination rules have also been suggested. These include the CQC3 rule, proposed by Menun and Kiureghian (C40). The Designer should exercise caution if using a correlation coefficient, γ, that is less than 1.0, as using a low γ factor may result in non-conservative demands.

Also refer to Article 12.3.3 Acceleration and Displacement Time Histories.

If member capacity/demand ratios are checked using corresponding forces and moments at each time step, then it is not necessary (or appropriate) to combine forces/moments in the longitudinal, transverse and longitudinal directions using the SRSS or 30% combination rules.

C8.1.6

LRFD Table 3.4.1-1 shows a 0.5 load factor for TU for the Strength Limit States, and LRFD Table 3.4.1-3 shows a 0.5 load factor for CR and SH for substructures supporting non- segmental superstructures. The 0.5 reduction factor for TU, CR and SH is not considered appropriate for long span bridges because they often have steel substructures that require design and detailing to ensure an appropriate level of ductility. Additionally, TU, CR, and SH effects tend to be more significant for long span design regardless of construction materials, and therefore require more rigorous analyses.

Note that the load factors are applied to the imposed deformations, not the induced load effects.

C8.1.7.1

This modification reflects industry practice.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

C8.1.7.2

This modification is consistent with PTI recommendations (C03).

C8.2.1

Long span bridges are often constructed segmentally. Hence construction loads and sequence of loading should be simulated to account for the stress history of the structural elements. Information regarding the design assumptions will also help the constructor assess the selection of his/her means and method.

C8.2.2

LRFD Articles 5.14.2.3.2 and 5.14.2.3.4 were developed for segmental construction, but contain provisions that may apply to long span bridge construction in general.

C8.2.3

Equation 8.2.3-1 is adapted from AASHTO LRFD. The combinations for stay and suspenders have been derived from the provisions of the PTI Recommendations.

C8.3

The erection of cable-stayed bridges with a composite deck is an example where the steel girders and floor beams may require significant temporary strain during erection, because the majority of the dead load is not yet in place to deflect the steel elements into their final shape. Erection sequence and timing of stitching deck to the steel in a composite section greatly affect the locked-in stresses.

C8.4

The effects of adjusting cable stay and suspender forces are similar to the secondary effects due to post-tensioning and are factored in a similar fashion.

C8.5

The minimum impact factor of 1.5 is based on the PTI Recommendations. Studies by Zoli (C05) and others have shown a more rational methodology than the equivalent static method reflected in these Guidelines. It has been noted that dynamic response is sensitive to the unloading duration of the loss event.

C9.0

A review of the HL-93 loading by Nowak, Lutomirska, and Ibrahim (C06) has concluded that the HL-93 live load specified in AASHTO LRFD (2007) can be applied for the design of new long span bridges. Site specific studies should be conducted in cases where a replacement bridge is planned for a site expected to see an inordinate level of heavy trucks. It is expected that existing long span bridges will be evaluated according to site specific studies in order to result in satisfactory ratings.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

C9.1

Based on studies conducted by Nowak, Lutomirska, and Ibrahim (C06).

C9.2

A study of dynamic effects presented in a report by the Calibration Task Group (Nowak 1992) (C07) contains details regarding the relationship between dynamic load allowance and vehicle configuration.

A site specific study may result in an increased loading. In this case, a corresponding reduction of impact from the heavier vehicles may be warranted.

C9.3

Comfort and confidence of pedestrian and parading crowds is an important consideration. Refer to Section B, Article 3.1.2 of these Guidelines.

The parade loading customarily used on smaller structures or pedestrian bridges may be too conservative for long span bridges. A study of the Millennium Bridge (C08) documented a pedestrian density of 1.3 ~ 1.5 persons/m2 or 19.3~22.3 psf. A study of the Verrazano-Narrows Bridge (C09) during the NYC marathon documented a runner density of 0.53 persons/m2 or 7.9 psf.

C9.4

Repair and inspection are often scheduled during non-peak hours. Provisions for maintenance of traffic would often keep the heaviest truck loading at a safe distance from the inspection and maintenance activities. These are reasonable cause to check maintenance and repair load cases with reduced vehicular/traffic loads. In the case where the use of an under-slung traveler is the only reasonable means to provide inspection and maintenance services to the bridge and that the traveler is an integral part of the design, the Designer should consider no reduction in vehicular / traffic loads.

C10.1

Refined numerical modeling for long-span bridges provides a versatile method for prediction of maximum wave heights. Improved modeling of the wave heights will benefit projects over simplified analysis techniques as uncertainty in wave heights could influence acceptable clearance heights and associated bridge superstructure elevations. Tsunami is an extreme event that is under review and expected to become a part of the design standards for coastal structures as sufficient research results become available. Until such time, tsunami effects should be considered on a case by case basis.

Wave loads and wave modeling is appropriate for any bridge structure open to the ocean, or over an inland bay, gulf, deep estuary, or large lake. Wave modeling may be considered for any scenario where depth limited waves are expected to be greater than 1-foot in depth. Pre- screening of bridge structures may involve consultation of local FEMA mapping for documentation of potential wave action.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

See references on analytical tools in the section below.

C10.2

Evaluation of scour potential should be performed by specialists in hydraulic/coastal engineering. Site specific studies, where warranted, should be documented as to approaches and assumptions.

Advanced two-dimensional hydraulic modeling provides flexibility in scour prediction. Two- dimensional modeling provides an increased view of velocity distributions across large systems and identifies areas of locally increased velocity that can be critical in prediction of scour. Two dimensional modeling can be performed using ADCIRC (C28) for coastal bridges and FESWMS (C29), RMA-2 (C30), or TUFLOW (C31) for riverine bridges.

Bedrock erodibility relationships have been proposed by Annandale (C32) that correlate measureable aspects of the rock to its potential erodibility. Further research into this topic is expected to refine the relationships and develop methods for prediction of scour depth reductions.

The use of HEC-18 (C33) equations in the prediction of scour depths for coastal bridges typically results in over-prediction of scour. Recent research suggests that the over-prediction is the result of the small time-scale of erosion experienced during a storm surge in a coastal area. Relationships independently postulated by Sheppard (C34) and FHWA (C32) provide guidance for time-dependent consideration of contraction and local scour.

Where scour poses a significant risk to the structure, a scour monitoring plan should be developed as a part of the bridge inspection and maintenance program. Inspection requirements, including method and frequencies, should be specified.

C11.1

Bridge types covered under these Guidelines may be of relatively short spans and would not necessarily exhibit sensitivity to wind effects. To address these cases, Section D, Article 17.2 of these Guidelines presents an approximated approach that would facilitate an early assessment of the structure’s wind-sensitivity. Bridges with the stability parameter Sw < 1.0 should have assessment of aerodynamic stability as one of the first priorities during design. Appendix D – Figure 3 is a flow chart that provides an overview of aerodynamic assessment and testing.

Preliminary assessment methods may also include but are not limited to - • Extrapolations from past projects with similar features and characteristics • Wind tunnel testing • Data obtained from various numerical simulations, including computational fluid dynamics (CFD).

C11.2

A Wind Climate Report is generally developed by a wind specialist. The Report should generally include the following: • All sources of wind data such as local meteorological stations, bridge site data and/or simulations are described including their location, type of observations and duration.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

• A brief description of the methods employed with all relevant references.

Recommended return periods for wind speeds are usually: • 20 years for design loads during construction • 100 years for design loads of completed bridge • 1,000 years for stability during construction • 10,000 years for stability of completed bridge

Recommended return periods are longer for stability evaluations than for design loads, because load factors are not applied to stability demands.

Wind directionality at the bridge site should be defined by fitting the wind data at the site with Weibull type distribution functions.

When applicable, ice and snow accumulation data should be included (expected thickness of ice accretion, maximum daily snowfall, etc).

Davenport or Von Karman spectral turbulence models could be used. It should be noted however that the original Davenport spectrum does not include the effect of elevation. A detailed explanation of turbulence is given by Simiu and Scanlan (C10). For bridges it is recommended to use the turbulence model provided by ESDU 85020/86010/86035 (C21) (C22) (C23) as being the most comprehensive. It is based on the modified Von Karman spectra which are generally recognized as the best model of isotropic turbulence. The isotropic model implies that the turbulence correlations are invariant of direction - an assumption commonly used for engineering applications.

C11.3

Wind tunnel and field testing are not performed merely to confirm aerodynamic stability of a design and to provide information needed for establishing design wind loading, they often serve as design tools in the selection of appropriate enhancement to the structure’s cross-sections that would bring about the desired performance.

Ice and snow accumulation may cause any cross-section to become aerodynamically unstable. In general, the concern is in the lower galloping and/or flutter speeds.

C11.3.1

To ensure the required 2 dimensional test conditions (2D), the length-to-width model ratio should be no less than 4. End plates attached to the model extremities can improve 2D flow conditions. The effect of turbulence on stability should be investigated. Turbulence is often found beneficial in reducing vortex shedding, but this is not necessarily true for all cases.

When simulating turbulence, a correction on its intensity should be implemented to simulate full scale turbulence effects.

Large scale turbulence is difficult to model unless sectional model tests are carried out in a very large wind tunnel facility. In smaller , it could be simulated using active gust generators.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

Special attention should be paid to scaling details such as barriers, rails and deck slots to the Reynolds number effects. When testing and/or interpreting test results from cross sections with curved surfaces, Reynolds number considerations must be implemented.

When vortex shedding is identified, the wind speed should be adjusted until the peak response is found.

Section model tests should allow for vertical and torsional motions (2 degrees of freedom) of the model. Depending on the project specific requirements, it may also be necessary to include an investigation of coupling effects with lateral motions.

The frequency ratio of vertical to torsional motions on a sectional model test (frequency ratio) should be taken as close as possible to that which has been computed for the bridge, and should in no case be taken as higher than the computed ratio.

Only modes with similar shapes should be considered as likely candidates for coupling. Theoretically, a bridge with frequency ratio that is close to unity (i.e. < 1.5) is susceptible to flutter at very low wind speeds.

Static force and moment coefficients should be measured outside of the vortex shedding lock-in range and at speeds lower than the flutter speed.

Drag coefficients should be normalized with respect to the deck height (including traffic barriers) whereas lift and moment coefficients should be normalized with respect to the width of the deck section (edge to edge). Drag refers to the air force along the direction of mean wind flow, whereas lift is across the flow force.

C11.3.2

The Designer and wind consultant should establish the need for aeroelastic model tests.

In spite of recent advances in analytical methods, aeroelastic model tests continue to be recognized as the most effective tools for evaluation of full scale stability of long span bridges under wind loads, addressing effects involving complex modal coupling, 3D wind effects and directionality, and wind turbulence.

Traffic congestions will cause higher loads on the deck due to the increased lateral projected area on the deck. In special cases such as long trains over a bridge, vortex shedding vibrations may occur. The probability of simultaneous occurrence of wind and traffic congestions should be considered. This typically amounts to reductions of the concurring wind speed.

When testing suspension bridges, the velocity scale = sqrt of length scale. This is based on the Froude number, a dimensionless number that considers gravitational effects. By following the Froude scaling principles, different scaling numbers can be derived that can be used to design an aeroelastic model and to re-scale the results. With the gravitational effects being important when designing an aeroelastic model for a suspension bridge, following the Froude scaling principles will also ensure that appropriate cable tensions are modeled. See reference: Simiu, E., Scanlan R. H., Wind Effect on Structures: Fundamentals and Applications to Design, John Wiley & Sons, Inc., New York, N.Y., 1996. (C10)

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

C11.3.3

The force-balance test involves direct measurements of mean forces and moments acting on rigid test sections that are representative of bridge deck sections (or trusses). The primary goal of the test is to understand the bridge sections’ sensitivity to wind directionality and sheltering effects.

The typical set-up in the wind tunnel comprises a tested section (strip) flanked by the rest of the structure to simulate the expected sheltering effects relative to different directions of wind attack. The test section is physically separated from the flanking elements and instrumented for measurements of lateral and longitudinal forces for various wind conditions and construction stages.

The wind directionality and sheltering effects obtained from the test can be used for derivation of wind loads due to wind buffeting. In addition to the normal loads due to drag, lift and torsion about the longitudinal axis, buffeting response analysis can be performed to generate wind loads acting along the structural elements, taking into account bridge dynamic properties and natural turbulence at the site. This response analysis can ultimately produce quasi-static wind loads, along with estimates of dynamic resonance and background (also known as direct gust) load effects.

C11.3.4

Examples of field tests include: • Modal identification • Damping measurements • Performance verifications such as stay tensions, damper system and user comfort • Installation and routine verification of vibration controlling devices • Measurements of local wind conditions and responses

C11.4

Flutter Speed can be found analytically using aeroelastic coefficients such as the Aerodynamic Derivatives.

Aerodynamic derivatives should be in accordance with the Scanlan definitions. (C24)

This information is required for 3D flutter response analyses.

C11.5

Bridges with acceptable stability should meet all imposed project specific stability and comfort criteria.

Mean wind loads vary with bridge elevation. Unless project specific requirements dictate otherwise, WX (extreme wind) may be calculated using reference wind velocity with 2000-year return period and scour with 100-year return period.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

Sheltering effects of one part of the bridge over another may be included when applying mean and direct gust loads. However, the extent of reduction is subject to verification by testing (see Article 11.7).

The direct gust (also called background) loads are caused by the fluctuating wind pressures acting on the exposed areas. They may be obtained from the mean pressures by applying appropriate gust factors. The effect associated with the lack of correlation over the bridge span shall be included.

Turbulent winds comprise gusts of various sizes. Whether or not a gust would be transmitted as a load on the cross-sections of the exposed bridge elements, i.e., deck, tower, depends on the size (and frequency) of the gust. The bridge elements act as frequency filters on the direct loads. This effect is typically implemented using the so called aerodynamic admittance. (C10)

The motion-induced loads (also called resonance) are proportional to various bridge modes of vibration.

Aerodynamic damping shall be taken into account and could greatly vary among different excited modes. It could modify the total damping very significantly.

The aerodynamic damping can best be estimated using the unsteady aerodynamic theory. In the absence of unsteady coefficients (aerodynamic derivatives), it could be evaluated using the quasi-steady buffeting theory. (C25) (C26) (C27)

The dynamic excitations caused by turbulence are complex where various bridge modes could be excited. Since the wind power in its spectra decreases rapidly with frequency, bridge modes of more than about 1.0 Hz are normally excited to much lower amplitudes. On some sections such as boxes, due to aerodynamic admittance amplification effects, higher frequencies may be excited. This effect is most pronounced for lift and moment loads and must be determined via admittance function measurements.

Limited amplitude vortex shedding causes motion- induced loads. These motion-induced loads are at the lock-in wind speed which is typically below the design wind speed. These loads have to be combined with the buffeting loads at that corresponding speed, which produce an additional load case that should not be neglected.

When ice and snow or traffic congestion cases are considered, a significant increase in the lateral loads can be expected due to the added exposed areas.

Ice and snow accumulation plus traffic congestion, if considered, should be combined at a reduced design wind speed. This project specific speed should be established by the wind consultant in conjunction with the owner. Different traffic should apply for different usage. One example would be traffic congestion on hurricane evacuation routes.

In the absence of a project specific speed, the reduced wind speed may be taken as 50% of the design speed, which is based on the ASCE recommendations (C11).

C11.7

Recent wind tunnel studies of long span bridges have uncovered that the recovery of mean pressures on the leeward sections occurs over surprisingly short separations. It has been

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

shown that the wake behind the windward elements could generate wind pressure that is even higher than the ambient gust of the incoming turbulent wind. The complex flow of air caused by the wake, when combined with the dynamic response of the leeward elements, could in fact result in peak loads in the leeward elements that are higher than the windward elements.

C12.1

The partition between Zone 1 and Zone 2 in LRFD is lowered from 0.15g to 0.10g. This is done in recognition of the criticality of long span bridges, as well as the substantial financial investment in construction.

Seismic Design Categories (SDCs) in the AASHTO Guide Specifications for LRFD Seismic Bridge Design (C28) correspond to Seismic Performance Zones in LRFD.

Long span bridges are generally considered to be critical bridges, which must continue to function as a part of the lifeline, social/survival network and as an important link for civil defense, police, fire department and/or public health agencies to respond to a disaster situation after the seismic event. Therefore these Guidelines are intended to achieve no to minimal damage to bridges during reasonably expected earthquake ground motions (Function Evaluation Earthquake) during the design life of the bridges and to achieve limited to moderate (repairable) damage during rare, high-amplitude earthquakes (Safety Evaluation Earthquake). Bridge owners may choose to adopt higher or lower levels of bridge performance based on the importance and the social impacts of the bridge.

The varying levels of damage during functional and safety evaluation earthquakes may be generally described as follows:

No Damage is defined as full serviceability without repair or replacement. For structural members, “no damage” can be defined as the nominal capacity as described in the AASHTO LRFD Bride Design Specifications (C01). Nominal material properties should be used instead of expected material properties. Increased member strength due to the effects of confinement steel should be ignored. For elements of the cable system, “no damage” is presumed to occur if stresses don’t exceed the proportional limits of the materials. For structural components such as bearings, expansion joints, etc. “no damage” implies preservation of full serviceability without repair or replacement.

Minimal Damage: Although minor inelastic response may occur, post-earthquake damage should be limited to narrow cracking in concrete, and inconsequential yielding of secondary steel members. There should be no apparent permanent deformations. Damage to non- structural components of the cable system would be allowed.

Moderate Damage is damage that can be repaired with a minimum risk of losing functionality, i.e. without closing the bridge. Inelastic response may occur, resulting in concrete cracking, reinforcement yielding, minor spalling of cover concrete and minor yielding of structural steel. The extent of damage should be sufficiently limited such that the structure can be restored essentially to its pre-earthquake condition without replacement of reinforcement or replacement of structural members. Small permanent offsets, which do not interfere with functionality of the bridge, would be allowed.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

Significant Damage: Although the risk of collapse should be minimal, significant damage may require closure of the structure for repair. Expected damage would consist of concrete cracking, reinforcement yielding, major spalling of concrete and deformations in minor bridge components which may require closure of the bridge for repair. Partial or complete replacement of secondary elements may be required in some cases. Secondary elements are those that are not a part of the gravity load resisting system. There may be permanent offset of the structure.

Allowable displacements are constrained by geometric, structural and geotechnical considerations. The most restrictive of these constraints will govern displacement capacity. These displacement constraints may apply to either transient displacements, as would occur during ground shaking, or permanent displacements, as may occur due to seismically induced ground failure or permanent structural deformations or dislocations, or a combination. The extent of allowable displacements depends on the desired performance level of the bridge design.

Geometric constraints generally relate to the usability of the bridge by traffic passing on or under it. Therefore, this constraint will usually apply to permanent displacements that occur as a result of the earthquake. The ability to repair such displacements or the desire not to be required to repair them should be considered when establishing displacement capacities. When uninterrupted or immediate service is desired, the permanent displacements should be small or non-existent, and should be at levels that are within an accepted tolerance for normally operational bridges of the type being considered. A bridge designed to a performance level of simply no collapse could be expected to be unusable after liquefaction, for example, and geometric constraints would have no influence. However, because life safety is at the heart of the no collapse requirement, jurisdictions may consider establishing some geometric displacement limits for this performance level.

C12.2

The conventional bridge design provisions have one single level of design earthquake ground motion hazard but use three implied performance objectives for small, moderate, and large earthquakes. The proposed two-level design earthquake hazards approach provides more definite performance objectives and damage states for two design earthquakes with explicit design checks to ensure that the performance objectives are met.

The upper-level event, termed the “rare” or Safety Evaluation Earthquake (SEE), describes ground motions that, for most locations, are defined probabilistically and have a probability of exceedance of 4 percent in 100 years. However, for locations close to highly active faults, the mapped SEE ground motions are deterministically bounded so that the levels of ground motions do not become unreasonably high. Deterministic bounds on the ground motions are calculated by assuming the occurrence of maximum magnitude earthquakes on the highly active faults. These are assumed to be equal to 150 percent of the median ground motions for the maximum magnitude earthquake. Deterministic bounds are generally applied in high-seismicity portions of California, in local areas along the California-Nevada border, along coastal Oregon and Washington State, and in high-seismicity portions of Alaska and Hawaii.

The lower level design event, termed the Functional Evaluation Earthquake (FEE) has ground motions corresponding to 50 percent probability of exceedance in 100 years. This event ensures that essentially elastic response is achieved in the bridge substructure for the more frequent or a reasonably expected earthquake during the design life of the bridge. This design level is similar to the flood and wind load design and has similar performance objectives.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

C12.3.1

The main purpose for conducting a site-specific probabilistic ground motion study is to develop ground motions that are more accurate for the local seismic and site conditions than can be determined from national ground motion maps (i.e., based on USGS study) and the general procedure of Section 12.3.2. The site-specific study is also necessary when the national hazard maps are not available for the design hazard levels (in terms of return-period or probability of exceedance) specified for the bridge. Accordingly, such studies should be comprehensive and incorporate current scientific interpretations on a regional scale. Because there are typically scientifically credible alternatives for models and parameter values used to characterize seismic sources and ground-motion attenuation, it is important to incorporate these uncertainties formally in a site-specific probabilistic analysis. Examples of these uncertainties include seismic source location, extent and geometry; maximum earthquake magnitude; earthquake recurrence rate; and ground-motion attenuation relationship.

Near-fault effects on horizontal response spectra include:

• Higher ground motions due to the proximity of the active fault, • Directivity effects that increase ground motions for periods greater than 0.5 seconds if the fault rupture propagates toward the site, and • Directionality effects that increase ground motions for periods greater than 0.5 seconds in the direction normal (perpendicular) to the strike of the fault.

If the active fault is included and appropriately modeled in the development of national ground motion maps, then the first effect is already included in the national ground motion maps. The second and third effects are not included in the national maps. These effects are significant for periods longer than 0.5 seconds and normally should be evaluated for long span bridges. Further discussions on the second and third effects are contained in Somerville (1997) (C12) and Somerville et al. (1997) (C13). The ratio of vertical-to-horizontal ground motions increases for short-period motions in the near-fault environment.

C12.3.2

Refer to the commentary in Article C3.4.1 of the AASHTO Guide Specifications for LRFD Seismic Bridge Design (C35).

C12.3.2.1

Procedures described in this article were originally developed for computing ground motions at the ground surface for relatively uniform site conditions. For long span bridges it is very likely that the soil conditions at different foundation and anchorage supports are different. Seismic response of long span bridges may be particularly sensitive to the spatially varying ground motion effects. These variations are not always easily handled by the simplified general procedures described herein. Therefore, to obtain a more accurate assessment of the site effects as well as spatially varying site effects, site-specific response analysis is recommended for long span bridges.

In addition, depending on the relative stiffness of the foundations to the surrounding ground and the level of the ground motion, the governing ground motion could be near the surface or at depth. This creates some issues as to the location of the motion to use in the bridge design. For

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

critical bridges such as long span bridges, it may be necessary to use more rigorous numerical modeling, taking into account soil-foundation interaction effects to represent these conditions.

C12.3.2.2

For long span bridge design, it is recommended that in-situ shear-wave velocity measurements be conducted, particularly if the site profile is non-uniform, or if the average velocity computed in the methods described herein does not appear reasonable, or if the project involves special design issues. In all evaluations of site classification, the shear-wave velocity should be viewed as the fundamental soil property, as this was used when conducting the original studies defining the site categories.

An alternative to applying Eqs. 12.3.2.2-2, 12.3.2.2-3, and 12.3.2.2-4 to obtain values for N ,

Nch and su is to convert the N-values or su values into estimated shear wave velocities and then to apply Eq. 12.3.2.2-1. Procedures given in Kramer (1996) (C14) can be used for these conversions.

Depth of Motion Determination: For conventional bridges supported on shallow spread footings, the motion computed at the near ground surface is appropriate. However, for long-span bridges where stiff/large-diameter deep foundations (i.e., driven piles or drilled shafts) are often used to support the bridge pier, the location of the governing ground motion will depend on the horizontal stiffness of the foundation system relative to the horizontal stiffness of the surrounding soil. If soil is very stiff relative to the foundation, then the motion should be defined near the pile cap. If the foundation system is stiff relative to the surrounding soil, then the controlling motion will likely be at some depth below the ground surface. Typically this will be approximately 4 to 7 pile diameters below the pile cap or where a large change in soil stiffness occurs. For cases where the controlling motion is more appropriately specified at depth, site- specific ground response analyses can be conducted to establish ground motions at the point of fixity. This approach or alternatives to this approach should be used only with the owner’s approval.

C12.3.2.3

Refer also to the commentary in Article C3.4.2.3 of the AASHTO Guide Specifications for LRFD Seismic Bridge Design (C28).

C12.3.3

Characteristics of the seismic environment of the site to be considered in selecting time histories include: tectonic environment (e.g., subduction zone; shallow crustal faults in western United States or similar crustal environment; eastern United States or similar crustal environment); earthquake magnitude; type of faulting (e.g., strike-slip; reverse; normal); seismic-source-to-site distance; local site conditions; and design or expected ground-motion characteristics (e.g., design response spectrum; duration of strong shaking; and special ground-motion characteristics such as near-fault characteristics). Dominant earthquake magnitudes and distances, which contribute principally to the probabilistic design response spectra at a site, can be determined from the site specific hazard analysis or, if national ground motion maps are used, from the deaggregation information on the U.S. Geological Survey website: http://geohazards.cr.usgs.gov/.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

It is desirable to select time histories that have been recorded under conditions similar to the seismic conditions at the site listed above, but compromises are usually required because of the multiple attributes of the seismic environment and the limited data bank of recorded time histories. Selection of time histories having similar earthquake magnitudes and distances, within reasonable ranges, is especially important because they have a strong influence on response spectral content, response spectral shape, duration of strong shaking, and near-source ground- motion characteristics. It is desirable that selected recorded motions be somewhat similar in overall ground motion level and spectral shape to the design spectrum to avoid using very large scaling factors with recorded motions and very large changes in spectral content in the spectrum-matching approach. If the site is located within 6 miles of an active fault, then intermediate-to-long-period ground-motion pulses that are characteristic of near-source time- histories should be included if these types of ground motion characteristics could significantly influence structural response.

Similarly, the high short-period spectral content of near-source vertical ground motions should be considered.

Ground-motion modeling methods of strong-motion seismology are being increasingly used to supplement the recorded ground-motion database. These methods are especially useful for seismic settings for which relatively few actual strong-motion recordings are available, such as in the central and eastern United States. Through analytical simulation of the earthquake rupture and wave-propagation process, these methods can produce seismologically reasonable time series.

Response spectrum matching approaches include methods in which time series adjustments are made in the time domain (Lilhanand and Tseng, 1988 (C15); Abrahamson, 1992 (C16)) and those in which the adjustments are made in the frequency domain (Gasparini and Vanmarcke, 1976 (C17); Silva and Lee, 1987 (C18); Bolt and Gregor, 1993 (C19)). Both of these approaches can be used to modify existing time histories to achieve a close match to the design response spectrum while maintaining fairly well the basic time-domain character of the recorded or simulated time histories. To minimize changes to the time-domain characteristics, it is desirable that the overall shape of the spectrum of the recorded time history not be greatly different from the shape of the design response spectrum and that the time history initially be scaled so that its spectrum is at the approximate level of the design spectrum before spectrum matching.

The requirements for the number of time histories to be used in nonlinear inelastic dynamic analysis and for the interpretation of the results take into account the dependence of response on the time domain character of the time histories (duration, pulse shape, pulse sequencing) in addition to their response spectral content. Additional guidance on developing acceleration time histories for dynamic analysis may be found in publications by the Caltrans Seismic Advisory Board.

C12.4

The requirements of this section shall also apply to temporary shoring systems that support only dead loads and construction live loads (no vehicular live load).

The option to use a reduced response coefficient and a reduced ground acceleration coefficient reflects the limited exposure period for a temporary bridge.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

In areas of low to moderate seismicity, in lieu of designing for 50% of the design response spectra for permanent structures, temporary bridges and the temporary shoring systems that support staged construction may be designed for an equivalent horizontal load equal to a (fixed) percentage of the total dead load to be supported at the point under consideration. In this case, the equivalent horizontal load should be determined on a case-by-case basis, and will vary depending on the seismicity at the site and the flexibility (response period) of the temporary structure. This equivalent horizontal design load should be applied separately in both the transverse and longitudinal directions. Values of 10% to 15% have been suggested for the equivalent horizontal load in areas of low to moderate seismicity.

In areas of significant to high seismicity, a dynamic analysis should be performed for seismic design of the temporary bridge or temporary shoring systems.

C13.1

An example of temperature variation within a structural component can be seen in the protective sheathing of a cable stay. The temperature difference between the HDPE, or other sheathing, and the main tensile element may affect the design of any expansion joint details for the PE pipe that protects the cable. To minimize temperature effects, consideration should be given to simply specifying light colored sheathing.

C13.1.1

The Design Guidelines for Cable-Supported Steel Bridges - KSCE (2006) (C38) recommends the following:

• Temperature differential between hangers and girder - 27°F (15°C) • Between hangers and main cables - 18°F (10°C) • Between girder and main cable - 45°F (25°C).

The PTI Recommendations (C03) suggest the following for temperature difference between stay cables and the rest of the structure:

• A range of 14°F (8°C) (for light colored stays) to 40°F (22°C) (for black stays) has been used to compute effects of temperature differential between cable-stays and deck.

• For temperature differential between stays left and right of pylons (as seen in longitudinal elevation),a range of 7°F (4°C) (for light colored stays) to 22°F (12°C) (for black stays) has been used in recent cable stayed bridge.

A differential gradient of 22°F (12 °C) has been specified for steel arch ribs and orthotropic steel box deck in recent designs.

A recent major cable-stayed bridge has used a temperature difference of 36°F (20°C) between its steel deck and the rest of the structure.

A temperature difference over the width of the deck was taken as 18°F (10°C) between the edges of the steel deck, and 9°F (5°C) between the edges of the concrete deck on a recently completed cable-stayed bridge.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary C

A temperature difference of 9°F (5°C) through the cross-section of the concrete tower has been specified in past projects.

C13.2

BS 5400 (Part 4 Appendix C) (C20) gives modification factors to account for the effect of reinforcement on creep and shrinkage.

C13.3

The effects of irreversible displacement are considered similar to differential settlement, which should be predicted and accounted for during design.

Locked-in stresses in flexible structures can often be partially relieved by adjusting cables, bearing heights, etc. This would be an acceptable design strategy if reasonable planning and provisions are made to detect the effects of impact and to repair the structure after impact.

C14.0

Site specific studies may involve a detailed study of the vessel population that can be expected during the design life of the bridge. Refer to the AASHTO Guide Specifications and Commentary for Vessel Collision Design of Highway Bridges (C39) for specific guidance.

The state of practice calls for the correlation between analytical and test results. The level of refinement in both analysis and testing is tied to the demand associated with the size of the design vessel and the climatic conditions at the bridge site. Detailed non-linear finite element analyses have been used to simulate the energy absorption associated with crushing of the vessel as well as the impact attenuators (such as dolphins and sand islands) to come to a more realistic assessment of the demand. Scaled tests have been used traditionally as proof testing, but in some cases used as a means to calibrate mathematical models that would then be relied upon in carrying out the design. (C37)

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

SECTION D COMMENTARY – STRUCTURAL ANALYSIS & EVALUATION

C15.0

Considerations addressed in these Guidelines will serve to highlight provisions of AASHTO LRFD that are most pertinent to long span design and to elaborate on the state of practice.

Performing complex modeling that articulates every member in the structure is not a requirement for the analysis and design of long span bridges, and often is not the preferred approach.

Relatively simple tools, such as 2-D models and simplified global 3-D models, are frequently used in design and analyses. Such tools are useful for the designers to acquire a concise understanding of the characteristics and sensitivity of the structure being considered. More sophisticated finite element analyses are often used to supplement the simpler 2-D and 3-D global models for the evaluation of localized stresses in the structure.

3-D FEM computer analyses, with representation of the torsional rigidity of the structure and soil-structure interaction, are typically needed for the evaluation of the structure for aerodynamic stability, seismic vulnerability and other extreme events.

Scour is not a load effect, but a condition that will affect stiffness and resultant forces. The load effects that should be considered under varying scour conditions include: • Earthquake • Ship collision load • Wind load

Refer to Article 15.4 of these Guidelines for a more detailed discussion of soil-structure interaction effects and analysis techniques to model these effects. Also, refer to Section G (Article 26.3) of these Guidelines for additional discussions on soil-structure interaction and foundation modeling and analysis for pile, drilled shaft and caisson foundations.

C15.1

Refer to Article 15.4 and Section G Article 26.3 of these Guidelines for discussions on the treatment of nonlinearities associated with foundation modeling and soil-structure interaction effects.

Suspension Bridges

The suspension cables are the primary source of nonlinearity of a suspension bridge. Figure 15.1-1 shows a generalized force-displacement relationship for cable-supported structures (e.g. suspension bridges).

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

Figure 15.1-1 Generalized Force-Displacement Relationship for Cable-Supported Structures

In addition to being the principal load carrying elements, the cables are usually the first elements to be constructed. This makes nonlinearity a more prevalent effect for suspension bridges, than for other long span bridges. Therefore, nonlinear modeling is essential for suspension bridges.

Cable-Stayed Bridges

The principal causes of nonlinearity in cable-stayed bridges are – • Nonlinearity due to cable sag. It is generally accepted that the true cable may be replaced with a straight elastic member that has a reduced modulus of elasticity to account for the sag effect. Figure 15.1-2 shows the cable sag and the principal parameters that affect its stiffness. The reduction factor r can be estimated by the following formula: 2 5 2 2 r = 1/ [1+ W COS αEcAc(H1+H2)/24H1 H2 ] where W = Total Weight of the Cable

H1 & H2 = Horizontal Component of Cable Force at the beginning and the end of loading, respectively α = Angle of Chord with Horizontal

Ac = Area of the Cable

Ec = Modulus of Elasticity of Cable It should be apparent that the reduction factor r approaches unity (i.e. minimal reduction of stiffness) when the weight of the cable, W, is small or the horizontal component of the cable force, H, is large. The above equation is in essence Equation 4.6.3.7-2 of AASHTO LRFD. The reduction factor is isolated above to better illustrate the relationship between cable weight and cable tension.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

Figure 15.1-2 Sag Effects for Stay Cables

• Beam column effect (P-Delta Effect). This is caused by axial forces acting on the deformed structure. This effect cannot be solved directly and can only be accounted for through an iterative solution. • Material Nonlinearity. While most materials exhibit nonlinear behavior, they may be assumed to behave linearly in their elastic or near elastic state for most loading combinations required of a cable-stayed bridge, except in the case of cable loss. (Refer to the discussion of the cable loss case under Article 17.4 of these Guidelines.)

Since nonlinearity of cable-stayed bridges varies depending on the contributing factors discussed above, its treatment should be bridge specific. Generally, nonlinearity becomes significant as span lengths increase. Also, nonlinearity is most pronounced during construction because the cable stresses are low.

Arch Bridges

Second order effects in the arch ribs may be very significant. It should be noted that buckling lengths given in AASHTO LRFD are not valid for tied arches.

Moment magnification (see AASHTO LRFD Article 4.5.3.2.2) may be used as an approximated method, but should not be used for final design without addressing its limitations, especially for bridges with slender arch ribs. It should be noted that the moment magnification equations do not take into account the reduced stiffness of steel structures under ultimate conditions. Also, destabilizing effects of spandrel columns should be addressed. With the advent of computing technology, it is advisable to incorporate nonlinear analyses into the design of long span arch bridges.

C15.1.1

There are many commercially available computer programs that can perform non-linear analyses for long span bridges. However, special computer programs will be necessary if a significant level of stress is expected to be locked into the final structure during construction. Beyond being able to account for nonlinearity generally, the appropriate program should be able to manage the incremental and partial results, and to store and retrieve the structural system (in terms of both stiffness and structural deformations – and not just stiffness) from a previous

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

construction step, and use this previous step as the starting point for performing the next non- linear load increment.

AASHTO LRFD Article 4.5.3.2.1 states that “Only factored loads shall be used and no superposition of force effects shall be applied in the nonlinear range. The order of load application in nonlinear analysis shall be consistent with that on the actual bridge”, and the corresponding commentary reads “Because large deflection analysis is inherently nonlinear, the loads are not proportional to the displacements, and superposition cannot be used. Therefore, the order of load application can be important and traditional approaches, such as influence functions, are not directly applicable.” However, superposition of force and deflection effects from separate nonlinear analyses for different loadings may be an acceptable practice for preliminary design purposes or for expediting design decisions, providing locked-in stresses in the structure and critical intermediate conditions are not overlooked.

C15.1.2

Previous investigations have demonstrated that the response of most long span bridges is close to linear under live load and other transient loads. Therefore, upon verification that the nonlinearity is insignificant, it is an acceptable practice to perform separate linear or nonlinear analyses for live loads and other transient loads, and have these loads combined with other nonlinear loadings, such as dead load effects (i.e. superposition of load effects from separate nonlinear analyses).

It may be possible to show that a long span cable-stayed bridge, or an arch bridge, can be analyzed linearly without resulting in inaccuracies associated with nonlinearities. In these cases, the use of superposition for expedience should not be ruled out.

The effects of creep and shrinkage should be computed in appropriate increments to ensure that their corresponding deformations and secondary forces are adequately represented in the analysis.

The Designer should be cognizant of the fact that superposition embodies approximation employed to expedite computational exercises. Results should therefore be applied to design with an appropriate degree of conservatism. Where analysis software has superposition built in, the limitations associated with the computation method should be stated in the design documentation.

C15.2

A Designer employed by an owner in a conventional design/bid/construction contract should exercise care to develop a viable scheme for the construction of the structure. Development of the scheme should include reasonable estimates of the weights and locations of major equipment, and the sequence of construction. While it would not be practical to conduct, during design, a stage-by-stage stress analysis of each operation with an assumed construction sequence, the analysis should be of sufficient detail to determine that the structure can be constructed by a viable construction scheme, and that scheme should be shown in the Contract Plans.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

C15.3

The stage-by-stage analysis is usually performed by a construction engineer who works closely with the contractor to validate the chosen construction scheme, to check stresses according to the step-by-step erection procedures, and to generate data needed for construction control.

An independent analysis is often also performed by the owner’s engineer who assesses the effects of the contractor’s proposed means-and-methods on the final structure.

A detailed stage-by-stage analysis that reflects the as-built conditions should form the base (dead load) system for rating and future assessment of the structure during its service life. This analysis is usually performed by an engineer under contract with the owner.

In general, the stage-by-stage analysis performed by the construction engineer requires a greater degree of accuracy, because the data from this analysis, such as camber and deflections, cable adjustment forces and elongations, are used to gauge and control the contractor’s field operations. Surveyed camber and measured cable forces not being reasonably close to the predicted values may be cause for stopping work until the difference can be explained.

C15.4

Soil-structure interaction effects can be included in the analytical (structural) models using either:

• The de-coupled approach (substructure method), where the foundation in the structural model is replaced by equivalent stiffness and damping matrices, or

• The fully-coupled approach, where the full length of the foundation shaft is explicitly modeled and the behavior of the surrounding soil is represented by equivalent soil springs (p-y, t-z and q-z springs). These soil springs can be either linear or nonlinear.

An approach that is used by some designers is to consider soil-structure interaction effects using the de-coupled approach for the conceptual and preliminary design phases. In the event that a more detailed representation of the complex interactions between the superstructure, foundation and the surrounding soil is required, a fully-coupled soil-structure interaction analysis would be conducted in the final design phase.

Soil-structure interaction considerations can be important for the design of both the structural systems and the foundation systems of long span bridges. There are four primary interaction effects that should be considered in any soil-structure analysis:

1. The influence of foundation stiffness on structural response. 2. The inertial structural loads imparted to the foundation system – termed as the inertial effect. 3. The ground displacement loads imparted to the foundation system (resulting from both free- field soil displacement and possible ground-failure conditions) – termed as the kinematic effect. 4. Modified (scattered) effective support motions due to the near field soil-structure interaction.

Depending on the degree of uncertainty in the soil parameters used for design and analysis, it is sometimes appropriate to perform two sets of analysis to bound the solution – one analysis

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

using “upper bound” stiffness properties and another analysis using “lower bound” stiffness properties.

When drilled shafts are used and the casings are to be left in place, the casings should be included in the model properties for an upper bound stiffness analysis. This should be done even where the casings are not included in calculating the resistance of the shafts.

Where significant, radiation damping effects and the elasto-dynamic behavior of the far-field soil outside the nonlinear soil region adjacent to the foundation should also be included in the analysis.

Additional recommendations and considerations regarding soil-foundation-structure interaction analysis techniques may be found in the literature, including the following: • Ho, et. al., “Seismic Retrofitting Guidelines for Complex Steel truss Highway Bridges”, Chapter 4 - Structural Analysis (D14) • Parsons Brinckerhoff, Inc., Geotechnical Engineering Circular No. 3, Chapter 8 – Geotechnical Seismic Design for Transportation Structures and Soil-Foundation-Structure Interaction (D38) • Tseng and Penzien, “Soil-Foundation-Structure Interaction”, Chapter 42, Bridge Engineering Handbook (D39) • Lam and Law, “Soil Structure Interaction of Bridges for Seismic Analysis”, Report MCEER- 00-0008 (D40)

Also refer to Appendix F of these Guidelines.

C15.4.1

A general procedure for developing pile foundation stiffness is outlined in Appendix F of these Guidelines.

C15.4.2

To incorporate the pile cap stiffness into the pile group, the following procedure should be used (D15).

(1) Estimate the ultimate passive soil pressure capacity on the front vertical face using the procedure depicted in Figure 15.4.2.1. For frictional soils, the passive pressure coefficient, Kp, should be derived using the log-spiral solution presented in Figure 15.4.2.2. (2) Construct an elasto-plastic load-deflection curve to represent the stiffness characteristics of the pile cap. Experimental data has indicated that the elastic limit occurs at a deflection of about 0.02 times the embedment depth, Z. The elastic slope is, therefore, defined as the ratio of the ultimate capacity to 0.02 times the embedment depth of the cap.

The foregoing procedure is applicable to stable level ground conditions. For cases where the footing is located adjacent to an slope or where poor, unstable soils exist, the pile cap stiffness would be significantly reduced. In liquefiable soils, stiffness attributable to the soils around the cap should be ignored.

It should be noted that it is generally uneconomical to allow the shear load to control the number of required piles, considering that the pile is effective in mobilizing soil resistance at only about

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

the upper five pile diameters. Therefore, the following design strategies can be used to increase the resistance to shear load (D05).

• Use of thicker or large footings and including the pile cap resistance at normal, stable soil sites. • Use of deeper pile footing embedment, which would increase the resistance of both the pile and pile cap • Modification of the pile top connection detail to achieve a greater degree of pile head fixity (e.g., embedding the pile top deeper into the pile cap) • Strengthening the structural capacity of the pile at approximately the upper ten pile diameters • Use of more ductile pile types that can develop soil resistance to a higher amplitude of pile deflection • Soil improvement at shallow depths around the pile footing and pile head at poor soil sites.

Figure 15.4.2.1 Method for Passive Pressure Capacity of Pile Cap

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

Figure 15.4.2.2 Passive Pressure Coefficient, Kp

C15.4.3

Ground displacement loading can be sub-divided into two categories: (1) free-field ground displacements, and (2) displacements due to unstable ground such as liquefaction induced lateral spread or unstable embankments/slopes. Ground displacements impose forces acting along the length of the piles and pile cap. Figure 15.4.3.1 schematically illustrates the two loading mechanisms. These two loading mechanisms could occur simultaneously during an earthquake, and therefore should be considered as such in the design. For the free-field ground displacements, the resulting forces can be estimated by imposing the estimated free-field ground displacement profile on the pile through p-y springs.

Proper selection of the non-linear p-y properties of the surrounding soil is crucial for the design. The displacement profile can be estimated from a site-specific response analysis (see Section C, Article 12.3 of these Guidelines). In competent sites, the free-field ground displacements generally do not govern the pile design because the curvature of the ground displacement is small. This effect, however, has to be considered for piles in soft soils and for sudden changes in soil stiffness or depth. The effect is particularly significant for large diameter piles or drilled caissons in soft soils.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

The ground displacement effect should be considered for piles/drilled shafts under the safety evaluation earthquake conditions. Whenever possible, stronger connection details and more ductile pile types should be used at poor soil sites to improve the chance of survivability of the foundations.

Pile design in liquefied soil deposits presents a more difficult case. The overall evaluation procedure would essentially be the same as that described above for the free-field ground displacement case. However, the choice of p-y characteristics must properly consider liquefaction effects of the soils. A number of studies have recently been conducted to address this issue. Recommended guidelines for pile design in liquefied soil deposits are provided by Lam, et al. (D15).

Figure 15.4.3.1 Mechanism in Soil-Pile Interaction Problem

C15.4.4

Due to the complex interaction between soil, pile, and structures, the effective support motions (i.e., the near field ground motions) at the foundation/structure interface differ from those in the free field. For regular shallow footings and flexible pile-supported footings (relative to the surrounding ground), free-field motions are considered satisfactory as the support motions in the structure response analysis.

For very large and stiff foundations, such as large caissons, very stiff battered pile groups, or very large pile groups (i.e., pile-reinforced soil mass mechanism), the effective support motions at the foundation/structure interface may differ considerably from the free-field motions. When this situation occurs, a more refined analysis taking into account the presence of the foundation should be performed to derive the effective support motions.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

C15.5

Distortion-induced loads – temperature (TU), creep (CR), shrinkage (SH) and secondary post- tensioning forces – may control the size and strength of columns and piers, particularly if the analysis is run using gross section properties. This may result in an overly conservative (stiffer) substructure design, which may in turn adversely affect the seismic performance and penalize the design for seismic loads. Therefore, particularly for long span bridges, it is recommended that the analysis for non-seismic, extreme wind and seismic loads be based on the use of effective, partially cracked moments of inertia for the columns, piers and drilled shafts.

Section 5.6 of the AASHTO Guide Specifications for LRFD Seismic Bridge Design (D01) and Section 5.6 of Caltrans Seismic Design Criteria (D09) present guidelines for the use of effective (partially cracked) section properties in analysis of bridge structures.

This is especially true for elements that have significant compressive loads, i.e. axial loads are relatively high compared to the axial capacity in absence of bending (Paxial/(Ag f’c) ≥ 0.35). An example of this case would include the pylon and lower tower leg of a single-pylon cable-stayed bridge.

Prestressed concrete elements are generally designed elastically and expected cracking in prestressed concrete is minimal. Therefore, gross section properties should generally be used for prestressed elements.

Load Factors and Section Properties used in Analysis (Effective vs. Gross)

Care should be exercised to maintain consistency in applying effective section properties to analyses. For example, curvature due to creep in a cracked, reinforced concrete section in flexure is only a fraction of what would be predicted using gross section properties.

Note that per Article 8.1.6 of these Guidelines (Section C), load factors associated with imposed deformations, i.e. temperature (TU), creep (CR), and shrinkage (SH), shall be taken as 1.0.

Effective Torsional Moment of Inertia

Reduction of torsional moment of inertia should be quantified if under-estimation may result in non-conservative designs. Prestressed elements are examples where the equation given in Section 5.6.5 of the AASHTO Guide Specifications for LRFD Seismic Bridge Design (D01) may not apply.

C16.1.1

NCHRP Report 543 (D22) and “Evaluation of Effective Slab Width for Composite Cable-Stayed Bridges” by Byers and McCabe (D08) include comprehensive studies of the analysis of cable- stayed decks.

An enormous amount of data being generated and managed and post-processed is normal for the analysis of long span bridges, and cable-stayed bridges in particular. The engineer’s judgment and command of the necessary computer software are often critical in limiting the size of the analytical tasks.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

Effective flange widths of a superstructure during construction stage analyses are often drastically different from those of the final structure. This is especially important for a cable- stayed bridge.

Large transverse elements, such as a diaphragm over the anchor piers of a cable-stayed bridge, may have significant effects on the local stresses in the deck. While under-estimating the effective flange widths may produce acceptable conservatism in the design of the deck elements, the discrepancies between the theoretical and the actual behavior of the structure may not be acceptable for construction engineering and as-built records purposes (See Article C 15.3 for the required accuracy of construction engineering analyses).

C16.1.2

It is common practice that assumptions such as effective flange widths of orthotropic deck are verified by refined analyses.

C16.2

Overly slender arch ribs and tower legs are seldom cost-effective solutions for long span designs. The reason is that slenderness, while being able to reduce the exposed area of the structure to wind, often causes increases in dynamic wind effects that would offset any savings associated with the reduced area. Also, constructability and aesthetics tend to set a limit on the slenderness of principal elements.

Approximate methods in the AASHTO LRFD specifications may not apply to tied arches as the elongation of the tie element and the displacement at the base of the arch are not addressed. The destabilizing effects of the spandrel columns, if used, are not addressed in the provisions. Also, the moment magnification equations do not take into account the reduction of stiffness in steel structures at the ultimate limit state.

C16.2.1

Second order analysis is described under “Direct Analysis Method” in Appendix 7 of the AISC (2005), where guidelines on accounting for imperfections and residual stresses can also be found (D04). It is noted that the AISC Specifications do not include the effects of camber and bowing.

C16.2.2

Construction of long span bridges is often guided by a special set of requirements, including frequent survey of the structure in each partially constructed stage, specified tolerance on deviations from the target alignment, deflection and grades, etc. Construction tolerances that reflect project specific expectations on workmanship should be a good basis for establishing effects on geometric imperfection.

When analyzing a tied arch, it has been noted that the suspenders would provide a certain degree of stabilizing effect against the lateral buckling of the arch.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

C16.2.3

There has not been a universally accepted approach in assessing global stability of the cable- stayed deck because the buckling modes are not easily defined in the modern cable-stayed decks supported by closely spaced stay cables. While designers continue to use the term “buckling”, many believe that modern cable-stayed decks do not actually buckle, but simply fail should they be subjected to increased overloading. Second order analysis with defined initial imperfections eliminates this problem.

Technical writings that articulated this matter with differing approaches include Tang (D28), Taylor et al. (D29) and Ren (D30).

• Tang (D28) proposed a procedure based on the energy method for determining the F.S. against buckling. The analysis was elastic and linear in nature, intended to estimate a theoretical buckling load at which the deck would see significant increases in deflection. This energy method may be extended to consider local buckling by providing additional nodes in the analytical model to simulate local areas with local deformation curves. Deflection curves based on various loadings on the deck in question were used to approximate the mode shape corresponding to the lowest critical load. The energy method was proposed as a general analytical procedure because it would not require pre-screening for vulnerable points on the deck nor did it require that structure specific loading scenarios be applied specifically to bring about failures at the target points of vulnerability. In design, Tang suggested the application of the computed axial force and a bending moment magnified by the AASHTO moment magnification factor.

• Taylor et al. (D29) and Ren (D30) did not present a general procedure for calculating a factor of safety against buckling. Instead, suggested nonlinear analyses that included geometric and material nonlinearities were used to assess deck buckling.

o Taylor et al. (D29) highlighted a number of potential vulnerabilities associated with recent trends in designing cable-stayed decks. They also reviewed cases where the main spans are loaded by a multiple of the dead load pattern until a target location of the deck would fail. They denoted the ratio of the selected load increment as the Buckling Load Factor or KDL. They did not recommend a minimum value for KDL but noted that some existing structures may have values as low as 1.3 for KDL.

Note: KDL is defined as the multiple, K, times dead load placed on the main span of a 3-span cable-stayed bridge that would cause buckling to occur, while the side spans remain loaded with 1.0 times dead loading.

o Ren (D30) reviewed a structure with a 605 meter main span. Based on the limit point instability concept, Ren traced the modes of failure of the cable-stayed deck to characterize the ultimate behavior of the deck under live load, and concluded that material nonlinearity is a significant aspect that should not be neglected when conducting nonlinear analysis. Ren did not distinguish the behavior of a cable-stayed deck under dead load vs. live loading.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

C16.2.4

The listed provisions in AASHTO LRFD point to the complex stress distribution in the orthotropic deck.

It is common practice that assumptions regarding the stability of orthotropic girders, such as effective flange widths of orthotropic deck, be verified by refined analyses.

C17.1

Field measurements have shown considerable scatter in the estimation of structural damping. Davenport (D32), ESDU 87035 (D33), and current design practice in North America show a typical range as follows:

• steel bridges – 0.3% to 1.0% • composite steel/concrete bridges – 0.5% to 1.5% • concrete bridges including substructures – 0.7% to 2.0%

These lower structural damping ratios can be applied to wind-related problems of low amplitudes, such as vortex shedding, where the onset of aerodynamic instability may be accompanied by relatively minor structural deformations.

Outside of the US, damping as low as 0.2% has been measured in steel box girder bridges. Designer should adopt damping values that are most representative of the bridge type under consideration.

Damping ratios recommended for seismic analyses are higher than the damping ratios recommended in AASHTO LRFD Article C4.7.1.4. For example, Caltrans Seismic Design Criteria, Version 1.4, June 2006, Section 2.1.5 recommends 5% damping for concrete bridges, while AASHTO LRFD suggests 2%, in lieu of field measurements.

C17.2

It is prudent to determine the need to conduct aerodynamic studies early in the design. Bridge designers should work closely with wind consultants to evaluate wind sensitivity.

Although a bridge with Sw values higher than the recommended limit can be considered less sensitive to wind, the given criterion must be applied with caution for special cases involving unusual bridge configurations.

The depth of the deck should include the solid height of the traffic barrier.

Depending on the selected erection scheme, long span bridge construction is typically more sensitive to wind during construction.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

Section B, Article 3.1.2 presents a characterization of the long span bridge in terms of flexibility and human comfort. This characterization is not strictly a set of criteria for aerodynamic stability but provides an indication of where aerodynamic stability of a flexible bridge may become a design issue.

The following are typical analytical steps taken during an aerodynamic investigation:

1. Structural dynamic analyses 2. Flutter stability analysis 3. Buffeting response analysis

Structural dynamic analysis is carried out for estimation of the frequencies and modes required for aerodynamic analyses. Typically, all modes up to about 2.0 Hz are used for stability and response estimates. Sectional model tests typically provide conservative estimates of flutter speed. Some of the sources of conservatism are in the omission of added mass and aerodynamic damping effects from other bridge elements such as towers, main cables and deck coupling with lateral motions. These effects are most notable on suspension bridges. If cables or towers of unusual configurations exhibit negative aerodynamic damping, then the full scale flutter speed may become lower than that measured of sectional models.” 3-D flutter analysis is considered to be the state-of-the-art method for theoretical estimation of flutter speed (D34).

On long-span suspension bridges, where modes tend to be coupled and closely spaced in frequency, a sectional model test alone may not be able to predict the lowest flutter speed.

The critical speed of flutter may be best predicted via aeroelastic model tests and/or 3-D flutter analysis. This analysis is based on the original theoretical multi-mode flutter analysis extended to include as fully as possible the actual 3-D behavior of bridges.

Buffeting analysis based on quasi-static theory is a universally recognized method for predicting bridge responses to turbulent winds. Among the accepted theoretical methods are those formulated by Davenport (D35) and Irwin (D36). The wind consultant should demonstrate the validity of the applied methods against aeroelastic model tests and/or full scale measurements.

It is recommended that the buffeting analysis incorporate as many 3-D loading effects as possible, including but not limited to loads on the towers, main cables and deck.

3-D finite element models used for buffeting analysis must have a sufficient degree of detail to ensure proper modal estimates for modes up to about 2.0 Hz.

When analyzing response and stability of cables, these may need to be modeled with many elements to cover higher cable modes. Special attention should be given when modeling boundary details such as the cable anchors, internal and/or external damper connections, and cross-cable or hanger collars.

If aerodynamic countermeasures are proposed to mitigate vibrations, the desired effect should be confirmed by experimental studies.

Parametric and external force excitation should be considered.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

Effects that may alter the surface roughness and shape of flexible elements, such as accumulation of ice on cables, should be considered in the design.

Wind load on towers and arch ribs should take into account dynamic and static wind pressure, buffeting, and vortex shedding. Shading effects on the leeward elements may be considered if verified by tests.

Refer to Sections B (Article 2.5) and C (Article 11.0) of these Guidelines for discussions on additional topics relating to wind effects and wind loads, including preliminary determination of the sensitivity of long span bridges to wind effects, site-specific wind climate studies, determination of design wind loads (including buffeting effects) and physical (wind tunnel) testing requirements.

C17.3

As used herein, the term “substructure” refers to the towers, pylons, piers or bents that support the bridge, including their foundations.

AASHTO LRFD Article 4.7.4.3 describes the following methods of seismic analysis for bridges: • uniform load elastic method (UL) • single-mode elastic method (SM) • multimode elastic method (MM) • time history method (TH)

Long span bridges are generally critical bridges and the geometric and structural configuration of long span bridges generally results in “irregular” geometry, i.e. the ratio of adjacent span lengths and the relative stiffness of the various substructure units are generally non-uniform. Therefore, the uniform load elastic method (UL) and single-mode elastic method (SM) are not appropriate for seismic analysis of long span bridges.

The multimode elastic analysis method, time history analysis method and inelastic static (pushover) analysis method are described in Articles 17.3.1, 17.3.2 and 17.3.3 of these Guidelines, respectively.

Bridge owners may consider requiring performance objectives on a case by case basis. For example, these objectives may include a stipulation that selected bridges be designed and detailed to ensure immediate limited usage by emergency vehicles following a design level seismic event, as opposed to a non-collapse criterion.

C17.3.1

Multimode response spectrum analysis is described in the following references:

• AASHTO Guide Specifications for LRFD Seismic Bridge Design, Article 5.4.3 – Procedure 2: Elastic Dynamic Analysis (EDA) (D01) • AASHTO Standard Specifications Division 1A – Seismic Design, Section 4.5, Multi-Mode Spectral Analysis Method – Procedure 3 and Section C4.5 in the Commentary to the 1998 Interim Specifications (D03) • Clough and Penzien (D10) • ATC-32, Articles 3.21.6 and C3.21.6 – Elastic Dynamic Analysis (D05)

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

• Buckle et al., Section 5.4.2.2 – Multi-Mode Spectral Analysis Method (D07) • Ho et al., Section 4.2.1.1 – Conventional Response Spectrum Analysis (RSA) (D14) • MCEER/ATC-49, Sections 5.4.2.3 and C5.4.2.3 (Commentary) – Multi-Mode Dynamic Analysis Method (D21) • Caltrans Seismic Design Criteria, Version 1.6, Caltrans, November 2010 (D09)

Also refer to Section C of these Guidelines, in particular the following articles:

• Article 12.3.1 – Design Response Spectra Based on Site-Specific Procedure • Article 12.3.2 – Design Response Spectra Based on General Procedure • Article 12.5 – Combination of Seismic Force Effects

Multi-mode response spectrum analysis can, in general, overestimate relative movements across expansion joints. It may be cost-effective to perform nonlinear time history analysis in order to arrive at more realistic expansion joint movements and possibly smaller expansion joints.

The multi-mode spectral analysis method considers coupling between structure responses in the three global directions, and the response of the structure in the higher modes of vibration, which are neglected in the simpler analysis methods (i.e. the uniform load method or single mode spectral method). In a multi-mode spectral analysis, the total structure response is determined by combining the maximum response of each vibration mode. The multi-mode spectral method may be used for structures with irregular geometry which induces coupling in the three coordinate directions within individual vibrations modes.

For multi-mode spectral analysis, seismic response should be determined as structure displacements and individual member forces using dynamic analysis techniques considering stiffness, damping and mass properties of the structure and soil. Modal spectral analysis based on the application of a response spectrum of ground acceleration to a lumped-mass space frame or finite element model shall be used. The analysis shall be performed independently in each of three orthogonal directions (longitudinal, transverse and vertical).

The number of degrees of freedom and number of modes considered in the analysis shall be sufficient to include all critical response modes including modes of vibration associated with the self-response of foundations. This requirement may generally be considered to be satisfied if the effective mass in the model (mass participation factor) is at least 90% of the total structure mass, and the effective mass for each segment or span is at least 90% of the total mass of that segment or span.

The earthquake response of structures with intermediate hinges, expansion joints or multi- bridge systems shall be considered in the analysis. This may be accomplished by evaluating several linear models with boundary conditions which bound the actual condition (e.g. stand- alone, compression and tension models), taking the design actions to be the governing force and displacement results from the various models analyzed.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

C17.3.2

Time history analysis (elastic and inelastic) is described in the following references: • AASHTO Guide Specifications for LRFD Seismic Bridge Design, Article 5.4.4 – Procedure 3: Nonlinear Time History Method (D01). • Clough and Penzien (D10) • ATC-32, Articles 3.21.8 and C3.21.8 – Inelastic Dynamic Analysis (D05) • Buckle et al., Section 5.4.2.3 – Elastic Time History Method and Section 5.7 – Method E: Nonlinear Dynamic Procedure (D07) • Ho et al., Section 4.2.1.2 – Time History Analyses – Elastic and Inelastic (D14)

Also refer to Section C of these Guidelines, in particular the following articles:

• Article 12.3.3 – Acceleration Time Histories • Article 12.5 – Combination of Seismic Force Effects In some cases, time history analysis can produce lower force and/or displacement demands than multimode response spectrum analysis. For instance, relative displacements across an expansion joint computed from an elastic or inelastic time history analysis may be smaller than those computed based on multimode response spectrum analysis.

For inelastic time history analysis, seismic response is determined as structure displacements and individual member forces using dynamic analysis techniques that consider geometric and material nonlinearities; damping; and mass properties of the structure and soil. Soil-structure interaction should be considered using non-linear springs to reflect the properties of the surrounding soil (see Article 15.4, above). Where significant, radiation damping effects and the elasto-dynamic behavior of the far-field soil outside the nonlinear soil region adjacent to the foundation should also be included in the analysis.

A response-history analysis of a lumped-mass space frame or finite element model of the structure should be used. The number of degrees of freedom considered in the analysis should be sufficient to excite all critical response modes.

The structural model should include the effects of concrete cracking, yielding of reinforcing steel and other material nonlinearities on the stiffness of members, and shall include the restraint of the surrounding soil. Inelastic response characteristics of the analytical model should be justified by experimental evidence; see Buckle et al. (D07). Viscous damping equal to 5% of the critical value or less shall be assumed for all critical response modes in addition to inelastic energy dissipation, except that higher viscous damping values are allowed where justified by experimental evidence and analysis.

Inertial mass shall consider dead loads only (any inertia effects from traffic live load are typically neglected). Gravity loads shall include all dead loads and the applicable portion of the design live load. Nonlinear (P-Delta) effects of gravity loads acting through lateral displacements and geometric stiffness (stress-stiffening) effects for cable elements shall be included in the analysis.

When using linear or nonlinear time history analysis with spatially varying ground motion inputs, an analysis using uniform ground motions shall also be performed. In some cases, uniform ground motions can produce more critical results than spatially varying ground motions.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

C17.3.3

Inelastic static (pushover) analysis is described in the following references: • Priestley et al. (D26) • ATC-32, Articles 3.21.7 and C3.21.7 – Inelastic Static Analysis (D05) • Buckle et al., Section 5.6 – Method D2: Structure Capacity/Demand (Pushover) Method (D07) • Ho et al., Section 4.2.2.1 – Static Pushover Analyses (D14) • MCEER/ATC-49, Sections 5.4.3 and C5.4.3 (Commentary) – Seismic Displacement Capacity Verification and Section (D21) • SCDOT Seismic Design Specifications for Highway Bridges, Section 6 – Seismic Displacement Capacity (D37) • Caltrans Seismic Design Criteria, Section 3 – Capacities of Structure Components, Section 4 – Demand vs. Capacity and Section 5 - Analysis (D09) • Ho, et. al., “Seismic Retrofitting Guidelines for Complex Steel Truss Highway Bridges”, Chapter 4 - Structural Analysis (Ref. D14)

Pushover analysis is best-suited for piers that have a concentrated location of mass, e.g. “typical” approach span pier configurations where the mass is mostly concentrated at deck level.

For cable-stayed bridge towers, the distribution of cable forces along the tower head can make it difficult to arrive at meaningful results using pushover analysis.

In the inelastic static (push-over) analysis, seismic responses are determined as local displacements, individual member deformations, and individual member forces using inelastic static analysis techniques, considering nonlinear stiffness properties of the structure and soil.

A step-by-step lateral displacement response analysis of a two-dimensional or three-dimensional model of the structure should be performed. Gravity loads should include all dead loads and the applicable portion of the design live load (see Section C, Article 8.1.4 of these Guidelines). Seismic loads may be assumed to act in one direction only. Nonlinear (P-Delta) effects of gravity loads acting through lateral displacements should be included in the analysis.

The structural model should include the effects of concrete cracking and other material nonlinearities on the stiffness of members, and should include the restraint of the surrounding soil. Inelastic response characteristics of the analytical model should be justified by experimental evidence.

For simple structures, i.e. single-level structures with single column bents or 2-dimensional multi-column frames, which can be adequately modeled as equivalent single-degree-of-freedom systems for horizontal motion, the center of mass of the superstructure should be displaced in steps until system failure mechanisms develop.

Local displacements and individual member deformations should be monitored at each step.

When the component model includes degradation of strength with increasing deformation, or when P-Delta effects counteract strain hardening, then the pushover analysis will show an increasing load with displacement to a maximum load and then the load may decrease with increasing displacement. This behavior can lead to large deformations and concentration of damage in degrading components. The reduction in lateral load may be large before a

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary D

component reaches its deformation capacity. Hence, the requirement to limit the maximum displacement to a point at which 80% of the peak lateral load is reached, and not permit further reduction in lateral load capacity.

Mander’s stress-strain model (D19) may be used to determine the ultimate strain capacity for confined concrete.

C17.3.4

Section 8 of the AASHTO Guide Specifications for LRFD Seismic Bridge Design (D01) presents guidelines for performing moment-curvature analysis.

C17.3.5

Dynamic interaction effects include inertia and damping effects, and depend on the shape and size of the submerged pier or pile, and the depth of water. The interaction with surrounding water tends to increase the fundamental period of vibration of the structure.

As a minimum, the hydrodynamic inertia effects should be represented in the analysis using the “added mass” approach, where the added mass is defined as a fraction of the mass of the displaced water. This should be done based on work by Goyal, Liaw and Chopra (D11 and D12).

Hydrodynamic damping effects may be neglected in the analysis, as this is conservative.

C17.3.6

The design for the Functional Evaluation Earthquake level earthquake should not count on any stiffness contribution from the abutment backwall. For the Safety Evaluation Earthquake level earthquake, the stiffness contribution from the abutment backwall (compression only) may be included in the analytical model, provided that the abutment backwall is suitably designed and detailed.

C17.3.7

Where nonlinear inelastic time-history analysis is performed, displacements calculated from the inelastic analysis may be used directly in the design.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary E

SECTION E COMMENTARY – CONCRETE STRUCTURES

C18.1.1

As long span bridges are generally constructed segmentally and/or in stages, it is appropriate to conduct more accurate estimates of the effects to include:

• Materials used • Principal dimensions of structural elements • Site conditions • Construction methods • Age of concrete at the time of load applications

Ignoring the effect of reinforcement may result in overestimating the effects of creep and shrinkage which is not necessarily conservative in design. Restraining forces at a heavily reinforced framed connection, for example, may be underestimated if the effects of the reinforcement are neglected. Appendix C of the British Code BS5400, Part 4, 1990 (E05) provides an example and discussions on accounting for the effects of reinforcement.

Given the variability of creep and shrinkage, it is prudent to be somewhat conservative in estimating the effects and making provisions in design. More rigorous analyses are encouraged to assess the sensitivity of structures to account for this variability.

The plans should show all the assumptions used in the creep and shrinkage model by the Designer. The Designer should consider specifying shrinkage and creep limits as part of the mix design for the project. Shrinkage can usually be easily controlled by aggregate sources and shrinkage reducing admixtures along with low water/cement ratios.

C18.1.2

Since the design service life of long span bridges tends to be longer than that of conventional structures, additional precautions should be taken in selecting the design mixture for construction. In addition to specifying the desired concrete properties, other measurable factors favorable to durability, including lower permeability, higher electrical resistivity, lower shrinkage, and lower cracking potential, should be specified.

To complement more thorough design practice, tighter field quality control and validation is generally required to ensure the desired outcome.

Various analytical models are available for predicting the design service life of concrete structures. fib Bulletin 34 – Model Code for Service Life Design (2006) is a resource that systematically outlines design procedures for enhanced durability (E06).

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary E

C19.1

The CEB-FIP model code provides guidance in addressing confidence limits within the mean values for creep and shrinkage coefficients. For reference, see CEB-FIP 1990, Articles 2.1.6.4.3 and 2.1.6.4.4 (E03).

In general, the design of long span bridges should be validated against creep and shrinkage coefficients based on the constructor proposed concrete mixes and materials.

Inelastic analysis according to Section D of these Guidelines may be used to determine the effects of time dependent deformations, restraint and force redistribution, as long as strain compatibility analyses demonstrate that the structural elements and connections provide the necessary ductility.

C19.1.1

For important structures, some bridge owners have extended the zero tensile stress limitation to permanent loads plus live load for non-replaceable elements.

Non-replaceable concrete superstructure elements are structural elements that cannot be designed for replacement without significant initial construction cost and/or subsequent rehabilitation cost. The reinforced concrete deck of a cable-stayed bridge and/or prestressed (pretensioned or post-tensioned) concrete decks that are under significant compressive stress due to permanent loads are generally considered non-replaceable elements.

To achieve the design life of non-replaceable concrete elements, stress limitations stated herein and other limitations in the AASHTO LRFD Specifications should be applied in conjunction with other forms of corrosion protection including the use of coated or stainless mild reinforcement; low permeability overlay; corrosion protected tendons; and inspection, monitoring and maintenance programs.

C20.1

The shear provisions in the AASHTO LRFD Bridge Design Specifications are not applicable for sections that are expected to accommodate a significant amount of plastic deformation. The concrete shear strength within the plastic hinge region degrades as the ductility demand increases but is improved with increasing transverse confinement.

Sections 7 and 8 of the Caltrans Seismic Design Criteria (E10) also present design and detailing procedures for ductile concrete components.

Particular attention should be paid to the design and detailing of moment-resisting connections such as knee joints, integral connections of pier columns to superstructure, connection of pier columns to footings and the connection of drilled shafts and piles to pile caps.

This stipulation is recommended because long span bridges are generally categorized as important structures. The most significant effect of this provision would be to invoke the transverse (confinement) steel requirements for SDC B for pier columns. In addition to the ability to resist seismic loads, the increased confinement steel will reduce the tendency of shrinkage cracking in columns and piers, and increase the structure’s resistance to blasts or other threats.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary E

Note that the designation of Seismic Design Category (SDC) B in the AASHTO Guide specifications for LRFD Seismic Bridge Design is equivalent to Seismic Zone 2 in the AASHTO LRFD Bridge Design Specifications.

C20.1.1

Hollow concrete sections are economical for the substructure of long span designs or large projects, where special design and validation by testing may be warranted. One such case was the Taiwan high speed rail project; see Ductility Design of Hollow Columns – by Mo and Nien (E07).

Hollow concrete piers with highly confined corner elements (corners confined with hoops) have been proven to be very ductile and have been used in high seismic regions in California; see Seismic Performance of Hollow Rectangular Piers - by Hines, Seible, and Priestley (E08).

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

SECTION F COMMENTARY – STEEL STRUCTURES

C22.1

Parallel-wire cables are typically erected by either the aerial spinning method or the prefabricated parallel wire strand method.

No. 6 galvanized steel wires are the most commonly used bridge wires for main cables in the U.S. These wires are 5 mm in diameter (0.196”) with zinc coating and 4.9 mm (0.192”) without the zinc.

Strength for the 5mm (0.196”) diameter wire may vary from 1517 MPa (220 ksi) to 1863 MPA (270 ksi).

The modulus of elasticity of parallel wire is approximately 29,000 ksi for the net section of the wire (without the zinc) and typically reduced to 28,500 ksi when using nominal diameter of wire. The zinc coating does not add to the strength of the wire; however for acceptance criteria it is customary to use the nominal diameter of the wire, and in order not to make errors in the calculations for elongation, the adjusted value 28,500 ksi is used for the modulus of elasticity.

Other materials may be considered. The stated Guidelines are intended to serve as the baseline for consideration of proposed alternatives.

The designer should always verify the availability of materials in the local area.

C22.1.1.3

The maximum primary axial fatigue stress range of 24 ksi for hot galvanized bridge wire is based on test data performed at John A. Roebling’s Sons Company (Re: H.A. Godfrey). This also fits into fatigue category A. Fatigue in new cable wires is generally not an issue. This is due to the fact that the maximum service load stress in most bridges is kept under 100 ksi (based on fu = 220 ksi) while the live load stress in the cable is usually less than 25% of the total stress. As a result the fatigue stresses are kept well below the 24 ksi limit.

In recent evaluation of remaining strength of degraded suspension bridge cables by Mahmoud (F02), it has been shown that cracking of the cable wire has driven degradation and reduced load carrying capacity of the cable.

C22.1.1.4

Yield strength may be estimated at 0.75 fu in the absence of material data.

φ=0.80 has been derived so that the total service load stress in the cable wire does not exceed the elastic range.

It is important to note that there are locations along the cable where stress rise occurs. These locations include strand shoes, splay castings, splay saddles, anchorage and tower saddles, and cable bands. The resistance factor value applied to the maximum uniform tension is expected to produce a reasonable cable size that will accommodate stress rises associated with

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

thoughtful detailing. It is important that the designer check and consider the total cable to account for the maximum stress anywhere along the full length of the cable.

C22.1.2

In the absence of project specific criteria, the following additional parameters should be specified:

• Design Elastic modulus – 29,000 ksi • Zinc Coating – Class A • Elongation in 10” – 4% minimum • Wire Diameter (including, zinc coating) – 0.196 inch • Reduction of Area – 35% min.

C22.2

Saddles are typically made of cast steel (see AASHTO LRFD 6.4.7).

Saddles for suspension cables are geometrically designed to minimize bending stresses, slip and fretting in the cable wires. Due to the global geometry of the main cable, the maximum axial service load stresses are typically found at the tower-side span interface. Additional bending stresses make the cable at the saddle one of the highest stressed areas on the bridge. While existing bridges provide a wealth of information as references for new designs, fatigue testing should be conducted during design to address variance from past design and detailing practice.

At the saddles, both at the towers and at the anchorages, the main cable cannot be compacted nor wrapped; as a result, the cable retains a hexagonal shape inside the saddle.

The saddles are designed for all group loads and limit states as specified in AASHTO LRFD. The global analyses of the bridge are performed through computer models covering the entire suspension bridge anchorage to anchorage. Once the global forces acting on the saddle have been determined (including vertical/inclined and lateral forces), the saddle design will be carried out like any other structural steel component.

COMPACTED CABLE / UNCOMPACTED CABLE Cross Section at Saddle

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

C22.2.1

Unbalanced horizontal force

Forces acting on tower saddle

C22.2.3

Anchorage saddles may bear directly on the anchorage structure or may be supported by a bent (as illustrated below).

Over time the anchorage saddle rollers rust and freeze in place, restricting saddle movement. Designers should account for such changing conditions and design for the appropriate forces acting on the saddle anchor bolts and the supporting structure.

Typical Anchorage Saddle Balance of forces at anchorage saddle Saddle on rollers

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

Typical Anchorage Saddle Bent Balance of forces at anchorage saddle bent Fixed Saddle

Anchorage saddles or cable bent saddles are typically designed to bear on rollers to accommodate the change in temperature and the cable tension adjustment during erection.

Similar to the tower saddle, the anchorage saddle carries the radial components of the main cable tension on both sides of the saddle as well as the transverse loads associated with eccentric live, wind, and seismic loads.

Over time the anchorage saddle rollers rust and freeze in place, restricting saddle movement. Designers should account for such conditions and design for the appropriate forces acting on the saddle anchor bolts or the saddle bent.

C23.1

Twisted wire ropes are commonly used in the suspenders of suspension bridges. The wire rope construction is usually 6 strands with an independent wire rope core (IWRC). The number of wires in each strand varies and depends on the design load of the rope. The diameter of a suspender rope is nominal and measured along the larger width.

Parallel wire strands are commonly used outside of the U.S., although there are currently no ASTM standards for this type of strands. Should circumstances call for the procurement of these strands from overseas sources, their mechanical properties should be established by the Designer on a case by case basis.

Cross Section of a Typical Suspender Rope

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

C23.1.1.2

Suspenders are typically designed for a higher safety factor than the main cable, usually 4 against the breaking load. This is generally due to the fact that the suspender rope wires are not protected by wrapping wire, as in the case of the main cable, and they are susceptible to corrosion, especially at their end sockets.

C23.1.1.4

Wire ropes commonly used as suspenders are based on testing.

φ = 0.50 has been derived with the goal of keeping the suspender in elastic range under total service load stress.

C23.1.2

The basic strength of the wires in a suspender rope is the same as in the parallel wire strand and typically specified as 1,517 MPa (220 ksi). However, due to the twisting of the wires, the strength of the rope cannot be defined in the same manner. Suspender ropes with twisted wires are specified by identifying the ultimate (breaking) load of the rope, by the amount of stretching at failure, and by the modulus of elasticity after pre-stretching. The loss of rope efficiency due to bend over cable bands should be accounted for in the design.

The Designer should always verify the availability of materials and their corresponding material properties in the local area.

C23.2

Suspenders are typically fabricated with a socket on each end.

The suspender socket consists of two parts; the first part is a cast iron wall and the second is the filling material. Both parts shall be designed for the same limit state criteria.

Hot pour zinc has been used in the past to bond the wire rope to the socket, but the excessive heat tends to degrade the wire strength at the interface. Resin is currently the preferred material. It should be noted that resin is less fire resistant than zinc. Test data vary by manufacturer and are usually available for the designer’s reference.

The size and details of the suspender sockets vary from fabricator to fabricator, but the overall behavior and function are the same.

C23.2.1.3

Threaded details have been used though not common.

C23.2.1.4

Socket design relies on both friction and radial bearing forces. The design is usually performed by vendors who often elect to use conservative assumptions. The proof of socket design has historically been based on testing.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

Sockets for suspenders are typically designed for the ultimate capacity of the suspender. In order to ensure that the connection is stronger than the member, a factor of 1.10 is introduced to the equation.

C23.2.3

Construction specifications should generally require no failure of wires inside the socket during prototype (system prequalification) or project specific testing.

C23.3

Cable bands are subject to bending and tensile stresses due to the bolt tightening forces. The specified cast steel (AASHTO LRFD 6.4.7.) needs to meet the high demand of the clamping forces.

Cable bands are the connections of the suspenders to the main cable. The bands are two halves traditionally made of cast steel connected by a number of high strength bolts which are tensioned to provide sufficient friction between the band inner surface and the main cable to resist the suspender load component along the main cable slope.

C23.3.2.4

Psu is a factored load obtained by analysis and not a resistance value.

While the resistance factor for bolts in tension is 0.80 as defined in AASHTO, it is recommended to use a smaller value of 0.4 for cable clamps. This will reduce the required bolt tension and the corresponding loss of clamping force due to creep effects.

C23.3.3

Design plans typically provide instructions for applying bolt tension in several passes.

Cable band bolt tension is subject to creep as the cable band acts as a compactor for the main cable wire, reducing the over time and in the process losing the bolt tension. Overtensioning the bolts will result in a faster creep effect.

C24.0

Some earlier cable-stayed bridges in the U.S. used parallel wire cables. Recently, stay cables utilizing prestressing steel have been the predominant choice in the construction of cable-stayed bridges in the U.S. Nevertheless, for very long span cable-stayed bridges, the use of parallel wire stay cables may be a cost-effective alternative. The reduction in wind drag on the more compact parallel wire cables is seen as a potential source of saving on substructure and foundation design.

C24.1.1

Both stress-relieved (normal-relaxation) and low-relaxation wire used in prestressed concrete applications are acceptable for stay cable applications since the creep (relaxation) in the stays is negligible under working stress. However, these Guidelines call for low-relaxation grade because the low-relaxation treatment process has proven to effect consistent tensile strength

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

throughout the production length of the strand, and result in better straightness (cast) for easy cable fabrication.

Outside the U.S., hot dip galvanized wire has been used for a number of cable-stayed bridges without low-relaxation treatment. At the present time, there is no ASTM specification for hot dip galvanized wire that is suitable for stay cable applications. Among the requirements for this type of wire are tensile strength, load at 1% extension, elongation, coating weight, straightness (cast) and fatigue resistance.

C24.1.2

The PTI Recommendations (F03) limit the frequency of occurrence of welds in wire rod at no more than one weld per each 500 kg of rod.

C24.1.3

A specialist stay supplier/contractor is typically prequalified and required to submit a stay system design prior to acceptance testing of the stay cable.

C24.1.4

The PTI Recommendations (F03) provide design and testing guidelines that reflect industry practice. Specific concerns in designing saddles include:

• Reduction in fatigue strength in the presence of lateral pressure • Notching and fretting associated with differential strains in strand bundles • Deterioration of force transfer between the stay and the saddle

Tests on bare post-tensioning strands in curved ducts performed at the University of Texas (F11) provide basic information regarding fretting fatigue in strand materials commonly used for stay cables.

C 24.2.3

The stress range available for design may be as low as 8 ksi for strands and 10.5 ksi for wire according to the PTI Recommendations. PTI Article 5.3.5 provides these lower bound values to account for heavy volume bridges.

C 24.2.4

Bending stresses may be calculated from the stay curvature (1/χ)Er, where r is the relevant distance of the strand extreme fiber to the neutral axis for bending and χ is the local radius of curvature in the cable at the local detail being examined.

Note:

The flexural stiffness of the stay cable and the relevant distance of the extreme fiber to the neutral axis for bending depend on the actual type of stay cable utilized. For individually protected main tension element (MTE) without grouting of the stay cable, the stiffness may be assumed as the sum of the stiffness of each individual MTE. If the MTE is a wire or bar, r may

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

be assumed as 50% of the nominal diameter of the MTE. If the MTE is a strand, r may be assumed to be the center-to-center distance between the king wire and an exterior wire. For existing stay cables with MTE that are fully grouted, the stiffness may be calculated for the composite steel-grout section with the nominal cable diameter, and r may be assumed as 50% of the nominal cable diameter.

C 24.3

Phi Factor for a combined axial plus bending condition should be determined for each stay cable and not intended to vary by loading cases.

MUTS is defined in Article C24.4-1.

The development of the variable phi factor was inspired by the works of Ogawa and Kasuga (F10).

Maximum resistance factor for extreme limit state is 0.95 to reflect minimum strength requirements for qualification of anchorage design.

At the fatigue limit state, elastic bending and axial stresses should be added directly.

C24.4

Stay cables are complex engineered products. Design and construction standards have been developed over time by stay cable suppliers with input and validation from academia, testing laboratories, designers and owners through participation in technical committees. The performance expectations associated with these design Guidelines are based on a state-of- practice that encompasses material specifications, component and system testing, design, quality control and construction practices that must meet or exceed minimum requirements. This state-of-practice is documented in various publications including the PTI Recommendations, the SETRA/CIP (F04), and the FIB Recommendations (F05).

The state of practice calls for the main tensile element to be protected against corrosion by two or more nested barriers that are individually protected against corrosion.

C24.4.1

Ductility requirements – Necking and bend tests are industry standards imposed in addition to the ASTM A421/A421M requirements.

PTI and FIB provide alternative numbers of test cycles with corresponding stress ranges.

The term Minimum Ultimate Tensile Strength (MUTS) is equal to f’s x ‘nominal area’ of steel.

C24.4.2

The One-Pin Test for ductility requirements is in addition to the requirements of ASTM A416/A416M, and defined in the PTI Recommendations.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

C24.4.3

PTI and FIB provide alternative numbers of test cycles with corresponding stress ranges. See also Article 24.2.1 of these Guidelines.

Acceptance may, at the discretion of the Engineer, be based on results of previous testing of similar applications. Generally, this would mean that the stay hardware is the same as the previous application, and that the previous and present applications share similar fatigue characteristics.

C24.4.4

Standards for acceptance of previous test results would be similar to those for stay anchorages.

The PTI Recommendations provide general testing guidelines that reflect industry practice.

C 25.1.1

LRFD provisions such as those in LRFD Articles 6.9.4.2, 6.11.8.2, and 6.1.4.4 may be considered pertinent for arch rib design, but tend to produce inconsistent design results. The equations proposed herein aim at achieving a uniform approach.

The current provisions in the AASHTO bridge design specifications ignore the interaction of shear and normal stresses in web panels. The only exceptions are the provisions for the bottom flange of box girders, 6.11.8.2.2, where the interaction of torsional shear and normal stress from the bending of the box girder is considered. Designers have often referred alternatively to Report FHWA-TS-80-205 (Proposed Design Specifications for Steel Box Girder Bridge, 1980) (F06) where consideration of the interaction is warranted.

The TS-80-205 approach is similar to the interaction buckling criterion of Article 9.11.4.4 of the BS-5400 code (F07). While these provisions are LFD (F08) based, they can be converted to LRFD by applying the LRFD loading modified with the LRFD load factors to calculate the factored stresses in the stiffened plate. Correspondingly, the resistance factors in LRFD Article 6.5.4.2 may be applied to the critical shear, bending and axial stresses, hence,

Where for strength limit state

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary F

Where for extreme event limit state

C25.1.1.1

Bending and axial compression are combined in one term in this quadratic interaction equation, where the strength of the element is based upon buckling capacity without direct consideration of post-buckling shear and axial effects.

C25.1.1.1.1

fb fa fc= fb + fa

=

f1= fb - fa

Bending Axial Combined

fc= fb + fa

f1= fb - fa

Combined

C25.2

Though encountered in short and medium span designs, deformation problems are usually exacerbated by the increased movement associated with a long span design.

C25.2.1

Article 5.9 of the PTI Recommendations indicates a minimum lateral load of 2.5% of the maximum static stay force for strength limit states, and the greater of 4% of the cable fatigue load or 1.5% of the maximum live load stay force for fatigue limit state.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

SECTION G COMMENTARY – FOUNDATIONS

C 26.0

Procedures and provisions specified in this Section are intended for arch and cable-supported bridges with long spans that require foundations that are larger in scale, support greater loads, and result in greater stress on the supporting ground than conventional bridges.

C 26.1

The main purpose of the investigation program is to sufficiently define the subsurface conditions for an appropriate and economical design of the bridge foundations to limit the risk during construction and to predict the long term performance of the completed structure.

For river crossings, the larger size foundations often result in greater depth of scour. Therefore, the investigation program must consider the typically greater foundation size and loading for long span bridges, and the potential for greater scour at water crossings.

The investigation program is typically performed in stages to provide initial data for defining subsurface conditions sufficient for performing alignment selection and bridge type selection studies; then additional data is obtained as the locations of the foundation elements are defined during preliminary design, and confirmatory borings are performed during the final design stage. A staged investigation can provide the information necessary for each phase of the project without committing the full budget for the investigation before locations and depths of the various foundation elements are sufficiently defined. A typical program for a long span bridge may include some or all of the following stages: • Conceptual Design Investigations • Preliminary Design Investigations • Final Design Investigations • Construction Control Investigations.

The conceptual design stage would begin with a desktop geologic study during the initial site studies. A literature search including collection of available existing borings and other subsurface data should be performed at this stage. The purpose of the initial geologic studies is to identify at an early stage any geologic hazard that may significantly impact the bridge location, type, and span arrangement. The conceptual design stage investigation may also include a limited boring and testing program for general assessment of subsurface conditions, particularly where there is little or no existing information for the project. During the preliminary design stage, geophysical and bathymetric surveys supplemented by geotechnical borings may be performed to advance the conceptual design to preliminary design. During the final design stage, additional geotechnical investigations may be performed to define subsurface conditions at the finalized foundation locations. During the construction control stage, confirmatory borings may be performed, if necessary. This may also include supplementary borings and testing to further evaluate unexpected site conditions encountered during construction.

A staged investigation program would be applicable to both a design-bid-build type project and a design-build project. However, the purpose and scope of the preliminary design investigation stage would differ for these two project delivery approaches. In a design-bid-build project, where the bridge type, arrangement and foundation elements are approximately defined during

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

preliminary design, the preliminary investigation program typically includes borings at the defined foundation locations. However, in a design-build contract, the preliminary borings are performed prior to contract award, and before the type, size and locations of the foundations are known. In this case, the primary purposes of the preliminary investigation program are a) to provide prospective bidders sufficient subsurface information to allow them to develop their pre- bid design and cost proposal, and b) to limit the risk to the owner of potential differing site condition claims after the project is awarded.

The need for additional investigation during construction should be identified in the Contract Documents, when known. For example, requiring a boring at the location of each non- redundant drilled shaft location where no boring was performed during the design stages, or requiring in situ testing to evaluate planned ground improvement measures.

C26.2.1

Alternative types of foundations should be considered and evaluated at the conceptual and/or preliminary design stages of a project to identify an appropriate and economical design for each foundation unit.

Foundations for major bridges typically fall into the following four types based on the way they transfer the load from the bridge substructure to the ground and/or their construction/installation method (Table C.26.2.1-1): • Spread Footings • Driven Piles • Drilled Shafts • Caissons – open well, pneumatic, etc.

Spread Footings Spread footings may be feasible where a competent bearing stratum such as sound bedrock or dense soil is present near the ground surface.

Requirements for large overturning moments, eccentric loads, and lateral loads may require a large footing. Generally, the weight and dimensions of the footing are designed to resist overturning and uplift. Occasionally, tiedown anchors, or tension piles have been used in conjunction with the spread footing.

Driven Piles Driven piles, including steel piles and precast prestressed concrete piles, are adaptable to a variety of soil conditions. Driven piles may not be suitable for sites with very dense soil strata or rock due to the difficulty in penetrating these materials to achieve a minimum embedment for lateral loading considerations or to achieve bearing below the design scour level. The driveability of piles is an important consideration and must be addressed at an early stage of design. The construction procedure of driven piles can be modified to suit the site conditions encountered. For example, where very dense soils are present, pre-drilling, spudding, or alternatively drilling and driving can be used to achieve the required design depth for the pile. An advantage of driven piles in comparison to drilled shafts is that the observed resistance to driving provides an indication of the nominal load resistance of the pile.

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Drilled Shafts Drilled shafts offer the advantages of high axial compressive and uplift capacity and high moment capacity. Also, with currently available technology, drilled shafts can be installed in almost any subsurface condition.

Since the performance and load of drilled shafts are influenced by the method and care in their construction, particular attention must be given to the details of the construction methods used to install the drilled shafts.

Post-grouting at the base of drilled shafts has been effectively used in granular soils to increase the end bearing resistance of the drilled shaft and to reduce shaft settlement.

Caissons Caisson foundations are large box-shaped structures that primarily act as gravity structures to support the loads from the bridge pier. Caissons can be extended to considerable depths to bear on suitable bearing material below the design scour level. Caissons may be installed by floating a prefabricated section of the caisson to the foundation site, and then sinking it into the ground in a controlled manner by balancing the rate of excavation below the caisson with the addition of weight to the caisson. Alternatively, caissons can be installed from a dry surface through temporary, man-made sand islands. Typical methods of construction include a) the open-well method in which material is removed by clamshell buckets through water-filled openings extending to the base of the caisson, and b) the pneumatic method, in which excavation at the bottom of the caisson is performed under compressed air at sufficient pressure to balance the external water pressure at the base of the caisson. Once the caisson reaches the required depth, concrete is placed at the base of the caisson to provide a continuous bearing surface. Due to their generally higher cost in comparison to drilled shaft and driven pile foundations, caissons are not commonly used for bridge foundation construction. However, they may be appropriate for very large loads (i.e., tower foundations and anchorage foundations for suspension bridges) or at deep water sites.

C26.2.2

Final selection of the foundation may be governed by factors other than subsurface conditions, including those listed herein. During the early stages of a project, it is common practice to consider alternative types and configurations of foundations to help identify an economical solution that best addresses ground conditions and site constraints.

The selection process may also consider ground improvement measures (cement grouting, chemical grouting, compaction grouting, or jet grouting) to improve the bearing capacity and deformation properties of soil or rock strata that might otherwise not provide the necessary foundation performance. These techniques could also be used to improve soils with a high potential to undergo liquefaction from dynamic earthquake loading.

Advantages and disadvantages of each type of foundation are summarized in Table C.26.2.2-1.

The type and size of the bridge as well as applicable loading conditions will govern the magnitude of the design loads along with the tolerable displacements and performance criteria. Such loading may be considerably different from that for conventional bridge structures. For example, the lateral loads at the anchor piers of a suspension bridge can be unusually large.

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The design of foundations for long-span bridges must also consider the often unusual construction loading conditions including unbalanced erection loads, hurricane loads on separate bridge elements (i.e., individual towers of a cable stayed bridge before linking the deck structure), construction equipment loads, etc.

Constructability of these typically deep, high capacity foundation elements is another major consideration during selection. Constructability issues may include the size and length of the foundation elements; the nature of the subsurface materials to be penetrated (i.e., dense soils, boulders, rock, etc.); site constraints (i.e., overhead clearance, seasonal flooding, etc); among others.

The stresses imposed on the bearing strata by these larger foundations extend to a greater depth, resulting in a greater cumulative strain. In other words, total settlement for these foundations is typically larger than foundations of conventional bridges. However, the large span lengths can generally permit displacement criteria that are greater than those typically used for conventional structures since the corresponding angular distortion between foundation units is typically smaller.

Considering the large size of foundations for long span bridges, local scour can be significantly deeper than that for conventional bridge foundations.

Considering the typically larger mass of long span bridges, seismic loading may greatly influence the bridge design. In addition, soil strength degradation from cyclic loading and liquefaction, ground displacements from lateral spreading and downdrag induced from seismic settlement can significantly increase the load on the foundations or decrease the resistance provided by these soils.

Cable Stayed Bridges Since much of the foundation settlement will occur during footing construction, tower erection and superstructure construction, before the adjoining sections of the bridge are joined, generally only post-construction settlement will impose additional stresses that the design must accommodate. The flexible cable-stayed bridges are therefore generally tolerant of foundation settlements.

Suspension Bridges Anchorages for suspension bridges are typically subjected to large uplift, overturning and horizontal loads. Generally, these large and unusual load conditions will require the construction of massive concrete anchor piers founded on caissons. Also, the weight of the bridge is mostly concentrated on the towers; caissons are therefore also a common foundation option for suspension bridge towers.

Arch Bridges Arch bridges include a) conventional arch bridges, and b) tied arches. In a conventional arch bridge the end piers or abutments transmit a considerable lateral thrust force to the foundation. Thus, the end supports must be capable of developing large lateral and vertical reactions. In a tied arch bridge, the lateral loads at the ends of the arch are resisted internally by the bridge roadway structure that links the two ends of the arch bridge. However, during construction, the construction stage loading may result in a temporary lateral thrust at the base of the arch. The choice between these two types of arch bridges should consider the suitability of the ground conditions for support of large thrust forces.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

Pile Caps at Waterline versus Mudline For water crossings, the pile caps can be located at the water line, or at the mudline (or at an intermediate depth). Each configuration offers advantages and disadvantages, as listed in Table 26.2.2-2.

Table C.26.2.2-1: Features of Various Foundations for Long Span Bridges

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Table C.26.2.2-2: Features of Waterline and Mudline Pile Caps

C26.3.1

Long span bridges pose a special challenge to bridge designers because these structures generally have a longer design life (as much as 150 years), require a greater degree of reliability, and have a greater cost of construction and maintenance. Also, such bridges typically

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

are considered life line structures that must be returned to service even after extreme loading conditions.

Often, given their longer spans, these structures are supported on multiple piers, with varying foundation types, subsurface conditions and loading conditions. For long span bridges over water, foundations supporting different piers may face significantly different loads such as at bridge towers and transition piers, and may be founded in considerably different bearing strata.

In seismic areas, during earthquakes, long span bridges often encounter spatially varying incoherent ground motions. For such conditions, a global analytical structural model of the bridge should be used along with a complete foundation model for final design. Varying ground motion time histories may be fed into each support location to model the incoherence effects. See Section D of these Guidelines for recommendations on structural analyses.

Under extreme limit state, in addition to preventing collapse and protecting life, the owner may require a design criterion of limiting damage (limited repair with short closure period). The Designer should accordingly establish limits for permanent displacements and damage under extreme loads.

For bridges located in severe marine environments, durability of the foundations requires added attention.

Bridge piers located in water need to be designed for vessel impact, or provided with a vessel impact protection system. Scour must also be considered for foundations in water.

See Section B, Article 3.2 for the state of practice employed for increasing design life for long span bridges, and Section C, Article 8.0 for current practice of varying load modifiers for an increase in reliability for different bridge elements.

Higher resistance factors may be considered for temporary support elements such as temporary raker piles to support temporary sheet piling. However, further investigation is needed to determine the appropriate adjustments to resistance factors for temporary support elements under construction loading.

C26.3.2

Where drilled shafts are installed in rock below soil overburden, the response of the drilled shafts to lateral loads is often governed by the properties of the overburden soils. Since the lateral load response of a drilled shaft is generally dependent on the soil in the upper five to eight diameters, reduction factors for group effects and minimum shaft spacing applicable to soils will be applicable for shafts in rock with overburden.

In the case where intact, strong rock is near or at the surface and a relatively low (service) level of loading does occur, reduction factors for group effects may be based on the theory of elasticity (G9).

The response of a laterally loaded drilled shaft in rock will be dominated by the secondary structure of rock, i.e., presence of joints, fissures, their orientation and in-fill materials. Therefore, a comprehensive, site specific geotechnical investigation will be required.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

C26.3.3

Permanent lateral displacements at the foundation level resulting from extreme loads are often significantly magnified at the levels of the deck, towers, and arch ribs, particularly for tall piers common to long-span bridges.

C26.3.4

It has been shown that dynamic formulae yield poor predictions of nominal pile resistance and provide no information on the stresses in the piles (G12). Pile resistance determined from dynamic formulae has shown poor correlation and wide scatter when statistically compared with static load test results (G13).

Wave Equation Analyses (WEA) overcome most shortcomings of the dynamic formulae. These analyses offer an approach to the complete mathematical modeling of the pile driving system including the hammer, cushions, helmet, pile and the soil. Moreover, pile resistance determined with WEA generally compares favorably with static load test results except in soils that exhibit a significant set-up or relaxation (increase or decrease in resistance with time after the end of driving). Considering the greater reliability required for the foundations of long span bridges, it is recommended that wave equation analyses be used for evaluating pile resistance, in conjunction with pile dynamic measurements and signal matching analyses, in accordance with the provisions of the AASHTO LRFD Specifications (5th Edition, 2010), with 2010 Interim Revisions. In addition, the test program should also include load testing as discussed in Article 26.6.

C26.3.5

Sliding resistance is determined by the lateral soil resistance on the base and partial passive pressure on the face of the caisson. Since the displacement needed to mobilize full passive resistance is larger than the displacement needed to develop full sliding resistance, the passive resistance must be appropriately reduced to be compatible with sliding resistance.

Side friction is generally ignored when calculating resistance to vertical loads because the methods used for caisson installation typically include jetting and/or bentonite slurry injection to lessen side resistance, which is necessary for advancing the caisson to the required foundation elevation.

Calculation of differential settlements should consider the lateral variation of the soil conditions, unsymmetrical weight distributions and predominant directional loading.

Modeling of Caissons After the caisson is sized using conventional limit equilibrium procedures for static analyses, dynamic analyses may be required in seismic areas.

Traditionally, the soil-foundation-structure interaction effects for bridges supported on large gravity type caissons have been considered by using sub-structuring methods (i.e., the uncoupled approach). In this method, the soil-foundation system is treated as an elastic or equivalent-elastic system in the frequency domain (i.e., an elasto-dynamic analysis).

Caissons are generally massive structures embedded in the ground and constitute a large portion of the weight of the bridge tower. Therefore, the fundamental period of vibration of the

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caisson should be small compared to the rest of the bridge. Since the modes of vibration of the long-span bridge structure and the gravity caisson are different, an uncoupled analysis would be expected to provide reasonable estimates of foundation behavior (Figure C26.3.5-1).

Reasonably accurate results can be derived from the elasto-dynamic approach because of the distinct modes of vibrations between the bridge and the caissons. However, if the stability of the caisson itself needs to be addressed in the analysis, either due to the large lateral loads, caissons being in weak soils, or when the rocking mode dominates, then the elasto-dynamic approach may not provide reliable results. In this case, a soil-foundation model with allowance for base separation and soil yielding should be incorporated into the analysis to capture the geometric as well as soil non-linearity effects.

Figure C26.3.5-1 Caisson Model for Dynamic or Vibration analyses

A comprehensive soil-foundation structure model subjected to spatially varying ground motions (i.e., different motions at each foundation) can be developed and incorporated into the global structural model. The soil stiffness and damping can be modeled by horizontal and shear springs with a hysteretic force-displacement relationship to represent the non-linear inelastic behavior as well as dissipated energy in the soils surrounding the foundations. The interfaces between the foundation and the surrounding soils (both at the base and along the sides) can be explicitly modeled to allow separation (i.e., gap elements) and slip, in the event that the shear/tensile strengths are exceeded during the simulated dynamic soil-foundation structure analysis. One of the advantages of this complete model is that the dynamic earth pressures induced along the exterior surfaces of the caisson wall can be directly derived from the response analysis of the bridge to assess the response of the caisson foundation.

C26.3.6

The number, size and shape of the foundations can have a significant effect on river or sea bottom hydraulics and related scour.

Scour can result in significant unbraced pile length, loss of friction within scour depth, reduced end bearing resistance, and reduced side resistance below the scour depth. These will impact both the axial and lateral resistance of the bridge foundation, and must be addressed in foundation design.

In lateral capacity analyses, any reduction in overburden stress due to global and local scour shall be incorporated as shown in Figure 26.3.6-1.

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C26.3.7

Refer to Section C of these Guidelines for more information.

C26.4

For cable-stayed or suspension bridges, much of the settlement occurs during construction when segments of the bridge are either unconnected or connected by flexible cable elements only. For such cases, total settlement may not be a primary consideration for foundation design. Of more significance would be settlement (in particular differential settlement) that would take place after sections of the bridge are connected and made continuous.

When necessary, adjustments should be made to the bridge bearing level to compensate for settlement that occurs during construction. This may be appropriate to maintain bridge clearance, or to reduce the differential settlement between adjacent foundations.

Stresses imposed by the generally larger foundations of a long span bridge often extend to greater depths than for more conventional bridges, and may therefore result in a greater cumulative strain. As a result, the total settlement of foundations for long span bridges may be greater than that for a conventional bridge. However, the settlement of foundations for both long span and conventional bridge foundations will likely be similar if the foundations are extended to competent rock or a dense soil bearing stratum. The settlement of the foundations for long span bridges can be computed using the same methods applicable to conventional bridge foundations, with the appropriate adjustments for foundation size and depth.

Analyses of foundation displacements should consider a range of soil and rock properties to assess the influence on the foundations due to the anticipated variability in subsurface conditions, the uncertainties inherent in the interpretation of the subsurface exploration data, and the limitations associated with analytical models used for estimating foundation displacements. A softer ground response would typically be associated with larger foundation displacements. However, a stiffer ground response may result in increased foundation loads, e.g. under seismic loading, or when combined with softer adjoining foundations.

Although the total settlement of the larger, more heavily loaded foundations of a long span bridge are often greater than the settlement of foundations for conventional bridges, long span bridges generally have a greater tolerance for settlement and lateral displacement than conventional bridges due to the much larger spans between foundation units. The serviceability criteria for long span bridges, including settlement, differential settlement (tilt) at any individual foundation, differential settlement between adjacent foundations, and lateral displacement, should be determined by the project structural engineer based on the type of bridge structure, the span arrangements, and the arrangement and details of the support bearings. In developing these serviceability criteria, consideration should be given to possible magnification of lateral displacements at the superstructure level due to the generally greater height of the bridge piers of a long span bridge.

For extreme event loadings, some damage to the structure can be considered acceptable, but the structure must be designed to prevent collapse. Also, considering the importance of long span bridges, the design criteria may require the structure to be quickly repaired and returned to service following an extreme event.

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C 26.5

Typically, long span bridges include some length of approach structure along which there is a general reduction in the height of the structure. The abutment, therefore, may have a height similar to abutments for conventional types of bridges. However, lower construction costs can generally be achieved by shortening the length of the approach structure and replacing it with high approach embankments and correspondingly higher abutments. In such cases, the design of the abutments should consider the following issues:

• Larger settlement of the abutment due to the possibly greater height and weight of the approach embankment, and the larger foundation loads. • Larger differential settlement between the approach embankment and the bridge deck due to the possibly greater height of the approach embankment. • Global stability of the approach embankment and abutment considering the potentially greater height of the embankment. • Downdrag loads on the abutment foundation. • Larger lateral earth pressures and corresponding overturning moments associated with a potentially higher approach embankment. • Increased dynamic earth pressures and wall inertia effects due to seismic loading for higher abutments. • Greater seismic forces from the bridge deck through bearing supports. • Constructability considerations.

For more information, refer to (G11) and (G14).

C 26.6

This section presents requirements for performing load tests on driven piles and drilled shafts, and the integrity testing of drilled shafts, with a focus on applications for foundations of long span bridges.

Load tests should be routinely performed for the foundations of long span bridges considering the generally high cost of the foundations, the consequences of a deficient foundation, and the greater reliability expected for the foundations of these structures. These guidelines include a requirement for at least one load test at each main pier and at each anchor pier, and at other foundation locations identified as separate “sites” within the project area.

Consideration should be given to performing load testing during the design stage of a project to help define an appropriate and economical foundation for the bridge, and to better estimate the size, length, and arrangement of the foundation piles or drilled shafts.

C26.6.1

A test program for driven pile foundations typically includes the following components:

- Wave Equation Analyses (WEA) - Dynamic testing during pile installation - Signal matching analysis of the dynamic test data - Static load testing - Revised WEA

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- QC testing during production pile driving

Additional guidance on testing of driven piles is provided by Hannigan, et al (G10).

C26.6.1.1

The results of Wave Equation Analysis are used to evaluate the suitability of the selected hammer in driving the pile to the design depth and resistance. WEA can also be used to predict driving stresses in the pile. Driving stresses are a particular concern when driving concrete piles.

WEA can be useful in evaluating hammer selection and pile cushion requirements to help avoid potentially damaging driving stresses. It can also be used to develop preliminary driving criteria for the test piles.

C26.6.1.2

Considering the greater reliability expected of the foundations of long span bridges, these guidelines include the routine use of dynamic pile testing of driven piles for these major structures.

Dynamic testing provides field measurements for assessing the suitability of the hammer and cushion system, measures the energy transferred to the pile top, and can identify excessive stresses that might occur in the pile during driving. Moreover, estimates of pile resistance are obtained from these measurements using signal matching methods. Also, information on set-up can be obtained by comparing dynamic measurements from the end of initial driving (EOID) with those obtained at the beginning of restrike (BOR). Dynamic test data can also provide information for assessing the structural integrity of a pile if damage during driving is suspected.

For piles larger than 48 inches in diameter, the increased number of strain gauges and accelerometers provides information regarding eccentric impact from a misaligned hammer.

Signal matching analysis provides an estimate of nominal pile resistance, the distribution of soil resistance along the length of the pile, and the various soil parameters (such as damping and quake values) required for a revised Wave Equation Analysis. Signal matching analysis should be performed for representative hammer blows both near the end of initial driving and at the beginning of a restrike.

C26.6.1.4

ASTM D 7383 specifies two alternative procedures, including “Procedure A” where the compression force pulse is applied at the top of the pile using a combustion gas pressure, and “Procedure B” where the compression force pulse is applied by dropping a mass onto the cushioned top of the pile. Each of these methods estimates resistance along the pile by evaluating the dynamic response of the pile and the dynamic characteristics of the supporting ground.

Currently, there are no generally accepted resistance factors for the force pulse (rapid) test method. Until resistance factors are codified for the force pulse (rapid) test method, foundation designers should exercise judgment in the application of the test data. A conservative approach

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may be to use the resistance factors specified for dynamic testing, or to calibrate the load test results with data from static load tests on the same or similar pile.

Advantages • Test equipment can be easily mobilized, erected, and disassembled. • The test is performed rapidly, with little preparation of the pile. • Can be performed for capacities as high as 3600 tons, or more.

Disadvantages • The test results require greater interpretation than static load test data.

Force pulse (rapid) load tests may not be as reliable as static load tests for predicting the static capacity of the piles in some soil conditions, particularly in cohesive soils, where the rate of loading can influence the resulting soil resistance.

C26.6.2

A test program for drilled shaft foundations typically includes the following components:

- Load testing - Integrity testing

Conventional static load tests and force pulse (rapid) load tests have been performed on drilled shafts. However, these methods may not be practical or feasible for testing high capacity drilled shafts that are typically used for long span bridge foundations. Bi-directional load cell testing, in which the load is applied by a hydraulic jacking mechanism cast within the shaft, is usually a more practical and economical method for determining the axial resistance of a high capacity drilled shaft, and currently (2010) is the most commonly used method for load testing drilled shafts.

Additional guidance on testing of drilled shafts is provided by Brown, Turner and Castelli (G8).

C26.6.2.1

Considering the large diameter and high axial resistance typical of drilled shaft foundations for long-span bridges, the use of conventional static load tests using a dead weight or reaction frame to load the top of the shaft is generally not practical.

C.26.6.2.2

The bi-directional load cell test involves inserting a sacrificial hydraulic jack system within the drilled shaft; when expanded, the bi-directional load cell reacts against the buoyant weight of the drilled shaft and shaft side resistance to develop a downward load on the section of shaft beneath the load cell(s). This process is continued until the load cell reaches its maximum load capacity, or until either the side resistance or end bearing reaches nominal resistance, or to the limiting extension of the load cell (typically about 6 inches), whichever occurs first.

Bi-directional load cell testing has become a reliable and generally accepted method of testing drilled shafts. However, at present (2010) there are no ASTM standards for bi-directional load cell testing. Accordingly, until such standards are available, caution must be exercised in the

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

planning, execution and interpretation of the results of load tests using this method. The guide specifications referenced herein include interim specifications with suggested procedures for performing bi-directional load cell tests based on current general practice.

Advantages • The bi-directional load cell test method provides a practical and economical means for testing high capacity drilled shafts. The load capacity of individual load cells can be as high as 3,000 tons, both upward and downward, for a maximum test load of 6,000 tons from a single load cell. For large diameter shafts that allow three load cells to be placed at a single level, the maximum test load is increased to 18,000 tons. Additional capacity may be available for long shafts with the use of multiple levels of load cells. • Instrumentation can provide direct determination of end bearing resistance as well as friction resistance along the length of the shaft.

Disadvantages • It requires advance installation of the load cells in the drilled shaft. Therefore, it is not feasible to conduct this test on a questionable production shaft after the shaft concrete is placed. • It requires that there be sufficient side resistance and end bearing resistance to develop the required test load, or a costly reaction frame may be necessary to provide additional resistance. • Since the test applies an upward load to the shaft, the friction resistance is in the opposite direction to the axial compression loading on a production shaft; however, this is considered to provide a conservative estimate of the nominal friction resistance.

C26.6.2.3

See Section C26.6.1.4.

C26.6.3

Drilled shafts can experience construction defects such as necking, bulging, voids, honeycombing, loss of concrete cover, material intrusion, etc., particularly drilled shafts installed using the wet method of construction. Therefore, drilled shaft integrity testing should be an essential part of long span bridge projects.

Integrity testing may be used for some or all of the following purposes, depending on the details of the drilled shaft foundation design:

• at the start of construction, to confirm the suitability of the contractor’s proposed shaft installation methods • routinely, or periodically, during drilled shaft production operations to continue to assess the suitability of the contractor’s approved shaft installation method • to evaluate the structural integrity of the shaft whenever there is a significant change to the contractor’s means and methods of drilled shaft installation • to investigate drilled shafts that are suspected to have defects in the shaft concrete • to evaluate the success of any remedial measures performed on a drilled shaft

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

The more common methods available for integrity testing include:

Internal Integrity Test Methods • Cross-hole Sonic Logging CSL and, • Gamma-Gamma Density Logging

External Integrity Test Methods • Sonic Echo • Impulse Response

External integrity testing methods require a greater degree of interpretation and do not provide the same reliability as the internal methods. For example, in sonic echo tests, toe reflections may be observed up to 30 shaft diameters in favorable conditions but defects such as cracks, bulging, necking or cold joints may reflect the low energy wave and prevent detection of deeper defects. Also, weak reflections may be observed for drilled shafts founded on or socketed into rock since the modulus of the rock may be comparable to, or greater than that of the concrete shaft.

Of the available test methods, Cross-hole Sonic Logging (CSL) is the most frequently used method. It provides test data for the entire length of the drilled shaft, can be used to estimate the size and location of an anomaly, and, when combined with tomographic imaging, can help develop an approximate 3D image of the anomaly.

CSL testing should be performed at all technique and test shafts, at all drilled shafts of a foundation with non-redundant shafts, and at all drilled shafts installed using the wet method of construction.

Consideration should be given to installing CSL tubes in all drilled shafts, and performing CSL testing at random drilled shaft locations during the progress of foundation construction to confirm the continued success of the contractor’s drilled shaft installation procedures. Care should be taken during construction to prevent foreign material from getting into the CSL tubes.

Whenever integrity testing identifies a significant anomaly in a drilled shaft, further investigations should be performed to determine if a defect is present, and to better define the nature and extent of this defect. The additional investigations may include: • re-evaluation of drilled shaft inspection records • additional integrity testing, such as CSL testing through additional tube combinations or use of tomography measurements (CSL measurements with the source and receiver probes at different levels) • use of different integrity testing methods, such as Gamma-Gamma logging, sonic echo testing, etc. • coring of the drilled shaft to retrieve samples of the concrete from the zone in question.

Further investigation measures may include down-hole camera inspection within the core holes; water pressure testing within a single core hole, or between two or more core holes; among others.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

C26.7.1

The term “cofferdam” is used to describe a temporary or permanent structure that is used to retain water and/or soil to enable a foundation site to be fully or partially dewatered for construction. These structures generally consist of 1) vertical steel sheet piling, 2) a bracing system composed of wales and struts or ring beams (or occasionally tensioned ground anchors for cofferdams onshore), and 3) a bottom tremie seal, if required, to prevent water seepage from the bottom of the excavation.

Cofferdams usually fall under the category of temporary structures necessary to construct bridge piers. Cofferdams are typically used for water crossings where the pile cap is below water level. Cofferdams may also be used on land for construction of deep footings.

As a temporary structure, the Contractor is typically responsible for design, construction, maintenance and removal of cofferdams. However, the contract documents must address the general requirements for design and performance of the cofferdam. The Engineer is responsible for evaluating the feasibility of using a cofferdam for construction of the foundation, and for checking and approving the Contractor’s shop drawings.

Rectangular cofferdams are typically used for bridge piers since their shape allows for the use of cross bracing. Elliptical and circular cofferdams have also been used for deep water piers. Circular cofferdams offer an advantage since perimeter ring beams are used for internal bracing, eliminating cross bracing and providing an unobstructed work area inside the cofferdam.

Single wall, steel sheet pile structures are most commonly used for excavations of moderate depth. Cantilevered sheet piling may be sufficient for shallow excavations or when the cofferdam is only partially dewatered (for construction of pile caps that extend partially below the water level). More typically, one or more levels of internal bracing are provided to support the cofferdam sheeting. Master pile systems and interlocking H-pile or pipe pile walls can be used for excavations extending deeper below the water level, or when a wider vertical spacing of bracing is desired. These guidelines only address cofferdams of the single wall type.

Other types of cofferdam structures include double walled cofferdams and cellular cofferdams. These types of cofferdams are generally limited to very large or deep excavations. These structures may also be suitable for sites where shallow rock prevents installation of sheet piles of a conventional cofferdam to the penetration necessary for lateral support. Double walled cofferdams comprise two parallel rows of sheet piles connected together by tie rods or sheet pile diaphragm walls, and the space between the walls is filled with granular materials such as sand, or crushed rock. Cellular cofferdams consist of rings of steel sheet piles filled with granular soil, and linked with sheet pile arcs to provide a continuous gravity wall structure.

C26.7.2

The design must define the sequence of construction to address all the load cases and corresponding support conditions and water levels associated with the construction and removal of the cofferdam. For each stage of construction, the Designer must calculate the required sheeting penetration, maximum bending moment and bracing loads. The structure elements must then be sized for the largest bending moments and bracing loads determined from all of the construction stages. It should be noted that bending moments and bracing loads during

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

intermediate stages of construction may be considerably greater than those in the completed cofferdam.

Sheet pile drivability should be evaluated as part of the design process, particularly if rock or dense soils are present at a shallow depth. This may control the penetration of the sheeting and the corresponding arrangement of the bracing system.

Since the majority of cofferdams are constructed as temporary works it may be uneconomical to design for all possible load cases. Appropriate risk assessment and decisions should be made by the Owner and the Contractor together to determine the level of risk that is acceptable. For example, when seasonal river flooding is unpredictable, it may be impractical to design for the maximum recorded flood level. In such cases, the design may include provisions for evacuation and flooding of the cofferdam when the river approaches a flood level corresponding to a return period more appropriate to the short duration of cofferdam service.

For cofferdams in water, an interlock sealant may be used to restrict the inflow of water. Since the problem of sealing a leaking cofferdam after construction may be difficult, it is prudent to use an interlock sealant as a matter of course.

Where practicable, the site to receive the cofferdam should be pre-dredged to the bottom elevation of the tremie concrete seal. Generally, a few extra feet are removed to permit placing a sand-and-gravel blanket. This blanket provides a level base for the tremie concrete seal.

Some of the more common causes of cofferdam failure include: • Failure to take into account the possible range of water levels and subsurface conditions • Failure to verify design assumptions with conditions observed during excavation • Overexcavation during construction • Inadequate bracing provided to support the loads • Insufficient penetration of sheeting to prevent piping or heave • Insufficient penetration or splitting of interlocks due to boulders or other obstructions. • Misalignment of the sheeting during setting and driving

A tremie seal is typically required to seal the entire bottom of the cofferdam to prevent water seepage into the excavation and the related risk of bottom instability due to piping. The tremie seal also provides a rigid support for the sheet piles at the bottom of the excavation. The tremie seal is designed to resist the hydrostatic pressure by its own buoyant weight, or by a combination of self-weight and the uplift resistance provided by piling or drilled shafts, as transferred by the bond between the concrete seal and the piles or drilled shafts. Concrete tremie seals are typically not reinforced, but reinforcement can be used to distribute the uplift loads to the piles or shafts.

A concrete tremie seal may not be necessary for shallow cofferdams where low seepage gradients result in a limited inflow that can be discharged by pumping, or for cases where the cofferdam sheeting extends to an impermeable soil stratum that serves as a cutoff to seepage. Whenever a tremie seal is omitted, it is essential to verify that the seepage gradients will not cause instability (piping) of the soils below the bottom of the excavation. Piping, and the related loss of soil, could undermine the cofferdam, lead to failure and uncontrolled flooding of the cofferdam, damage previously installed foundation elements, and cause subsidence or lateral displacement of adjoining existing structures.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Commentary G

Often, the design engineer will determine a preliminary depth and thickness of the tremie seal to assess the influence of the seal on the bridge foundation, and for preliminary cost estimating purposes. However, since the tremie seal is used only for support of construction loading, the design of the tremie seal is typically the responsibility of the construction contractor. The design of the tremie seal and any other temporary works designed by the contractor should be reviewed by the Engineer-of-Record, or a designated representative.

C 26.8

Rockfill islands are an effective means of protecting bridge foundations from damage due to vessel impact. When a vessel collides with the island, it will plow into the island and/or ride up the side slope of the island, depending on the water level, ship ballast and bow geometry. Rockfill islands are an alternative to protection dolphins and conventional fendering systems. However, they require large quantities of materials such as quarry rock, and have a large footprint within a waterway.

Rockfill islands protect the bridge pier and foundation by dissipating the energy from vessel impact by converting the vessel’s kinetic energy into: • potential energy as it glides up the island. • strain energy as the rock island material shears, crushes and displaces and the hull of the ship is crushed. • frictional energy as the ship plows into the island. • dynamic pore pressures generated within the rock fill.

Other potential benefits of rockfill islands include: • Longer life and less maintenance than dolphins and fendering. • Provides lateral support to the pier foundation elements. • Rockfill island with waterline pile cap is an economical alternative to mudline pile cap and associated temporary cofferdam. • Facilitates construction of foundation pile cap (eliminates suspended bottom form).

Additional guidelines are provided in the AASHTO “Guide Specification for Vessel Collision Design of Highway Bridges” (G1).

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

i FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

TABLE OF CONTENTS

A 1.0 Proportions of Suspension Bridges ...... 1 A 1.1 Cable Sag ...... 1 A 1.2 Superstructure Depth ...... 1 A 1.3 Tower Height and Size ...... 2 A 2.0 Layout and Points of Fixity ...... 2 A 2.1 Curved Alignment ...... 2 A 2.2 Fixity at Cable Anchorages and Expansion at Towers ...... 2 A 2.3 Continuous Superstructure ...... 2 A 3.0 Initial Sizing of Principal Elements ...... 3 A 3.1 Cable Sizing ...... 3 A 3.2 Suspender Sizing ...... 3 A 4.0 Main Cable Fabrication, Erection, and Maintenance ...... 3 A 4.1 Air Spun Cables vs. Prefabricated Parallel Wire Strands (PPWS) ...... 3 A 4.2 Cable-Compaction Void Ratio ...... 5 A 4.3 Corrosion Protection of Main Cables ...... 5 A 4.3.1 Wire Wrapping ...... 5 A 4.3.2 Plastic Cover ...... 6 A 4.3.3 Dehumidification ...... 6 A 4.4 Center Cable Tie ...... 6 A 4.5 Wire Splices ...... 7 A 4.6 Main Cable Strengthening ...... 7 A 4.7 Main Cable Replacement ...... 7 A 5.0 Suspension Cable Anchorage ...... 8 A 5.1 Corrosion Protection Inside the Anchorage ...... 8 A 5.2 Self-Anchored Suspension Bridges ...... 9 A 6.0 Erection ...... 9 A 6.1 Erection Stage and Stress Analysis ...... 9 A 6.2 Superstructure Erection ...... 10 A 6.3 Saddle Offsets at Top of Towers ...... 10 A 6.4 Creep and Shrinkage Effects with Concrete Towers ...... 11 A 7.0 Suspenders and Cable Bands ...... 11 A 7.1 Suspender Materials ...... 11 A 7.2 Suspender Fabricated Length and Adjustments ...... 11 A 7.3 Cable Bands ...... 12 A 7.4 Suspender Sockets ...... 12 A 7.5 Suspender Stress ...... 13 A 7.6 Suspender Oscillation in Wind...... 13 A 8.0 Superstructure Details ...... 13 A 8.1 Torsional Box ...... 13 A 8.2 Steel Box Segment Shipping ...... 14 A 8.3 Wind Locks ...... 14 A 8.4 Rocker Links ...... 14 A 9.0 Vertical Suspenders vs. Inclined Suspenders ...... 15 A 10.0 Vertical Cable vs. Inclined Cable ...... 15

ii FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

LIST OF FIGURES

Figure A 1 Side Span Types ...... 16 Figure A 2 Operation Principle of Aerial Spinning ...... 17 Figure A 3 Prefabricated Parallel Wire Strand ...... 18 Figure A 4 PPWS Installation ...... 19 Figure A 5 Main Cable Compactor ...... 20 Figure A 6 Main Cable Compactor ...... 21 Figure A 7 Cable Compaction Void Ratio ...... 22 Figure A 8 Elastomeric Wrapping ...... 23 Figure A 9 Corrosion Protection of Main Cable ...... 24 Figure A 10 Center Tie Detail ...... 25 Figure A 11 Center Tie Detail ...... 26 Figure A 12 Wire Splice Detail ...... 27 Figure A 13 Main Cable Strengthening ...... 28 Figure A 14 Main Cable Replacement ...... 29 Figure A 15 Anchorage Types ...... 30 Figure A 16 Corrosion Protection Inside Anchorage ...... 31 Figure A 17 Suspension Bridge Types ...... 32 Figure A 18 Example of Erection Load Induced Bending Moments ...... 33 Figure A 19 Superstructure Erection Methods...... 34 Figure A 20 Cable Saddle ...... 35 Figure A 21 Temporary Cable Pull at Tower ...... 36 Figure A 22 Creep and Shrinkage Effect of Concrete Tower ...... 37 Figure A 23 Suspender Materials ...... 38 Figure A 24 Suspender Fabrication Length ...... 39 Figure A 25 Suspender Socket ...... 40 Figure A 26 Relative Movement between Suspenders ...... 41 Figure A 27 Suspender Oscillation in Wind ...... 42 Figure A 28 Torsional Box ...... 43 Figure A 29 Wind Lock ...... 44 Figure A 30 Rocker Link Detail ...... 45 Figure A 31 Suspender Geometry ...... 46 Figure A 32 Cable Geometry ...... 47

iii FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

APPENDIX A – SUSPENSION BRIDGES

Topics selected are those deemed likely to be of immediate application for the readers of these Guidelines. Materials presented are available in the literature and provided to supplement the Guidelines as a convenient source of reference. An overview1 is suggested for the readers who seek general and historic perspectives of this bridge type.

A 1.0 Proportions of Suspension Bridges

A bridge deck suspended from a pair of cables is the basic configuration of a suspension bridge. The main cables are supported by towers and the ends of the cables are typically anchored either to rock or gravity anchorages. The main cable follows a natural catenary, although the actual shape depends on the distribution of dead loads.

A 1.1 Cable Sag

The span to sag ratio of the suspension cable determines the characteristics of a suspension bridge. Span is defined as the distance between two towers. The sag is defined as the vertical distance measured from the top of the tower to the lowest point of the cable (see Figure A1). The following are preferable ratios of the cable sag to the span for the more commonly used suspended span arrangements - • Approximately 1/9 for bridges with a single suspended span (only main span is suspended) • Approximately 1/8 for bridges with 3 suspended spans (suspended main and side spans) • Approximately 1/10 to 1/12 for highway bridges with 3 suspended spans and lighter superstructure, including narrower and orthotropic deck sections

The lower sag/span ratio will increase the cable force and, consequently, increase the cable size and weight. However, a bridge with lower sag/span ratio will be stiffer.

A 1.2 Superstructure Depth

The stiffness of a suspension bridge is characterized by the resistance of the cable against changing its shape. The depth of the stiffening girder is controlled, structurally, by local stiffness requirements for distributing service and construction loadings. For highway bridges with superstructure comprised of stiffening trusses, the following ratio of the depth of the stiffening trusses to the length of the main span are often found to be feasible: • Span length < 1000 feet, ratio approximately 1/60 • Span length < 2000 feet, ratio approximately 1/70 - 1/90 • Span length < 3000 feet, ratio approximately 1/90 – 1/150

A streamlined steel box superstructure usually has a much lower ratio.

To provide lateral stiffness, the center to center of the stiffening trusses is preferably not less than 1/30 of the span length.

1 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

The ratio for side span to main span should be about 0.5 for 3 suspended span bridges and 0.25 for those with a single suspended span. Lower ratios tend to make the stresses in the backstays greater than those in the cables over the main span.

A 1.3 Tower Height and Size

The height of towers above the deck should be the additive of the desired cable sag and the following: 1. The elevation difference between the cable and the roadway surface at the lowest point of the cable. 2. The elevation difference between the roadway at the towers and at the cable low point.

The tower top cross section needs to be large enough to accommodate the cable saddle. Access to the saddle should be considered when laying out the tower head.

A 2.0 Layout and Points of Fixity

A 2.1 Horizontal Alignment

The suspension cables of a ground anchored suspension bridge are always erected first and each cable will naturally hang plumb. The superstructure of major suspension bridges is typically erected by segments hoisted vertically by lifting equipment mounted on the suspension cables. It is therefore natural for major suspension bridges to be used only for a straight alignment.

Some bridges have one or both side spans on piers instead of being supported by the suspension cables. The George Washington Bridge is a good example. The roadway not supported by the suspension cables can make a turn off the straight alignment.

A 2.2 Fixity at Cable Anchorages and Expansion at Towers

In a typical suspension bridge, the superstructure is generally connected to the cables at the main span via a center tie and expansion of the superstructure is provided at the towers. Vertical restraint at the towers is usually provided by a pair of vertical links, which allow longitudinal movement. The side span superstructure is usually fixed at the anchorage. Vertical links are again provided at the towers to allow longitudinal movement associated with temperature variation. Due to the flexible connection, the tower tends to take less earthquake load. Instead, the connection between the anchorage and the superstructure tends to receive large earthquake load. The short suspenders will not bend in the longitudinal direction since the cable and the deck are connected positively. The longer suspenders are located where the movement is largest. Overall, this arrangement serves to minimize bending of the suspenders.

A 2.3 Continuous Superstructure

In some cases, the superstructure is made continuous from anchorage to anchorage. The Great Belt (Storebaelt) Bridge in Denmark is a major suspension bridge that follows this design approach. The superstructure is supported by suspenders at the tower instead of links. The design reduces the number of expansion joints from the normal 6 to 2 large ones at each anchorage location to accommodate all superstructure movement and length variation. This

2 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

design causes significant relative movement between the deck and cable near the anchorages where the suspenders are very short. In order to mitigate the otherwise damaging bending of the suspenders, hinges were installed at the ends of the suspenders. The advantage of this type of design is the superstructure is isolated from the towers and anchorages, which would help minimize earthquake loading. Supporting the superstructure by suspenders at the towers removes the “spikes” in negative moment associated with the stiffness of link supports.

A 3.0 Initial Sizing of Cable Elements

A 3.1 Cable Sizing

The horizontal component of the cable force is constant throughout the suspension cable from one anchorage to the other. However, the internal force along the length of the cable varies depending on its inclination. The inclination is always largest at the tower saddle. Moreover, the cable inclination on the main span side is not necessarily the same as that of the side span side. Usually the inclination on the side span side is greater due to a shorter side span and the lower position of the cable anchorage. As a result, the highest internal force in the cable tends to be located at the side span side of the saddle.

The suspension cable can be sized to accommodate the highest force along the cable. This will result in a uniformly sized cable with the same number of wires throughout its entire length. This would be easier for construction but material quantity will be higher. An alternative approach for minimum material would have the number of wires in the cable designed according to the maximum cable force over the main span, and additional wires would be added as needed on the side span side to satisfy the higher demand there. This approach will require anchorages at the tower saddles, which could offset savings from minimizing wire quantity.

The cable force can be estimated based on assumed deck and cable weights; a fully loaded uniform live load; and the assumed cable sag. The cable force divided by a selected factor of safety would produce the design cable force.

A 3.2 Suspender Sizing

The axial force in a suspender is computed based primarily on dead and live load. An initial estimate of the axial force can be computed by simply using uniform load and tributary area. A factor of safety equal to or greater than five (5) may be applied to the estimated axial force for the design axial load. The actual sizing of the suspender would follow by selecting a wire rope from a manufacturer’s product list which provides a breaking strength that exceeds the design load.

A 4.0 Main Cable Fabrication, Erection, and Maintenance

A 4.1 Air Spun Cables vs. Prefabricated Parallel Wire Strands (PPWS)

The aerial spinning method of parallel wire cables was invented by John A. Roebling and used for the first time on the Niagara Falls Bridge which was completed in 1855. The basic principle behind the aerial spinning method is quite simple. After the first wire in a strand is pulled from the wire reel and is anchored at the far end, a spinning wheel is used to weave additional loops of wires into the strand shoes at the cable anchorages (see Figure A2). Though simple in principle, significant know-how and planning is required to carry out the air spinning operations 3 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

properly. One significant challenge is to ensure that the wires in the assembled cables are of equal length. In the aerial spinning method, individual wires would be spun in the free-hanging condition. This implies that the sag of each wire had to be individually adjusted to ensure that all wires would end up with an equal length. Reverse engineering is needed to determine the sag required of the first wire being installed. Advance planning and continual monitoring of the changing sag in the partially assembled cable is necessary to make sure the final cable would have the proper sag. A new method, sometimes called the tension-control method, was developed in Japan. The idea is to keep the tension in the wire constant during cable spinning to achieve the desired uniform wire length. Nevertheless, this method would still require some adjustment of the individual strand.

The Prefabricated Parallel Wire Strands (PPWS) method requires wires to be to predetermined unstressed (prefabricated) length and placed in a bundle to form a strand. The strand is socketed at both ends prior to installation (see Figure A3). The prefabrication process may be done off site or may be done on site. The trade-off is simply the cost of shipping vs. the cost of setting up assembling facilities at the site plus the higher cost of site labor. During installation, the strands are erected one-by-one from one anchorage to the other anchorage (see Figure A4).

Until the 1960s, aerial spinning was the suspension cable erection method of most suspension bridges in the United States. The PPWS method was developed to compete against the aerial spinning for its economy. While it is generally accepted that the PPWS method is a faster way to install suspension cables, the overall cost-effectiveness of the PPWS method is not universally recognized. The Newport Bridge, when completed in 1969, was the only suspension bridge in the United States that used the PPWS method. The recently completed new Tacoma Narrows Suspension Bridge2 and the Carquinez Bridge used the aerial spinning method to install their suspension cables.

The PPWS method needs special equipment and facilities to fabricate the strands to a specific length in a controlled environment. The return on investment appears to have been the reason why countries such as Japan and China, with many suspension bridges constructed and more in planning, have developed the equipments and facilities, and employed the PPWS method in many of their major projects. The Akashi Kaikyo Bridge in Japan, which is currently the longest bridge in the world, was constructed using the PPWS method. An exception to the rule would be the Tsing Ma Bridge in Hong Kong, China, where the contractor elected to install the cables by the aerial spinning method. Interestingly, the self-anchored suspension span of the new East span of the San Francisco Oakland Bay Bridge, having received a waiver on the Buy America Clause, is expected to use foreign materials and the PPWS method for its suspension cables.3

Due to the capacity limitations of erection equipment, the PPWS method will tend to limit the number of wires in each strand and result in a large number of smaller strands in a cable. On the other hand, the aerial spinning method could carry one to four wires at each run of the spinning wheel. Therefore, the number of wires per strand is theoretically unlimited and larger strands may be preferable when the aerial spinning method is used. The size of the strand would tend to be limited by other factors, such as the size of the tie rod. The size of the anchorage chamber is determined by the number of tie rods.

In general, the PPWS method would tend to shorten the duration of cable installation although this is not necessarily an indication that it is more cost effective than the aerial spinning method. The need for a larger splay chamber, for example, may offset the savings associated with the PPWS method. The Designer should therefore assess the trade-offs, including input from interested constructors, and design for one method that would most likely be cost-effective for

4 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

the given site. Consideration may be given to permit contractor proposed alternative design of the chamber and anchorage system.

A 4.2 Cable-Compaction Void Ratio

The suspension bridge main cable is usually comprised of thousands of wires, each with a diameter of approximately 5 mm. Once all wires are assembled, cable bands, typically made of cast steel to tight tolerance, will be installed to clamp around the cable, to provide enough friction to prevent the cable bands from sliding along the inclination of the cable sag. Also, the cable will generally be wrapped by steel wires as part of its corrosion protection. To facilitate both clamping by the cable band and wire wrapping, the wires in the cable must be compacted into a circular shape. See Figures A5 & A6 for schematics of the cable compactor. The theoretical void ratio (i.e. the ratio of void to steel-plus-void) is approximately 9.31%. Most bridge cables can achieve 18% to 19% (see Figure A7). The radius of the cable band to the cable contact surface is required for fabrication. The Designer needs to predict the diameter of the compacted cable, so that the cable band and other appurtenances can be designed.

It should be noted that the void ratio changes during construction as additional load is applied to the cable. Provisions should be made for retightening cable clamps for suspenders during construction. Cable clamps should also be tightened periodically through the life of the structure to address steel relaxation.

A 4.3 Corrosion Protection of Main Cables

A 4.3.1 Wire Wrapping

The traditional cable corrosion protection method consists of the application of lead rich paste onto the compacted parallel wire surface and then wrapping the parallel wires with galvanized soft steel wires (see Figure A9). The soft wire is typically 3.5 mm in diameter. The soft wire is typically stressed to around 150 MPa by a wrapping machine.4 The surface would be further painted to prevent water intrusion. However, the length of the cable changes with bridge loading and temperature. As a result, gaps tend to form between the wrapping wires, which lead to paint cracking along the gaps. With the lead rich paste drying up and cracking over time, rain water would eventually get into the cable through these cracks.

In spite of the fact that steel wire wrapping would not form a water tight barrier, it does provide a strong surface to protect the main cable wires from being dented or cut during construction and in service. Consequently, most modern suspension bridge cables would still use the steel wire wrapping while providing additional measures to keep the interior of the cable dry. On some bridges the round wrapping wire has been replaced by S-shaped wire to reduce paint cracking. Kurushima Kaikyo Bridge in Japan, completed in 1999, has used S-shaped wire and the San Francisco Bay Bridge will be using this S-shaped wire as well. However, the Akashi Kaikyo Bridge used round wire.5

On recent bridges, lead rich paste has been replaced by zinc rich paste for improved performance and environmental concerns. Elastomeric base systems with metallic zinc filler have been used for long-term flexibility of the coating and hence greater weather resistance.

5 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

A 4.3.2 Plastic Cover

An improved waterproofing system utilizing elastomeric cable wrapping (Figures A8 and A9) was developed in the 1960s.6 The idea was to combine the use of superior waterproofing membranes and coatings into a system that would totally encapsulate and protect the cable from the outside environment. The membrane weighs about 20% of the steel wire wrapping. The total cost of the elastomeric system is also lower than the steel wire wrapping. The Newport Bridge in Rhode Island and then the William Preston Lane, Jr. Memorial Bridge over the Chesapeake Bay at Annapolis, Maryland, used this type of elastomeric wrapping. The Akashi Kaikyo Bridge used waterproofing elastomeric wrapping on top of the steel wire wrapping.3

A 4.3.3 Dehumidification

Experience has shown that a good waterproofing wrapping alone cannot prevent moisture from getting into the cable prior to wrapping and moisture from seeping into the cable through cracks in caulking around the cable band, saddles and splay points at the anchorages. During the 1980s, a cable dehumidification method was developed in Japan to remove moisture and maintain the interior of the cables at or below a relative humidity of 40%. The first of such systems was installed at the Akashi Kaikyo Bridge. The method was developed based on the premise that steel will not rust when the relative humidity is lower than 40%. Dry air is continuously pumped into the cable to drive out moisture inside the cable. In the first few months, the moisture already inside the cable is removed. The dehumidifier will operate periodically to remove any moisture that seeps into the cable and keep the interior space of the cable dry. Several existing and new suspension bridges have since adopted the dehumidification system to protect their main cables from corrosion.7

At this time, the dehumidification method, and a closely related method utilizing inert gas to displace moist air, may still be considered to be in their development stage. In contrast to the traditional corrosion protection method, dehumidification and the inert gas system require machinery to work continuously over the life of the bridge. Enforcement of the maintenance plan, reliability of the mechanical / electrical operating system, and the effectiveness of the system in protecting all points along the cable will have to be tested by time.

A 4.4 Center Cable Tie

When the first Tacoma Narrows Bridge collapsed, its main span was vibrating in the second torsional mode. The low point of the main cable at mid span moved back and forth along the bridge, while the deck exhibited minimal displacement at that location. The observed cable movement is widely regarded as an amplification of the second torsional mode of the structure which contributed to its demise. After the Tacoma collapse, all new suspension bridges have been installed with center cable ties (see Figure A10), which eliminated the longitudinal cable movement at the mid span locations, increased the stiffness of the suspension bridge system and improved its aerodynamic behavior.

There are different combinations of center cable tie and deck fixity that will affect the stiffness, torsional mode, and forces in the cable tie. It can range from having fixity at the center cable tie only, as in the case of the Carquinez Bridge, to having the deck fixed at the center tie, the towers, as well as the anchorages. Fixing the deck at towers by shock transmission units is another possibility. The Bronx-Whitestone Bridge was one of the first suspension bridges to have

6 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

a center cable tie. The center cable tie photo of Hoga Kusten Bridge in Sweden8 and sketch of the Newport Bridge9 (see Figure A11) demonstrate two different details.

The forces sustained by the center cable tie are mainly associated with asymmetric live load loading and vibration caused by wind. The demands on the center cable tie may be computed by half main span loaded with live load or from the wind loads developed from buffeting analysis.

A 4.5 Wire Splices

The aerial spinning method uses a very long and continuous length of cold drawn wire to form each of its strands. A ferrule, or mechanical splice, is used to connect successive lengths of wire delivered by the reels (see Figure A12). A special clamping tool is used to clamp the ferrule to form the splice. This type of splice looks like a small steel tube with a serrated grip on the inside.

During cable inspections, there is a frequent need to collect wire samples from the subject cable. After extracting a sample, the two ends of the cut wire will be spliced with turnbuckle ferrules which essentially comprise of two ferrules connected with a small turnbuckle used to restore the tension lost between the cable clamps when the wire was cut.

Ferrule connections should be tested to demonstrate that the splice is stronger than the wire itself. In other words, the wire, instead of the splice, is expected to fail first in a pulling test. The ferrule also should be tested for fatigue, and the test should show that the ferrule does not reduce the fatigue resistance of the wire.

A 4.6 Main Cable Strengthening

The factors of safety (F.S.) for new suspension cables vary from 4.1 on the Williamsburg Bridge in New York City to 2.2 on the Akashi Kaikyo Bridge in Japan.10 In general, the newer bridges are designed with a lower factor of safety. The reasoning behind lowering the FS includes (a) better understanding of materials and better quality control, (b) better computational tools to estimate the stress level in the cable, and (c) improved corrosion protection systems.

When a bridge cable is found to be corroded or additional load must be carried by the bridge, the F.S. of the main cables drops. When the F.S. drops to an unacceptable level, strengthening of the existing cables or replacement of the cables will have to be considered.

One way of strengthening the suspension cables is by adding new cables to carry the additional load or to share the existing load (see Figure A13). The cables of the Tagus River Bridge in Lisbon, Portugal, were strengthened to carry two more traffic lanes and two heavy rail tracks.11 A pair of new cables was installed above the existing cables to carry the additional load.

A 4.7 Main Cable Replacement

In cases where strengthening is not a practical option, replacement of the existing cables may be considered. Replacement of the main cable will be very expensive because new anchorages will have to be constructed.

Cable replacement requires new cable anchorages, and new towers or tower modifications (see Figure A14). A suspension bridge is very flexible. Any change of load or geometry of the structure will cause large stress variation. Load transferring from the existing cables to the new cables is a complicated process. It requires detailed analysis of every stage to ensure that no member of the bridge will be overstressed during the load transferring operation.

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One of the advantages of cable replacement as compared to cable strengthening is that new cables can last much longer than the existing ones, even when they are rehabilitated to some extent during reconstruction. New technologies, especially improved corrosion protection methods, should be used for the new cables.

A 5.0 Suspension Cable Anchorage

There are two types of suspension cable anchorage (see Figure A15): • Anchoring to Rock • Gravity

Where shallow competent rock is available, anchoring the main cables directly to the rock may be an economical solution. The volume of the rock mobilized should be large enough to resist all loads required.

There are two ways to engage the rock:

1. A concrete block is anchored to the rock by rock anchors. Cable strands are then attached to the concrete block which would carry the cable load to the rock.12 The reliability of the rock anchors is therefore of utmost importance. Hennepin Avenue Bridge12 over the Mississippi River in Minneapolis and Hoga Kusten Bridge in Sweden13 use this type rock anchorage.

2. The friction type of rock anchorage would directly transfer the cable pulling force from the concrete block to the rock through friction. A large size cavity is excavated in the rock and the cavity is filled with concrete. Tie rods are cast in the concrete that will, in turn, connect to the cable strands. The Middle Fork Feather River Bridge in California used this type of rock anchorage.14

Where sound rock is not available, a large concrete anchorage is required to resist the vertical and horizontal pull of the suspension cables. The anchorage needs to be made large enough to resist the horizontal shear under the anchorage, the overturning moment and any uplift force generated by the cable force. The Jianyin Bridge over Yangtze River in China15 and the Bronx- Whitestone Bridge in New York City used this type of anchorage.

A 5.1 Corrosion Protection Inside the Anchorage

The wires of the main cable are separated into groups of strands inside the cable anchorage (Figure A16). The strands are attached to tie rods which are in turn anchored to concrete or rock. Access is required to perform inspection, maintenance, and future rehabilitation of the wires and tie rods. While providing access would often mean that the chambers have to be made larger at an increased construction cost, experience shows that some degree of corrosion and, therefore, maintenance / rehabilitation activities is inevitable and that proper access is a worthwhile investment for the longevity of the cables and the anchorages.

The interior of the chamber formed in rock or concrete is always cool. Condensation on the steel wire and tie rod surfaces is a permanent condition at most bridge sites. Since moisture is the main source of steel corrosion, it is a matter of time before the galvanizing protection over the wire is consumed and steel corrosion begins. In recent years, bridge engineers began to

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consider the prospect of controlling corrosion in steel by keeping the relative humidity within the anchorage chamber below 40%. Dehumidification inside the anchorage chamber (see Figure A16) is now an option for corrosion protection that should be given serious consideration during design. The surface of the concrete or rock should be sealed to reduce the power requirement for the dehumidification. One alternative to sealing the concrete or rock surfaces, and to dehumidifying the entire chamber, is to wrap the wires and tie rods and provide even more localized dehumidification. Assurance of continuous power supply and maintenance of the dehumidifier is required to make this option feasible.

A 5.2 Self-Anchored Suspension Bridges

For self anchored suspension bridges, the suspension cables are anchored to the deck. While large and costly anchorages can be avoided, the superstructure has to be designed to resist a large compression force (see Figure A17). This type of cable anchored bridge is called self-anchored suspension bridge. Their span range is typically much shorter than the conventional suspension bridges. The Three Sisters Bridges crossing the Allegheny River in Pittsburgh are the earliest examples (1924-28) of self-anchored suspension bridges in the US.16 Currently the largest self-anchored suspension bridges are the Konohana Bridge in Japan17 and the Yongjong Grand Bridge in South Korea; both have 300 m main spans.18, 19 The new San Francisco Bay Bridge will be the latest self-anchored suspension bridge in the US.20

The stiffening girders and deck of a self-anchored suspension bridge are under high compression as in the case of a cable-stayed bridge. A catenary suspension cable is very low in stiffness in all directions. Therefore, global buckling of the suspended superstructure in compression needs special attention. By contrast, the straight stay cables of a cable-stayed bridge provide significantly greater stiffness to the superstructure.

A 6.0 Superstructure Erection

A 6.1 Erection Stage and Stress Analysis

A suspension bridge is generally a very flexible structure. Bending moments tend to be very small in the completed structure. However, the structure may be subjected to significant bending moments during construction, depending on the method and sequence of construction. For this reason, structural analyses of the construction stages become extremely important to ensure the structure will not be overstressed. See Figure A18 for an example of a potentially critical loading condition for a suspension bridge. Traditionally the construction means and methods are determined by the contractor. Therefore, it is usually the contractor’s responsibility to perform the construction stage structural analyses based on equipment loads and positions, and erection sequence. Frequently, the Contractor has to choose between taking more erection steps to avoid overstressing the structure or investing in temporary strengthening of the structure to shorten the erection duration. Aerodynamic stability is an important aspect of the contractor’s stage analyses that can sometimes only be maintained by providing external systems of stabilization, which have to be modeled in the stage analyses. By-products of the stage analyses include fabrication lengths of cables and suspenders, cambers for fabrication of structural members, and deflections that the contractor will need to monitor during erection of the partially completed structure.

Although it is the contractor’s responsibility to develop the detailed construction sequence and accompanying stage analyses, the Designer is generally required to perform sufficient studies for at least one workable erection scheme that will be shown on the contract plans. 9 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

A 6.2 Erection Sequence

Suspension bridge construction usually starts with the towers and anchorages. Installation of the suspension cables and suspenders will follow. The last major operation is the erection of the deck segments. During the installation of the initial few segments, the low stiffness of the cables exhibits itself through significant deflection where the segments are hung. In order to not overstress the deck, the segments are temporarily pinned together to allow them to follow the exaggerated sag of the cable without taking stresses. As more segments are erected, the deck segments start to come to a smooth curve that closely resembles the design profile. At this time, the segments may be permanently joined together by bolting or welding.

Segment erection may start from the mid span or from the towers (see Figure A19). The erection sequence typically reflects the constraints at the site, associated delivery method(s) and logistics, schedule, and risk of being exposed to destabilizing wind. Other factors include the extent to which the cables are capable of resisting the bending stress caused by the concentrated loading of the initial segments. Most deep stiffening trusses are erected from the towers, while the steel box segments, being more streamlined, have been erected in many cases from the mid span.

A 6.3 Saddle Offsets at Top of Towers

Saddle Offsets (sometimes referred to as Setbacks) are often required for suspension bridges. During erection of the superstructure, the top of the tower and the cable saddle that is temporarily fixed to the tower top will tend to deflect toward the main span. In order to avoid permanent moment in the tower(s), saddles are often set back away from the direction of the tower deflection by an amount equal to the deflection. As the superstructure loads are placed, the top of the tower is jacked back to plumb against the cable saddle (see Figure A20). The timing and the number of intermediate jacking repetitions required would depend on the strength of the tower against construction demands. The Golden Gate Bridge21 and the Askoy Bridge in Norway22 used this method.

An alternative to this method involves pulling the tower top away from the direction of the expected deflection before the cable installation starts (see Figure A21). In this case, the saddle is permanently fixed to the centerline of the tower. As the superstructure is being erected, the pull on the tower is gradually released. Ideally, the tower is plumb without any bending when construction of the bridge completes. The Great Belt Bridge23, second longest bridge in the world, and the Carquinez Bridge in California24, used this alternative method.

The stress level in the tower column and the stability of the tower should be checked for all critical loading cases, including construction stages. For the same amount of movement on the top of the tower, a tall tower will generate less bending moment and, consequently, lower stress. Therefore, the pulling method is generally more appropriate for a tall slender tower. Offsetting the saddle tends to work better for a short and stubby tower.

It is common practice for the Designer to specify the expected method to use by showing such a method schematically in the construction plans. In order to not place undue restriction on the contractor’s operations, the contractor is commonly made responsible to determine and propose the required timing and frequency of these jacking operations.

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A 6.4 Creep and Shrinkage Effects with Concrete Towers

In addition to elastic shortening, a concrete tower will shorten due to creep and shrinkage. Most of the shrinkage will take place within the first year after the concrete is placed. For a long span suspension bridge, where the construction duration is generally long, most of the shrinkage would have occurred prior to the closure of the superstructure. Creep is a more long term phenomenon that is triggered by loads. Therefore, erection of the deck will initiate creep in the tower that will result in discernable shortening of the tower for around a decade. Creep is not known to be a prohibiting effect for suspension bridge designs, but it is an effect that should be evaluated. At the least, creep will cause slope break in the roadway profile at the towers and anchorages (see Figure A22). The details of the connections have to be designed to accommodate the angle changes.

A 7.0 Suspenders and Cable Bands

A 7.1 Suspender Materials

Galvanized steel wire rope and prefabricated galvanized steel strands are used most frequently for long span suspension bridges (see Figure A23). Suspenders comprised of parallel wires, parallel strands, as well as lock-coil wires have also been used when attachments to cable bands are located below the cable. Older suspension bridges used wire rope for its flexibility in bending. This flexibility allows the rope to be wrapped around the main cable band without overstressing the wire rope. This flexibility also makes the rope tolerant of bending caused by bridge movements. However, the paint on the wire rope tends to crack due to the twist effect. And as the saddle area on top of the cable band also collects moisture and dirt, corrosion at the top of the saddles is a common problem that is not simple to resolve. Inspection and painting of the long suspenders adjacent to the tower are particularly difficult and expensive.25

Prefabricated galvanized strand (PWS) suspenders are usually encased in PE pipes or other types of sheathing to provide additional layer(s) of corrosion protection. The PE pipe does not require painting and there is no saddle that can cause corrosion. Therefore, the PWS is widely regarded as the preferred choice for suspenders.

A 7.2 Suspender Fabricated Length and Adjustments

The main suspension cable drapes naturally in a catenary depending on the cable length and load. Adjusting the lengths of the suspenders is the means to bring the deck to the design profile, which can be computed independently of the cable profile. The suspenders for a long span suspension bridge are usually long and their elastic elongations due to dead load tension cannot be ignored. The suspenders should be fabricated to their unstressed lengths (see Figure A24). The suspenders will stretch to their design lengths after the dead load is fully loaded.

Due to the tolerance of the fabricated suspender length, provisions for length adjustment** should be provided. The adjustment is usually performed during erection when an obvious error is noted. However, consideration should be given to make such provisions a permanent feature that can be used in the future for possible suspender replacement. A suspension bridge is a highly indeterminate and flexible system. There are interlocking forces (moment, axial force and shear) in all members of the bridge. The adjustment of length in just one suspender after the bridge is complete would trigger changes in forces and geometry in the entire bridge. Therefore,

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using suspender adjustments to achieve target forces in the structure after completing span closures is generally not recommended.

** Note: Tolerance will vary depending on the type of materials used for the suspenders. Suspenders with twisted wires may warrant more generous provisions and will require more care during socketing, storage, transport and erection because rotation (or untwisting) of the rope will result in the lengthening or shortening of the suspender. Marking a paint strip on the entire length of the rope when it is measured is one way to detect unwanted rotation.

A 7.3 Cable Bands

The main cables are connected to the suspenders by means of cable bands. Cable bands are typically made out of cast steel in two cylindrical halves that are clamped by high strength bolts. The wall thickness of each half is relatively thin in order to provide flexibility that facilitates conformance on the inside surface to the diameter of the compacted cable. When used with a loop type suspender, the outside surface is usually grooved to guide the suspender rope around the main cable. The suspender loop adds to the clamping force provided by the high strength bolts. The cable bands can also be shaped in their lower section into gusset plates and provided with pin holes that will engage suspenders equipped with sockets.

Since the cable slope varies along the span length, the suspenders are usually not perpendicular to the cable and, as a result, the suspenders tend to pull the cable band downward along the cable. High strength bolts should be designed to provide sufficient clamping force, and hence friction, to prevent the cable band from slipping. The resistance force against slippage is calculated in a fashion similar to designing slip-critical bolted connections. The friction coefficient between the wires of the main cable and the cable band material (steel or cast steel) is 30% at ultimate. Typically a safety factor of two (2) is used by the allowable stress method. In the case of grooved cable band, additional clamping forces can be considered from the looping action that the suspenders exert around the cable band.

The tension in the high strength bolts tends to drop over the years as the cable changes shape due to additional compacting effects created by the cable bands. The bolt tension needs to be verified particularly when signs of slippage are visible. Experience shows that over-tightening of the high strength bolts is not a viable way to safeguard against slippage of the cable bands. Excessive clamping force may simply accelerate the creep effect mentioned above.

Cable bands should be painted and sealed all around with provisions for drainage at the low side of each cable band.

A 7.4 Suspender Sockets

Traditional suspender sockets are filled with zinc to transfer load (see Figure A25). Epoxy filling was developed in the past few decades to improve the fatigue performance of this type of socket. Neither of these socket types is fire resistant. Since the sockets are located invariably near the deck, truck or car fire will have a high probability of causing failure in the zinc or epoxy filled sockets long before the steel wires in the suspender would fail. A new cement based filling has recently been developed which promises to not fail before the suspender wires break under high temperature.

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Suspender sockets should be tested for fatigue. The current practice is to have the designers develop project specific test programs. See Section F of these Guidelines, which summarizes the state of practice.

A 7.5 Suspender Stress

Where wire rope is used as suspenders, the bending stress associated with wrapping the rope around the main cable should be considered in the design. The size of the suspender wire rope should be selected based on the allowable bending radius specified by the wire rope manufacturer and the minimum radius in the cable band, considering the diameter of the main cable plus the thickness of the cable band.

The deck moves relative to the main cable, in both the longitudinal and transverse directions, under live load, temperature variation, wind induced vibration and other infrequent loads (see Figure A26). As a result, suspenders and their connections may be susceptible to significant bending stresses. In service conditions, the maximum bending moment and highest stress is likely to occur in the shorter suspenders where a given relative movement will translate into greater curvatures. The Akashi Kaikyo Bridge is one structure where extensive stress checking was conducted. The result of the stress checking required spherical bearings to be installed to the middle length suspenders.25 Spherical bearings may also be required due to demands imposed during the construction stages. Second order effects can be taken into account when analyzing combined axial and bending stresses of suspenders, which will greatly reduce stresses due to bending.

A 7.6 Suspender Oscillation in Wind

The suspender may oscillate itself in the wind independent of the main cable. It is mostly caused by vortex shedding. This type of oscillation needs to be suppressed, because the vortex shedding induced oscillation can lead to fatigue failure of the suspender. Reducing the free length of the suspender by installing spacers is the most commonly used method (see Figure A27).

The Akashi Kaikyo Bridge installed spacers with high damping rubber that was designed to suppress the oscillation induced by vortex shedding. However, the spacer/damper broke during a typhoon. Subsequent investigation suggested that wake-induced flutter was the cause. The suspenders are now wound up with 10 mm diameter spiral rope. After introducing this countermeasure, neither wake-induced flutter nor vortex-induced oscillation has been observed.26

A 8.0 Superstructure Details

A 8.1 Torsional Box

After the Tacoma Narrows Bridge collapse, bridge engineers recognized the importance of the aerodynamic stability of long span bridges. Studies revealed that the onset of flutter, the phenomenon that caused the failure of the Tacoma Narrows Bridge, can occur at a relatively low wind speed – also referred to as critical wind speed. Therefore, current practice would require that long span bridges be designed to achieve a critical wind speed that is much higher than the design wind speed.

A structure with low torsional mode frequency will generally exhibit a correspondingly low critical wind speed for flutter. The H-shaped superstructure of the collapsed Tacoma Narrows Bridge, 13 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix A

comprised of two vertical edge girders and transverse floor beams, is now widely recognized as a section that is economical in construction but of relatively low torsional stiffness. Its application in very long span designs is therefore also limited.

The most commonly used alternative that exhibits much greater torsional stiffness is the box shape superstructure. The box can be formed by two stiffening trusses linked across the top and bottom chords by diagonals or by closed plate elements made of steel or concrete (see Figure A28).

A 8.2 Steel Box Segment Shipping

A steel box superstructure needs to be prefabricated in segments in the shop and then shipped to the bridge site for erection. The box segments are usually supported by 4 or more blocks. The unevenness of the supports may cause significant stress in the box. Furthermore, when the box segments are loaded on a barge, ship or other vehicles, the movement or deformation of the vehicle may cause significant stress in the box as well.

The design engineer should consider these effects and design into the steel box some degree of robustness that would enable it to resist a certain amount of distortion associated with storage and shipping.

A 8.3 Wind Locks

Transverse wind load on the deck needs a load path to the towers and the anchorages. A wind tongue, or wind lock, is required to provide the load path (see Figure A29). While the deck is supported by flexible cables, the roadway profile and bridge alignment has to be fixed at the tower and the anchorage in vertical and transverse directions. Constraint in the vertical direction is provided by rocker links (or bearings capable of taking uplift), while constraint in the transverse direction is provided by the wind tongue. The wind tongue must allow longitudinal movement of the deck as its length changes due to temperature variation. The deck will rotate about the vertical and horizontal axes, while its longitudinal axis rotation is limited by the pair of rocker links. The wind tongue needs to be free in 4 out of the 6 degrees of freedom, except for lateral translation and rotation about its longitudinal axis.

Usually the shear blocks are set on the tower and anchorage. The tongue is attached to the deck superstructure. The members, which transfer the wind load from the deck to the tongue, are main load carrying members.

A 8.4 Rocker Links

The deck suspended from the cables is free to move in 3 dimensions. However, the deck must be on alignment and grade at the towers and anchorage locations. Wind tongues keep the deck from moving laterally. A pair of rocker links keeps the deck from moving vertically and rotating about the longitudinal axis. The links allow the deck to move longitudinally relative to the tower and anchorage (see Figure A30).

A link often comprises a steel element (usually a steel beam) with a pinned connection at each end. The link can be placed above or below the deck. The link above the deck is normally in tension, while a link placed below the deck would operate in compression. In addition to designing the link for axial loads, the rocker link should also be designed for a certain moment capacity to resist the increased moment generated by the old corroded pins. Designing a stronger rocker link will either force the pin to rotate or keep the rocker link bending stress level

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below allowable. A lubrication port should be designed into the pin to allow for future maintenance and to keep the friction low.

Alternatives to the rocker link may include the use of conventional bearings, or making the deck continuous and eliminating any rigid support at the towers. While there are numerous possibilities for alternatives, the previous discussions on rocker links point to the principal functional aspects that designers must satisfy.

A 9.0 Vertical Suspenders vs. Inclined Suspenders

Vertical suspenders are commonly used in traditional designs. The suspenders link the cable and deck to move in unison in the vertical direction. They carry the vertical load from the deck.

The suspenders may also be inclined and arranged in a net pattern (see Figure A31). The net system not only carries the vertical load, but also restrains the relative movement between the cable and the deck in the longitudinal direction. It could greatly increase the overall stiffness of the bridge, which is beneficial to its aerodynamic behavior. Since inclined suspenders tend to see greater stress variation than vertical suspenders, special attention should be paid to fatigue design.

A 10.0 Vertical Cable vs. Inclined Cable

Cables would normally hang in their natural vertical catenary plane.27, 28 However, a number of bridges have their cables pulled laterally by the suspenders to form inclined cable/suspenders planes.17, 18, 19 As the two cables are pulled apart laterally, the deck has to resist the transverse components of the suspender load in compression (see Figure A32).

This inclined cable design increases the torsional stiffness of the bridge which would improve the bridge’s aerodynamic performance. However, construction of the bridge will be more difficult. A pair of cables would first be erected in the vertical plane. During the erection of deck segments, the cables will be pulled apart gradually. This is a delicate operation, because all connections will be fabricated to complex geometries that must be tracked and controlled in three dimensions at each stage of erection. Additionally, secondary stresses should be considered in the main cables and the cable bands due to the progressive loading and twisting of the cables during erection.

The tower saddles and splay saddles are set in their final inclined position. They must be detailed to accommodate the vertical hang of the cables before the deck is in place. The challenge is in developing details and procedures that will protect the wires from damage while the cable slides horizontally in the saddle during construction.

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Figure A 1 Side Span Types

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Figure A 2 Operation Principle of Aerial Spinning

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Figure A 3 Prefabricated Parallel Wire Strand

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Figure A 4 PPWS Installation

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Figure A 5 Main Cable Compactor

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Figure A 6 Main Cable Compactor

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Figure A 7 Cable Compaction Void Ratio

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Figure A 8 Elastomeric Wrapping

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Figure A 9 Corrosion Protection of Main Cable

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Figure A 10 Center Tie Detail

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Figure A 11 Center Tie Detail

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Figure A 12 Wire Splice Detail

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Figure A 13 Main Cable Strengthening

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Figure A 14 Main Cable Replacement

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Figure A 15 Anchorage Types

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Figure A 16 Corrosion Protection Inside Anchorage

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Figure A 17 Suspension Bridge Types

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Figure A 18 Example of Erection Load Induced Bending Moments

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Figure A 19 Superstructure Erection Methods

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Figure A 20 Cable Saddle

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Figure A 21 Temporary Cable Pull at Tower

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Figure A 22 Creep and Shrinkage Effect of Concrete Tower

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Figure A 23 Suspender Materials

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Figure A 24 Suspender Fabrication Length

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Figure A 25 Suspender Socket

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Figure A 26 Relative Movement between Suspenders

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Figure A 27 Suspender Oscillation in Wind

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Figure A 28 Torsional Box

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Figure A 29 Wind Lock

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Figure A 30 Rocker Link Detail

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Figure A 31 Suspender Geometry

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Figure A 32 Cable Geometry

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REFERENCES

1 Okukawa, Atsushi, Shuichi Suzuki, and Ikuo Harazaki. “Suspension Bridges.” Bridge Engineering Handbook. Ed. Wai-Fah Chen and Lian Duan. 2000.

2 Spoth, Tom and Joe Viola. “A New Suspension Bridge Across the Historic Tacoma Narrows.” Structure Magazine. National Council of Engineers Associations. February 2009.

3 “In Bay Foundation/SAS Superstructure Constructability Workshop.” SFOBB East Span Seismic Safety Project. California Department of Transportation. 20 November 2002: Slide 34.

4 Sun, John, Rafael Manzanarez, and Marwan Nader. “Suspension Cable Design of the New San Francisco-Oakland Bay Bridge.” Journal of Bridge Engineering. January/February 2004: 101-106.

5 Kitagawa, Makoto, Shuichi Suzuki, and Motoi Okuda. “Assessment of Cable Maintenance Technologies for Honshu-Shikoku Bridges.” Journal of Bridge Engineering. November/December 2001: 418-424.

6 “NeoCablewrap: Elastomeric Cablewrap System for Suspension and Cable-stayed Bridges.” Republic Powdered Metals, Inc.

7 Bloomstine, Matthew L. and Ove Sorensen. “State-Of-The-Art Main Cable Corrosion Protection by Dehumidification.” 3rd New York City Bridge Conference. Bridge Engineering Association. 12- 13 September 2005.

8 Trimbath, Karen. “Dehumidification System Implemented on Swedish Bridge.” Civil Engineering. 12 December 2005: 27-28.

9 Hedefine, Alfred and Louis G. Silano. “Newport Bridge Superstructure.” Journal of the Structural Division, Proceedings of the American Society of Civil Engineers. November 1971: 2653-2678.

10 Guidelines for Inspection and Strength Evaluation of Suspension Bridge Parallel Wire Cables. NCHRP Report 534. Transportation Research Board.

11 “High Wire Live Action.” New Civil Engineer. 21-28 August 1997: 26-29.

12 Hughes, William R. The Design of a Suspension Bridge Anchorage System. Bridge/Tunnel Service Group of Howard Needles Tammen & Bergendoff. 1990.

13 Jorgensen, Gerner Rorso, Anton Petersen, and Lars Pettersson. “Hoga Kusten Bridge, Sweden.” Structural Engineering International. February 1999: 106-108.

14 “General Plan & Anchor Details 1, Middle Fork Feather River Bridge.” State of California Department of Water Resources.

15 Zhou, Shizhong. “Construction of the Jiangyin Yangtze Suspension Bridge.” Structural Engineering International. January 2004: 30-31.

16 Ochsendorf, John A. and David P. Billington. “Self-Anchored Suspension Bridges.” Journal of Bridge Engineering. August 1999: 151-156.

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17 Kamei, M., T. Maruyama, and H. Tanaka. “Konohana Bridge, Japan.” Structural Engineering International. February 1992.

18 Cho, Choong-Young, Seung-Woo Lee, Soo-Young Park, Myeongjae Lee. “Yongjong Self- anchored Suspension Bridge.” Structural Engineering International. January 2001: 21- 23.

19 “Korea’s New Gateway.” International Construction. May 1994: 32-33.

20 Sun, John, Rafael Manzanarez, and Marwan Nader. “Design of Looping Cable Anchorage System for New San Francisco-Oakland Bay Bridge Main Suspension Span.” Journal of Bridge Engineering. November/December 2002: 315-324.

21 Mensch, Ernest C. The Golden Gate Bridge: A Technical Description in Ordinary Language. 1935.

22 Blom-Bakke, Lars and Jostein Hellesland. “The Askoy Suspension Bridge.” International Bridge Conference, Pittsburgh. 14-16 June 1993.

23 Reina, Peter. “For Historic Link, Team Focuses on Earlier Experience.” ENR. 13 May 1996: 24-28.

24 “State of Suspense.” Bridge Design & Engineering. Third Quarter 2002: 32-38.

25 Kondoh, Munenobu, Motoi Okuda, Kouji Kawaguchi, and Takefumi Yamazaki. “Design Method of a Hanger System for Long-Span Suspension Bridge.” Journal of Bridge Engineering. May/June 2001: 176-182.

26 Kashima, Satoshi, Yukikazu Yanaka, Shuichi Suzuki, and Kunihisa Mori. “Monitoring the Akashi Kaikyo Bridge: First Experiences.” Structural Engineering International. February 2001: 120-123.

27 Pylons, Anchor Blocks & Main Cables of the Suspension Bridge. Storebaelt.

28 Reina, Peter. “China’s Newest Suspension Bridge is its Longest.” ENR. 1 November 1999: 23.

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TABLE OF CONTENTS

B 1.0 Proportions and Sizing of Cable-Stayed Bridges ...... 1 B 1.1 Deck Characteristics ...... 1 B 1.2 Tower Height Above Deck ...... 1 B 1.3 Tower Sizing ...... 1 B 1.4 Superstructure Depth ...... 2 B 1.5 Main Span to Back Span Ratio ...... 2 B 1.6 Stay Cable Sizing...... 2 B 1.7 Feasibility of Cable-Stayed Bridge on Curved Alignment ...... 3 B 1.8 Deck Vertical Support at Tower ...... 3 B 1.9 Longitudinal Restraint at Tower ...... 3 B 1.10 Lateral Restraint at Tower ...... 4 B 2.0 Minimum Design Force in Stay Cables ...... 4 B 2.1 In-Service Condition ...... 4 B 2.2 During Construction ...... 4 B 2.3 During an Extreme Wind Event...... 4 B 2.4 During a Seismic Event ...... 5 B 3.0 Cable Vibration Control ...... 5 B 4.0 Cable Corrosion Protection ...... 5 B 5.0 Cable Anchorage ...... 6 B 5.1 Cable to Girder Connections ...... 6 B 5.2 Cable to Tower Connections ...... 7 B 6.0 Construction and Erection Methods ...... 7 B 7.0 Erection Stage Stress Analysis ...... 8 B 8.0 Construction Geometry Control ...... 9 B 9.0 Uplift at Anchor Piers ...... 9 B 10.0 Deck Replacement ...... 10

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

LIST OF FIGURES

Figure B 1 Cable Vibration...... 12 Figure B 2 Cable Corrosion Protection ...... 13 Figure B 3 Cable to Girder Anchorage ...... 14 Figure B 4 Cable to Girder Anchorage ...... 15 Figure B 5 Cable to Tower Connection ...... 16 Figure B 6 Cable to Tower Connection ...... 17 Figure B 7 Construction Erection Method ...... 18 Figure B 8 Construction Geometry ...... 19 Figure B 9 Tie Down ...... 20 Figure B 10 Deck Replacement ...... 21

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

APPENDIX B – CABLE-STAYED BRIDGES

Topics selected are those deemed likely to be of immediate application for the readers of these Guidelines. Materials presented are available in the literature and provided to supplement the Guidelines as a convenient source of reference. An overview1 is suggested for the readers who seek general and historic perspectives of this bridge type.

B 1.0 Proportions and Sizing of Cable-Stayed Bridges

B 1.1 General Characteristics

The stiffness of the tower, the size of the back-stay (anchoring) cables, and the inclination of the stay cables are the principal factors that affect the stiffness of the bridge superstructure, often simply referred to as the deck. Cable sag reduces cable stiffness. However, the reduction is normally negligible except when the main spans exceed 2000 feet.

Bending moments associated with dead loads can be manipulated in the longitudinal girders by cable adjustments. As a result, designers often adjust cables to optimize bending moments in the deck. Generally, the deck behaves as a beam on elastic foundations. Therefore, post-tensioning the concrete deck will result mainly in primary effects, i.e. axial post-tensioning force x eccentricity, and very minor secondary effects except near hard supports at the anchor piers or towers, if bearings exist.

B 1.2 Tower Height Above Deck

The height of the tower above the deck is determined by the location of the highest stay cable connection. Generally, the higher the cable connection point, the lower the compression in the deck. The compression in the deck is the accumulation of the horizontal component of the cable tensile forces. The height of the tower is usually approximately 22% of the main span length. However, architectural requirements may sometimes require that the tower height be set significantly higher or lower than this norm.

B 1.3 Tower Sizing

The following considerations usually determine the tower cross section: 1. Structural stability and stress – H-shape, needle-shape, and A-shape are three examples of the more popular configurations among tower designs. Wider cross-sections are selected when the design calls for great laterally unsupported length. Likewise, a larger cross-section is generally suitable when the tower axial and bending loads are high. The H-shaped tower is the least favorable in terms of providing aerodynamic stability, but tends to be more economical for construction.

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2. Cable to Tower Connection – Unless dictated by exceptional designs, the cable size is generally limited to the capacity of the testing facilities as well as the sizes supported by US cable suppliers. The highest cable in the tower is usually also the largest one. This rule may not apply for cables in the harp configuration. Except when a design calls for cables to pass continuously through the tower, the vertical cable spacing and lateral dimensions in the tower should be selected to accommodate the cable anchors and cable installation requirements. The number of cable planes and the flare of the cable planes are additional factors that would control the tower lateral dimension. 3. Access – Adequate space should be provided to allow access for construction, inspection and maintenance personnel. The space should also allow for material and equipment to be delivered from the deck or ground level to the cable anchors for repair and replacement work. 4. Architectural Design – While there should be no limitation on selecting tower cross-sections that are architecturally pleasing, it should be noted that overly complex cross-sections may seriously impact construction cost and construction schedule, because tower construction is usually on the critical path of cable-stayed construction.

B 1.4 Superstructure Depth

Modern cable-stayed bridges are required to be designed for cable loss and cable replacement load cases. Due to these requirements, the cable spacing along the deck is sometimes reduced in order to allow the girder to span the missing cable. For bridges with dual planes of cables where one cable is assumed lost, a 6-foot deep girder is usually adequate when cables are spaced 40 feet to 45 feet apart. Box girders supported by a single cable plane would have to be approximately 10 to 12 feet deep, and the cable spacing should be reduced to about 20 feet. The depth of the box girder is most likely governed by the torsional strength requirement rather than the cable replacement or cable loss cases.

B 1.5 Main Span to Back Span Ratio

Unless otherwise constrained, most cable-stayed bridges have their back spans laid out to be approximately 45% of the length of the main span. The benefit of staying close to this span ratio is that the anchor (or back stay) cables can be kept taut all the time. This span ratio will likely cause uplift at the anchor pier that must be resisted by a tie-down or counterweight.

B 1.6 Stay Cable Sizing

When sizing cables comprised of 7-wire parallel strands, an allowable stress to use for dead plus live load is 45% of GUTS (Guaranteed Ultimate Tensile Strength). Dead load is approximately 80% of the combined dead plus live load for median span bridges. The dead load percentage is higher for longer span bridges.

Dead load reactions at all cable support points can be estimated by treating the deck as a continuous beam. These reactions would be the vertical component of the cable force. The cable forces can be estimated by considering the cable inclination. The rough sizes of the cables 2

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may be computed by dividing those cable forces by 40% GUTS. The rough size should be refined by additional analyses.

B 1.7 Feasibility of Cable-Stayed Bridge on Curved Alignment

A horizontally curved cable-stayed bridge is feasible. Many cable-stayed bridges have been built around the world with the superstructure on a single or double horizontal curvature. Nevertheless, most cable-stayed bridges are on a straight alignment for ease of construction and economy, with a few having their back spans on a mild horizontal curve.

A cable-stayed deck on curved alignments, vertically and/or horizontally, requires that attention be paid to stability issues especially during construction. A horizontally curved deck will also result in additional lateral load to the tower.

B 1.8 Deck Vertical Support at Tower

Three common types of deck support at the tower are (a) deck supported by vertical bearings, (b) deck supported by cables, and (c) deck constructed integral with the tower. As stated above, dead load bending moment in a cable-stayed deck can be estimated the same way as one would for a continuous beam. When the deck is supported by bearing or is integral with the tower, a sharp spike in negative live load bending moment can be expected at the centerline of the tower support.

B 1.9 Longitudinal Restraint at Tower

The cable-stayed system is relatively flexible in the longitudinal direction. There is a need to provide longitudinal restraints to minimize the longitudinal deck movement at the expansion joints. The deck can be longitudinally restrained against one tower and free at the other tower. The tower that restrains the deck will resist longitudinal force due to live load, wind load, earthquake load, temperature load and creep and shrinkage load. Most cable-stayed bridges adopt this bearing arrangement.

Some bridges have no longitudinal constraint at the towers. Instead, the longitudinal restraint is provided by the anchor piers. Since the anchor piers are typically more flexible than the towers, this arrangement tends to have lower stiffness in the longitudinal direction and, consequently, the frequency of the first longitudinal mode is lower. This arrangement has the potential of attracting less earthquake load overall with minimal influence to the aerodynamic behavior of the bridge.

If the deck is integral at the towers, effects from temperature variation, creep and shrinkage could become a major concern. Length change in the deck due to the temperature variation, creep and shrinkage pushes and pulls the tower along the bridge alignment, which would cause significant bending stresses in the tower and tower foundation. When the deck shortens, significant tension is likely to occur in the mid span of the deck. Design of the towers and their foundations may be governed by the increased axial force in the deck against the tower.

Some bridges utilize locked-up devices installed between the deck and an otherwise free standing tower. The locked-up device has very low resistance for slow movement but the resistance becomes high when the movement is fast. This allows the bridge deck to move freely under temperature load and creep and shrinkage, but lock up when fast motion of an earthquake

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occurs. The advantage is to allow the strong tower to participate in resisting the earthquake load, without interfering with the other slow movement caused by temperature, creep and shrinkage.

B 1.10 Lateral Restraint at Tower

Where the deck is free to move relative to the tower, a bumper is usually installed to transfer lateral wind load from the tower to the deck. The bumper will also transfer lateral earthquake loading from the tower to the deck. Bumpers installed at each side of the deck slab will transfer the wind and earthquake load directly from and to the tower. The bumpers are simple devices that can be easily replaced should a design level event cause damage to them. The bumpers should be installed with a gap that is large enough to accommodate the thermal expansion of the deck in the transverse direction, as well as any deformation (such as deck bowing) associated with temperature gradient in the structure.

B 2.0 Minimum Design Force in Stay Cables

A stay cable should, by design, be kept under significant tension for most of the service loading cases. However, under certain loading conditions, the tension may be significantly reduced - to near zero in the extreme cases. Some of the situations where such loading conditions may occur include:

B 2.1 In-Service Condition

The anchor cables connect anchor pier and tower. When the back span is loaded with live load, the tower is pulled towards the anchor pier and the tension in the anchor cables will decrease. The amount of reduction depends on the stiffness of the tower and the deck. For typical double tower cable stayed bridges, half of the main span is generally longer than the back span. This typical span arrangement would generally result in high tension in the anchor cables and make them less susceptible to becoming slack.

B 2.2 During Construction

Tensioning a cable reduces the tension in the adjacent cables. Depending on the flexibility of the girder, the initial stressing against the weight of the erected structure may be so low that the wedges in the stay anchorage would not adequately “bite” into the strands. It is generally the construction engineer’s responsibility to prevent this from happening during any stage of erection. Wedge restrainers should be used in case wedge slip is possible.

B 2.3 During an Extreme Wind Event

Buffeting effects on a structure during construction can potentially unload stay cables. Buffeting analyses or measuring cable tension in wind tunnel tests should be preformed to plan the proper preloading necessary to prevent unloading, in both permanent and / or temporary stays. In certain situations, wedge restrainers should be considered to ensure that the wedges at the cable anchors will never come loose.

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B 2.4 During a Seismic Event

When the deck experiences significant displacements during an earthquake, the tension in the cables will fluctuate. It is generally not a concern of the cable-stayed bridge design, because the cable-stayed bridge is very flexible and the first mode usually has a period longer than 3 seconds. But it may be a concern for short span cable-stayed bridges and bridges in high seismic areas. When complete unloading of a stay cable is predicted, wedge restrainers should be installed at the anchors to prevent wedges from coming loose.

For socket type cable anchors, stay cables with the potential of unloading should have anchors positively attached to the deck and tower to prevent dislocation.

B 3.0 Cable Vibration Control

The modern parallel wire or strand cables are inherently low in damping. The damping ratio is in the range of 0.05% to 0.5%.2 At this level of damping, the cables will be susceptible to both galloping and rain/wind-induced vibration. When the cable is excited by wind, deck movement, and / or earthquake, the cable will start to oscillate. Because of the low damping, the energy accumulated and the amplitude of the vibration will increase. Such vibration will bend the cable at its anchors which could cause fatigue failure of the cable.

Installing cable vibration dampers is the common method for increasing damping and limiting the vibration amplitude. Most dampers are installed at the lower end of the cable because of easy access for construction as well as subsequent inspection and maintenance (see Figure B1). There are many different types of dampers. The latest development would utilize high damping rubber, which claims to have no moving parts and promises only minor deterioration over time.3

Cables can also be tied together by a series of cross-ties.4 This will increase the fundamental frequency of the stays and reduce their vibration amplitude. It should be noted that the cross-ties are normally effective only in suppressing in-cable-plane vibrations, and have negligible effect on the out-of-plane (lateral) frequencies. The ties are usually made of wire rope, which has higher damping than the stay cables. Cross-ties are often pretensioned to avoid snap loading but may also be simply oversized. The use of cross-ties is a relatively low cost and reliable method for suppressing cable vibration. However, there are limitations to its application. It can be attached to grouted cables very easily but will require special details when dealing with ungrouted cables. It is noted that the latest trend is to use ungrouted cables. Also, for cable-stayed bridges having main spans longer than 1,500 feet, the attachment points of the ties may be too high above the roadway for crane access for installing and maintaining the cross-ties. As a result, cross-ties are not the first choice for very long span bridges, especially when ungrouted cables are used.

B 4.0 Cable Corrosion Protection

Worldwide, most cable-stayed bridges use high strength steel wires to form cables as main load carrying elements. In the United States 0.6” dia. 7-wire strands are used on most cable-stayed bridges.

Many of the first generation of cable-stayed bridges built in the 1960s have stay cable damage due to corrosion. Some cables have been replaced. Some bridges built in the 1970s and 1980s 5

FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

have had their cable stays replaced. At least one cable-stayed bridge in China was demolished because of cable corrosion. Stay cables of the Maracaibo Bridge in Venezuela have been replaced.5 The State of Louisiana has plans to replace the cables on the Luling Bridge.6

With proper protection and maintenance, steel and concrete structural members can generally last more than a hundred years, because of their bulk size and small surface area-to-volume ratio. Stay cables, on the other hand, utilize high strength steel wires as the main load carrying elements. With the high surface-to-volume ratio, the high strength steel wires can develop corrosion at a rapid pace once oxygen and moisture penetrate the protective barriers of the stays. This is the reason why corrosion protection has been given considerable attention by engineers involved with the design of cable-stayed bridges.

Galvanization is one of the best first layers of protection for steel wires. However, galvanized strands are not available in the United States. Since the “Buy America” law has in the past prevented the import of galvanized wires, most cable-stayed bridges in the US have used strands without galvanization.

Epoxy coating and cementitous grout was commonly used for corrosion protection in the 1980s. This system is no longer considered adequate, as the construction industry has moved toward multi-layer protective systems involving strands that are individually protected by grease or wax and polyethylene or polypropylene sheathing (see Figure B2). The sheathed strands would be ungrouted. At the anchorage, the protective PE sheathing is removed to allow the strand wedges to bite into bare steel wires. Cable suppliers tend to develop their own corrosion protection method for the cable anchor. The exposed steel strands are generally sealed and the voids are filled with grease or wax. This type of ungrouted stay and corrosion protection was first adopted on the Leonard P. Zakim Bunker Hill Bridge in 1998 and has been adopted by most cable-stayed bridges in the United States ever since. The current standards also require stringent testing for water tightness of multi-layer protective systems.

The concept of controlling the immediate environment around cables as a means of corrosion protection has been utilized in at least one bridge in the US. The Penobscot Narrows Bridge uses pressurized dry nitrogen to provide additional corrosion protection to epoxy coated strands.7

B 5.0 Cable Anchorage

New ideas for the cable to girder connections and cable to tower connections have been developed continually since the construction of the first cable-stayed bridges in the US.8,9 The following provide a brief description of the more popular approaches:

B 5.1 Cable to Girder Connections

Cable anchors can be located at the bottom, middle or top of the girder (see Figure B3). They can also be connected to the side of the steel edge girders; in this case, the design should account for the eccentricity of the cable force with respect to the girder.

To anchor the cable at the top of a steel girder, a girder web extension is required to receive the cable anchor. The extension may be bolted to the girder web or the web and extension can be cut from one piece of steel plate. The extension plate could also be welded to the top flange of the steel girder; in this case, the cable force would be transferred through the top flange plate to 6

FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

the web plate (see Figure B4). In the case where the extension plate is welded to the top flange, the cable force is transmitted from the web of the girder to the stay through tension across potential weak layers of the flange plate. Special X-ray examination is therefore required to identify any defects that may exist in the production of the base material as well as those resulting from the welding process. While this approach has been used on a few bridges, its application has been limited.

B 5.2 Cable to Tower Connections

A pair of cables can be anchored at the opposite faces of a tower. The opposing cable forces will compress the solid concrete tower and result in an effective way of exploiting the compressive strength of concrete. However, the cable anchors should be properly protected from the elements. Attention should be given to provide access for inspection and future maintenance work (see Figure B5).

Cables being supported by a saddle, or a series of saddles, at the top of the tower is another connection method. High transverse pressure to the strands at a saddle must be considered to prevent potential damage to the coating or sheathing of the strands. Individual saddles for epoxy coated strands were developed to keep the strand corrosion protection through the saddle area. Friction at the saddle has been reported to vary from 0.13 to 0.5. Design should be based on test results to ensure that the strands would not slip due to the unequal cable forces at the two ends of the saddle. Bending of the strands and friction between the strand and the saddle require special attention for fatigue and fretting.

Cables can be anchored to the opposite walls of a hollow concrete tower. Post-tensioning could be used to transfer the horizontal pulling forces from a pair of opposing cables (see Figure B6). Corrosion protection of the post-tensioning tendons requires special attention and the effectiveness of the short tendons should be accounted for in design.

Steel frames became popular in the early 1990s for connecting cables across a hollow concrete tower. The first bridge where this was used is the William Natcher Bridge in Kentucky. The horizontal components of a pair of opposing cable forces are taken up as tension in a steel frame. The unbalanced horizontal components and the vertical component of the cable forces are transferred from the steel frame to the concrete tower through shear studs. The unbalanced horizontal forces that occur during cable installation, cable replacement or cable loss are transferred to the tower wall through shear studs as well. The exposed surfaces of the steel frame are usually painted. These exposed surfaces can be directly accessed for inspection. The frame can be used as a working platform during construction, inspection and maintenance.

B 6.0 Construction and Erection Methods

Most long span cable-stayed bridges are constructed by the balanced cantilever method. Towers or pylons are generally constructed first. Then the superstructure or deck construction begins by having segments installed, one additional segment at a time with its supporting cable(s), to alternating sides of the tower. Steel box girder superstructures are usually preassembled into segments and lifted in place.

Occasionally, other construction methods such as incremental launching, traveling formwork, and falsework construction may be appropriate for the site. In these cases, construction of the 7

FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

superstructure may proceed in longer segments, each of which being supported by several cables. Even in these cases, the stay cables are typically stressed alternately from the two opposing sides of the pylon.

A steel composite superstructure is usually erected by having the steel frame preassembled, which includes steel edge girders and floorbeams, lifted up to the deck level and spliced to the previous segment (see Figure B7). Stay cables are then installed, followed by precast concrete deck panels. After the cast-in-place concrete joints are placed and cured between the precast panels, the concrete deck slab becomes composite with the steel edge girder and floorbeams.

A concrete superstructure could be precast or cast-in-place. Erection of precast segments and erection of steel box girder is similar. The cast-in-place method requires the use of form travelers for casting the deck segment by segment.10

As erection progresses, the cantilevered superstructure will continue to extend toward the anchor pier on the back span side. If the contractor elects to erect the anchor pier segment in advance, then a side span closure is required. After the closure is made at the anchor pier, erection of segments on the main span side will continue. When both halves of the bridge are complete, a mid span closure will be placed.

The length of the closure piece should be determined by making adjustments for temperature variation to the actual gap width. The gap width is always measured because it reflects the accumulation of tolerances of all previous segments and is therefore impossible to predict. Temperature variation throughout the day causes the gap width to vary constantly. However, the seasonal temperature variations cause the biggest adjustment required for closure. Since it is desirable to keep the deck permanently in compression, the length of the closure piece is usually cut according to the upper range of measured gap widths. If the gap is not large enough to fit the closure section, jacks are usually used to help increase the gap for a fit.

Before span closure, the partially completed bridge is very flexible and therefore quite vulnerable to wind. Wind tunnel testing is usually conducted to investigate the stability of the construction stages. Should the test indicate instability at a certain stage, temporary tie-downs can usually be installed to provide stability. At bridge sites where the water is deep and the bridge deck is high, wire rope or strands may be used to tie the deck down against the tower footing. Otherwise, struts that can take tension and compression can be installed to stabilize the deck in the back span. The type and configuration of the tie down should not only consider the aerodynamic requirements, but also the potential risk of obstructing the navigational channel and being damaged by debris carried by flood current.

B 7.0 Erection Stage Stress Analysis

The loading on a partially completed bridge, such as the weight and location of construction equipment and stored construction material, is generally very different from those carried by the completed bridge. Additionally, the effect of camber (rotational and length) on the partially complete structure may be additive to the construction loading and result in temporary stresses that could govern the design of certain elements.

In order to ensure the safe construction of a project, the responsibility to check the structure for all loading conditions is generally assigned to the contractor and his construction engineer. The

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

construction engineer may assist the contractor in selecting his construction equipment, developing a detailed construction sequence, as well as procedures for monitoring and controlling stresses in the structure during construction. An erection stage stress analysis should be of sufficient detail to account for the stress history of the structure, including consideration of creep and shrinkage effects, shear lag, construction loading, wind loading and any other loading that may cause a critical load case during the construction of the structure.

B 8.0 Construction Geometry Control

It is also common to assign the responsibility of geometry control to the contractor. The contract plans specify the target geometry of the bridge which the Contractor has to deliver at the end of construction (see Figure B8). As a part of construction stage analysis, the contractor’s construction engineer should develop a set of theoretical cable installation forces, unstressed cable length, a corresponding schedule of shim (or ring nut) adjustments, and progressive cambers that predict the displacement of the structure at a given stage of construction. Along with this set of data, a set of step-by-step instructions should be prepared for the contractor that indicate the expected locations of all major erection equipments on the structure that might affect cable forces and displacement of the structure. It is of great importance that a detailed record of construction activities, including equipment locations, cable installation records, and attained geometry, be kept for each stage of construction. When the actual cable forces and / or camber start to deviate from the predicted values, an explanation can often be found without resulting in delay to construction if detailed records are available for review.

Fixing a set of construction tolerances for cable installation forces, camber etc. is not always easy because deviations in virtually every construction stage should be tracked, and trends assessed and projected before the state of the structure can be evaluated. Experience shows that the contractor, his construction engineer and the owner’s representative should share as much of the database from the construction analysis as possible so that all parties have equal appreciation of the constraints and leeway inherent in the structure under construction. A set of project specific tolerances may be established initially for cable forces and camber based on reserve capacities in the design. However, this set of tolerances should be continually reviewed and updated to reflect the actual and anticipated stresses in the structure. Generally, more generous provisions for shim or ring nut adjustments can alleviate the need for holding to tight tolerances. Finally, open and continual communication between the construction engineer and the owner’s representative is always helpful in expediting the engineering process and avoiding unnecessary construction delays.

B 9.0 Uplift at Anchor Piers

A cable-stayed bridge generally has a longer, and hence heavier, main span than the back span. This arrangement creates an uplift force at the anchor piers. This uplift force is traditionally transferred to the anchor pier through tie-down devices. The tie-down devices mobilize the weight of the anchor piers to counteract the uplift force.

Several different types of tie-down devices have been used (see Figure B9): • Pin • Welded member with hinges at both ends

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• Cable

Tie-down devices such as those shown are fracture critical unless redundant elements are incorporated in the design. Design considerations should include replaceability of the tie-down, corrosion protection, as well as details to accommodate horizontal movement when the tie-down is located at an expansion joint. Fracture critical devices will require close inspection and maintenance, since their failure could cause a collapse of the bridge.

One alternative to a tie-down is to eliminate the uplift force altogether. Utilizing concrete counterweight made composite with the superstructure has been found to be an economical option for counterbalancing the uplift force. When a counterweight is designed to counterbalance the worst case of uplift comprised of the dead load uplift plus the maximum live load uplift, then the bridge will experience no uplift under any circumstances. The concrete counterweight is usually encased inside the deck, floorbeams, diaphragms and other enclosure plates. Only the underside of the counterweight is partially visible. The counterweight concrete is generally a maintenance free element that should be considered especially in cases where seismic effects are not dominant. The concrete counterweight method has been used in the William Natcher Bridge, for example. 11

For cable-stayed bridges that are flanked by approach spans, the weight of the approach spans can be activated against the uplift by making the cable-stayed deck continuous over the anchor pier. Negative moment over the anchor pier can be limited if necessary by introducing a rotational joint in the first flanking span.

Lightweight concrete may also be used in the main span to reduce the overall uplift force at the anchor piers.

B 10.0 Deck Replacement

A common strategy for designing durable cable-stayed bridge concrete decks is to minimize tensile stresses in the deck to limit or eliminate cracking, which is the principal source of deck corrosion. For designs involving edge girders, floor beams and two planes of cables, the deck slab is naturally under compression in the transverse direction due to the positive moment in the floor system associated with the two cable planes. In the longitudinal direction, there are small areas adjacent to the anchor piers and in the middle of the main span that may see tensile stresses. Some designers would elect to use localized post-tensioning to limit or eliminate longitudinal tension, while others would satisfy serviceablility requirements by checking stresses and providing adequate reinforcement to limit the potential of deck cracking. For designs with a single plane of cables, transverse and longitudinal post-tensioning are generally used to limit or eliminate deck cracking. Additionally, designers often specify coated reinforcing bars, an impervious wearing course, and inspection program to ensure longevity of the structure. This strategy appears to be satisfactory as there have not been reported problems with cable-stayed bridge decks that were designed, built, and maintained according to the above-mentioned strategy. In fact, there has not been any reported deck replacement even amongst cable-stayed bridges known to have been built and maintained by a lower set of standards.

Nevertheless, experience associated with conventional bridges has bridge owners concerned with the longevity of the cable-stayed bridge deck, so they often require that cable-stayed decks be designed for future replacement.

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It can be demonstrated that the replacement of a small portion of a cable-stayed deck is technically feasible. It would require decompression of the deck slab locally by transferring the compression to a temporary structure. Figure B10 illustrates this approach. The opening of the removed deck area shown must be enlarged by jacking to compensate for the elastic shortening and creep and shrinkage of the new concrete. After the concrete is cured to its design strength, the jacks would then be released and the new deck slab is supposed to take compression as the removed old concrete did. In this manner, the stress level in the girder web and bottom flange will not increase. However, it should be kept in mind that while this is theoretically feasible, in practice it is very difficult to regain the compression in the new deck as the compression field has moved away from the replacement area. It should be further noted that the stability of the bridge against transverse wind loads must also be provided by temporary structures which would transfer wind loads to the tower and anchor piers. The need to de-compress and re-compress the deck also applies to steel deck replacement.

While the above concept can be applied in stages and result in a global deck replacement, the high cost, including the requirement of maintaining traffic, would make this approach uneconomical if such an operation is considered in the life cycle cost analysis. For composite designs, the floor system can also be designed to carry axial loads without the concrete slab present. However, the deck slab of a typical cable-stayed bridge expends up to 80% of its capacity in resisting compression load from the cable stays. The addition of the compression element is nearly equivalent to providing two decks. It is therefore evident that this option will not make total deck replacement any more appealing. With due consideration of the above, designers often promote the strategy of exercising control of tensile stresses, investing in resources to provide a sound concrete design to start, and exercising a rigorous maintenance and corrosion protection program to avoid global deck replacement.

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Figure B 1 Cable Vibration

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Figure B 2 Cable Corrosion Protection

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Figure B 3 Cable to Girder Anchorage

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Figure B 4 Cable to Girder Anchorage

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

Figure B 5 Cable to Tower Connection

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

Figure B 6 Cable to Tower Connection

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

Figure B 7 Construction Erection Method

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

Figure B 8 Construction Geometry

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

Figure B 9 Tie Down

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

Figure B 10 Deck Replacement

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix B

REFERENCES

1 Tang, Man-Chung. “Cable-Stayed Bridges.” Bridge Engineering Handbook. Ed. Wai-Fah Chen and Lian Duan. 2000.

2 “Wind-Induced Cable Vibrations.” RWDI. 8 April 2004.

3 Bournand, Yves. “Development of New Stay Cable Dampers.” Proceedings of Cable-Stayed Bridges – Past, Present and Future, IABSE Conference, Malmö, Sweden. 1999.

4 Ferguson, Guy. “Ironton Russell Bridge: Crossties Control Vertical Cable Movement.” At the Moment – Motion Control News & Views from Motioneering. 2006.

5 Manzanarez, Rafael, Charles Seim, and Dennis Smith. “Beyond Engineering: The Politics of Maracaibo.” Civil Engineering. March 1991: 60-62.

6 Grant, Arvid. “Cables Not in Trouble.” Civil Engineering. May 1991: 61-63.

7 Cho, Aileen. “Unique Methods and Techniques Fuel A New Maine Attraction.” ENR. 10 July 2006: 26-32.

8 Chandra, Vijay and Ruchu Hsu. “Owensboro Bridge.” International Bridge Conference. 16 June 1998.

9 Gale, Lori and Julio Valvezan. Spotlight on the Owensboro Bridge: Innovative Cable Anchorages.” PB Network. Summer 1994.

10 Green, Peter. “Form Travelers Cast Bridge on Six-Day Cycle.” ENR. 7 January 1988: 32-34.

11 Hamawi, Fadi and Peter Wahl. “Spotlight on the Owensboro Bridge: Creative Solution at Anchor Piers.” PB Network. Summer 1994.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

TABLE OF CONTENTS

C 1.0 Arch Types ...... 1 C 2.0 Global Geometry and Load Path ...... 1 C 2.1 Truss Rib Load Path ...... 1 C 3.0 Stability of Arch Ribs ...... 2 C 4.0 Erection ...... 2 C 4.1 Length Adjustment for Rib, Tie and Hanger ...... 3 C 5.0 Redundancy of Tie ...... 3 C 6.0 Distortion Induced (Secondary Stress) Fatigue ...... 3 C 7.0 Hanger: Materials, Corrosion Protection and Fatigue ...... 4 C 7.1 Oscillation of Suspenders / Hangers ...... 4 C 8.0 Maintenance ...... 4 C 8.1 Corrosion Protection ...... 4 C 8.2 Accessibility to the Interior Spaces of Box Members ...... 5

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

LIST OF FIGURES

Figure C 1 Arch Bridge Types ...... 6 Figure C 2 Arch Bridge Types ...... 7 Figure C 3 Arch Bridge Types ...... 8 Figure C 4 Ideal Rib Shape ...... 9 Figure C 5 Arch Rib Buckling ...... 10 Figure C 6 Length Adjustments ...... 11 Figure C 7 Redundancy of Tie ...... 12 Figure C 8 Secondary Stress ...... 13 Figure C 9 Hanger Type ...... 14 Figure C 10 Hanger Stability in Wind ...... 15 Figure C 11 Corrosion Protection of Box Sections ...... 16 Figure C 12 Box Section Accessibility ...... 17 Figure C 13 Typical Bracing Systems for Arch Spans1 ...... 18

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

APPENDIX C – ARCH BRIDGES

Topics selected are those deemed likely to be of immediate application for the readers of these Guidelines. Materials presented are available in the literature and provided to supplement the Guidelines as a convenient source of reference. An overview1 is suggested for the readers who seek general and historic perspectives of this bridge type.

C 1.0 Arch Types (See Figures C1 to C3)

When there is sound rock for foundation, an arch bridge can be an economical and elegant solution. Where suitable ground is not available, tied arch, which needs only vertical support, can be an alternative.

A deck arch has the arch supporting the deck from below. A through arch has the arch support the deck from above. The selection of the deck or through arch depends on the underclearance available. A half-through (or semi-through) arch has portions of the arch below the deck near the supports and above the deck in the middle part of the main span.

Modern arch ribs are usually made of steel, concrete, or concrete filled steel tubes (CFST). Whether the arch ribs should be solid or hollow depends on the span length and the width of the bridge.

C 2.0 Global Geometry and Load Path

The most efficient design of the arch rib is keeping the compression through its neutral axis along the arch. A parabolic curve is the closest shape to satisfy this requirement. For long span tied arches, the shape of the arch must be derived from its dead load distribution. If the span length is not too large and the bending caused by the eccentricity is tolerable, a parabolic shape would be acceptable. A circular shape is used for some shorter span arch bridges (see Figure C4).

The rise-to-span ratio for arch bridges vary widely, as the arch can be very shallow or, at the other extreme, take the form of a half circle. Most arch bridges have a rise-to-span ratio in the range of 1:4.5 to 1:6.1

C 2.1 Truss Rib Load Path

A rib not only carries axial compression load but also resists bending moment due to unsymmetrical live load. A truss arch rib resists bending moments through its upper and lower chord members. Both the top and lower chords of a truss rib also resist high axial load. Because bearings usually support lower chords only, the axial force in the top chord must transfer to the lower chord through diagonal members and finally to the support bearings.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

C 3.0 Stability of Arch Ribs

The arch rib is a compression member that also carries bending. The rib is usually slender, which makes buckling a critical issue for the design of an arch bridge. Both global and local buckling should be checked (see Figure C5).

The critical buckling load of an arch rib can be computed by performing non-linear analysis by adding incrementally load until the deflection diverges. An eigenvalue solution can also be used to determine the critical buckling load.

To prevent buckling, arch ribs are generally externally braced, although some arch bridges have been designed without external bracing. The bracing types that are generally used include the K- type bracing shown in Figure C13.a, the diamond shape bracing shown in Figure C13.b, and the Vierendeel type bracing shown in Figure C13.c.

For a detailed discussion on the stability of arches, refer to the literature, including the following:

• Galambos, Stability Design Criteria for Metal Structures (Chapter 17 – Arches)2. The subject of stability of arches is very well handled in this reference. Values to use in formulas for critical buckling loads are listed in tables for many different cases of loading and various arch configurations. • Xanthakos, Theory and Design of Bridges (Chapter 10 – Elastic Arch Bridges)3

C 4.0 Erection

Most arch bridges span over water and gorges. Some arch bridges also span over land. The erection method of arches varies depending on site condition and constraints. The most common erection methods are falsework, cantilever and floating-in.

Falsework When the profile is low, falsework is the most economical method to erect an arch bridge. Falsework directly supports the form to construct the concrete ribs. A few struts can be used to temporarily support rib segments before closure and the arch can stand by itself.4

Cantilever Method In case the arch is built high above a deep gorge, an arch is commonly built by cantilever method. Hoover Dam Bridge used temporary stay cables and towers to support the cantilever arch until its closure.5 Another example of an arch constructed by cantilever method using temporary stay cables is the Natchez Trace Parkway Bridge in Tennessee.6

When the second Navajo Bridge over the Marble Canyon was constructed, the cantilevered truss arch was supported by horizontal tie-backs.7

Floating-in Method A tied arch could be completely assembled on a barge. Then the loaded barge is towed to the bridge site. For a low profile bridge, the bridge can be lowered and set onto its bearings by pumping water into the barge or following the receding tide.8 For a high profile bridge,

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

when high capacity floating cranes are available, the complete bridge could be lifted from the barge and directly set onto its bearings.9,10

C 4.1 Length Adjustment for Rib, Tie and Hangers

Constructing bridges that conform to geometric requirements laid out in the design plans is a basic requirement. The long span tied arch structure, being very flexible, poses special challenges. The tie, rib, and hangers will elongate or shorten (deform) significantly under the self weight of the bridge. All structural members must be fabricated with compensations made for these deformations – i.e., fabricate the arch ribs longer and the tie girders and hangers shorter (see Figure C6).

The amount of adjustment in the fabricated lengths of each element may be specified by the design on the contract drawings or by the construction engineer before the preparation of shop drawings.

C 5.0 Redundancy of Tie

Most tied arch bridges are supported by a pair of tied arches. In the case of a tied arch bridge, even the failure of one of two ties will cause a collapse of the bridge. In other words, the tied arch bridge lacks global redundancy. To increase the bridge safety, modern tied-arch bridges are designed to provide internal redundancy to the ties. The basic concept is to use multiple tension elements in the tie in lieu of a tie comprised of a single tension member (see Figure C7).

Three commonly used methods to provide internal redundancy to the tie are:

• Stitching plates by bolts at corners to form a box. The Blue Water Bridge11 and the Blennerhassett Bridge12 used this method. • Providing post-tensioning bars or strands to resist the tension and pre-compress the tie box that resists the bending and prevent cracking. The Lupu Bridge used this method.13 • Making the deck slab composite with the tie member in order to prevent abrupt failure. The Rancocas Creek Bridge used this method.8

Designing the details to make the tie to be Category B for fatigue and using high performance steel with greater toughness are used to increase reliability when increasing redundancy is not feasible.

C 6.0 Distortion Induced (Secondary Stress) Fatigue

Many existing tied arch bridges have fatigue cracks due to secondary stresses. This type of crack is seen frequently at the connection between the floorbeams and the tie girder. When the tie girder extends under live load, the floor beams want to move with the tie girder at their ends while being restrained by the deck slab (see Figure C8). Even with a design where the deck /stringers are supported by expansion bearings on the floorbeams, the accumulation of the friction from all the bearings will be high enough to engage the deck in preventing stress-free movement of the floorbeams. As the bearings get old and freeze up, the deck /stringer system behaves as if it was a composite deck / floorbeam design. The distance between the first bearing and the tie is always limited. The transverse deflection of the floorbeam induces bending moment

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

about its weak axis that causes high stresses. Since the live load is repetitive, these stresses cause fatigue type cracks. The cracks in the floorbeam may propagate to the tie girder when a welded connection is used. The FHWA Technical Advisory T5140.4, September 28, 1978 addresses this problem.14 One way to eliminate this secondary stress is to make the deck composite with the tie girder. In this approach, the deck should be connected to the tie girder after all dead load is in place. This will preclude the majority of the tie girder elongation from affecting the floor beam connections. The composite action between the deck and the tie girder will prevent relative movement between the two elements thus protecting the floorbeam end from any distortion and eliminating the development of fatigue cracking.

C 7.0 Hanger: Materials, Corrosion Protection and Fatigue

Wire ropes, solid rods, rolled shapes, built-up members, 7-wire strands, and parallel wires have all been used for hangers (see Figure C9).

Galvanization is the basic corrosion protection of the hangers. Painting may replace the galvanization or can be applied on top of the galvanization. Hangers made of 7-wire strands have been used while enclosed inside a PE pipe for corrosion protection and aerodynamic considerations.

Corrosion at the lower socket is a common problem due to accumulation of debris and moisture. Protection of the lower socket should include self cleaning features, extra layers of protection and considerations for regular cleaning to prevent the collection of debris and moisture.

The hangers support the floorbeams, which directly carry truck loading near their ends. The tensile stress range is large and the number of cycles is usually high. The hangers should therefore be designed for fatigue load.

C 7.1 Oscillation of Suspenders / Hangers

A commonly observed phenomenon is the oscillation of suspenders associated with wind effects. This is caused mainly by vortex shedding. This type of oscillation should be suppressed, because the vortex shedding induced oscillation can result in fatigue failure. Reducing the free length of the suspenders by installing spacers is the most commonly used solution (see Figure C10).

C 8.0 Maintenance

C 8.1 Corrosion Protection

All steel surfaces must be protected from corrosion. Paint is the most commonly used method. Box members may be sealed and filled with dry air. Provisions for checking for leaks and refilling with dry air should be provided (see Figure C11).

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

C 8.2 Accessibility to the Interior Spaces of Box Members

The interior surfaces of box-shaped members must be accessible for inspection and painting or other means of corrosion protection. Large interior spaces should be provided with adequately sized openings for personnel access and delivery of maintenance equipment. Ventilation should also be provided to supply oxygen as well as to vent fumes associated with painting or welding (see Figure C12).

For boxes that are too small for personnel access, a series of hand-holes should be provided to allow inspection and painting. These hand-holes should face down to prevent retention of rain water or the accumulation of moisture. Hand-holes should be covered by screens to keep birds or other animals out.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 1 Arch Bridge Types

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 2 Arch Bridge Types

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 3 Arch Bridge Types

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 4 Ideal Rib Shape

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 5 Arch Rib Buckling

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Figure C 6 Length Adjustments

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 7 Redundancy of Tie

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 8 Secondary Stress

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Figure C 9 Hanger Type

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 10 Hanger Stability in Wind

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Figure C 11 Corrosion Protection of Box Sections

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C 12 Box Section Accessibility

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

Figure C13.a K-Type of Bracing

Figure C13.b Diamond Type of Bracing

Figure C13.c Vierendeel Type of Bracing

1 Figure C 13 Typical Bracing Systems for Arch Spans (Plan View)

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix C

REFERENCES

1 Fox, Gerard F. “Arch Bridges.” Bridge Engineering Handbook. Ed. Wai-Fah Chen and Lian Duan. CRC Press, Boca Raton, 2000.

2 Galambos, T. V., Stability Design Criteria for Metal Structures, 5th ed., John Wiley & Sons, New York, 1998.

3 Xanthakos, Petros P., Theory and Design of Bridges, John Wiley & Sons, New York, 1994.

4 Hoeckman, Wim. “Bridge over the River Loire in Orleans, France.” Structural Engineering International. February 2001: 94-98.

5 Goodyear, David and Robert Turton. “The New Colorado River Bridge.” Structure Magazine. January 2006: 29-32.

6 Green, Peter. “Technology Advances Arch Construction.” ENR. 8 March 1993: 28-32.

7 “Arch Rivals at Grand Canyon.” ENR. 13 February 1995.

8 Hsu, Ruchu and Lori Gale. “Rancocas Creek Bridge – A State of the Art Design of a Railroad Tied-Arch Bridge.” Proceedings from the International Bridge Conference. 2004.

9 “Fremont Bridge: Portland, Oregon.” Oregon State Highway Department. Completed 1974.

10 Normile, Dennis. “Double-Deck Arch Lifted Atop Piers.” ENR. 22 February 1993: 26-27.

11 “Second Blue Water Bridge: Port Huron, Michigan and Point Edward, Ontario.” Michigan Department of Transportation. Completed 1997.

12 Deneault, Joe. “Bridging the Past.” Modern Steel Construction. January 2007: 33-36.

13 Yuanpei, Lin and Zenhang Zhang. “Lupu Arch Bridge, Shanghai.” Structural Engineering International. January 2004: 24-26.

14 “Tied Arch Bridges.” U.S. Department of Transportation, Federal Highway Administration, Office of Engineering. Technical Advisory T 5140.4. 28 September 1978.

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FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix D

Appendix D – General Design Process for Long Span & Complex Bridges

Arch and cable-supported bridges are often complex and long span structures located at physically challenging sites. Site specific wind, geotechnical, and hydraulic data are some of the essential information required for designing these structures. Since collection and interpretation of these data would take longer for complex and long span bridges than for conventional bridges, additional time should be allotted for collection and studies at the beginning of the project. This is especially true when a limited amount of data is available from past projects or for other purposes in the vicinity of the bridge site. Data may include-

Wind Data: • Wind Climate Study • Site Analysis • Studies and Analyses

o Historical data o On-site wind speed measurements o Characteristics of local terrain

Geotechnical Data: • Subsurface investigation (boring) • Rock shear wave velocity measurements

Hydraulic Data: • Historical flood and tide data • Flood and tide current measurements • River and sea bed material data • Historical scour data • Scour measurements • Historical wave data • Wave measurements

Other than the long lead investigations, developing the design criteria is very important to the success of the project. Depending on project specific requirements, the design criteria may include: • Design Specifications

o Owner or State modifications to national design and construction specifications o Pedestrian, bike, and transit requirements (current or future provisions) • Bridge underclearances, vertical and horizontal • Bridge alignment and profile • Roadway requirements • Air traffic (current or future provisions) • Bridge type: suspension, cable-stayed, arch, etc.

1 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix D

• Other constraints

Based on the design criteria and the preliminary geotechnical, wind and hydraulic data, preliminary design may start. By the completion of the preliminary design, all major members should be sized. All major boundary conditions should be determined as well. These major boundary conditions may include: • Fixed and expansion locations • Bearing or no bearing • Hinge locations

Wind load is the governing design consideration for most cable-supported bridges. An aerodynamic stability check needs to be performed as early as possible. Aerodynamics is a highly specialized discipline, so in most cases, stability assessments and checks should be performed with the input of a consultant who specializes in this area. Sectional model wind tunnel tests may be performed during the preliminary design stage, and the aeroelastic wind tunnel tests may be performed early in the final design stage. The structure needs to be modified when the wind tunnel test result is not satisfactory. The time required for wind tunnel tests depends on the complexity of the bridge. For a simple case, section tests take 2 months, while aeroelastic model tests could take 4 months or more, depending on the complexity of the structure and the site conditions, as well as availability of testing facilities.

Initial boring of a few holes in the general area of the proposed bridge would provide a general understanding of the subsurface condition. Boring holes at bridge foundations may be required to provide data for preliminary design. When the foundation locations are finalized, boring holes at the exact foundation locations should be drilled for foundation detail design. Each boring program would take a few months to complete. It will take a longer time to work on water, especially on the open sea.

Following the preliminary design, final design can start – usually with the benefit of additional geotechnical and interdisciplinary input to completely define the design requirements. Final design includes analysis and checking of the structural members, followed by detailing (Drawings) and developing specifications and cost estimates.

The accompanying Figure 1 shows a flowchart that summarizes the main activities and their relationships in the design of a typical long span bridge. Figure 2 shows a flow diagram for aerodynamic studies.

2 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix D

3 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix D

4 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

TABLE OF CONTENTS

E 1.0 Introduction ...... 1 E 2.0 The Development of a Structural Health Monitoring Program ...... 5 E 2.1 Overall Automated Bridge Health Monitoring Systems for Cable-Supported Bridges ...... 5 E 3.0 Descriptions of Automated Detection Methods...... 11 E 3.1 Inert Gas Dehumidification ...... 11 E 3.2 Vibration-Based Structural Health Monitoring ...... 12 E 3.2.1 Accelerometers ...... 13 E 3.2.2 Laser Vibrometers ...... 17 E 3.3 Weigh-In-Motion Methods ...... 17 E 3.4 Strain Gauges ...... 19 E 3.5 Global Positioning System (GPS) ...... 20 E 3.6 Acoustic Monitoring ...... 20 E 3.7 Peak Displacement Sensor ...... 21 E 3.8 Elasto-Magnetic Stress Sensor ...... 22 E 3.9 Fiber Optic Sensors ...... 22 E 3.10 Weather Sensors: Temperature Sensors, Anemometers, Barometers, Rainfall Gauges, and Hygrometers ...... 26 E 3.10.1 Temperature Sensors ...... 26 E 3.10.2 Anemometers ...... 27 E 3.10.3 Barometers ...... 27 E 3.10.4 Pluviometers (Rainfall Gauges) ...... 28 E 3.10.5 Hygrometers ...... 28 E 3.11 Digital Video Cameras ...... 29 E 3.12 Corrosion Cells ...... 29 E 3.13 Tiltmeters ...... 29 E 3.14 Infrared Thermography (IR) ...... 30 E 3.15 Bearing Sensors ...... 30 E 3.16 Scour Detection ...... 31 E 4.0 Information Transfer - Fiber Optic Networks/Wireless Communication ...... 31

i FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

LIST OF FIGURES

Figure E 1 Typical System for Automatic Sensor Data ...... 4 Figure E 2 Layout of Sensors on the Akashi Kaikyo Bridge (Japan) ...... 7 Figure E 3 Layout of Sensors on the Tsing Ma Bridge (Hong Kong)...... 8 Figure E 4 Layout of Sensors and Data Acquisition Units on the Sutong Bridge (China) ...... 9 Figure E 5 Sensor Configuration for Tatara Bridge (Japan) ...... 9 Figure E 6 Algorithm for a Vibration-Based Structural Health Monitoring System ...... 13 Figure E 7 Plan and Elevation View of Vincent Thomas Bridge, Showing Sensors ...... 14 Figure E 8 Isometric View of Vincent Thomas Bridge, Showing Sensor Locations ...... 15 Figure E 9 Application of Peak Displacement Sensor to Monitor Peak Displacement of Bridge Pedestal Damper ...... 21 Figure E 10 Application of Elasto-Magnetic Stress Sensor for Monitoring True Stress in Bridge Cables ...... 22 Figure E 11 View of the Turin, Italy Lingotto Pedestrian Cable-Stayed Bridge...... 23 Figure E 12 Fiber Optic Sensor on a Cable of Lingotto Bridge ...... 23 Figure E 13 Schematic of the FBG Cable Sensor on Lingotto Bridge ...... 24 Figure E 14 Layout of Fiber Optic Sensors on the Jiangyin Bridge (China) ...... 24 Figure E 15 Fiber Optic Sensors Embedded Within a Bridge Cable Cross Section ...... 25 Figure E 16 A Fiber Optic Crack Sensor ...... 25 Figure E 17 A Crack Sensor Used to Monitor a Bridge Pier ...... 26

ii FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

LIST OF TABLES

Table E 1 Suspension Bridge Typical Structural Health Monitoring System ...... 5 Table E 2 Cable-Stayed Bridge Typical Structural Health Monitoring System ...... 6 Table E 3 Damage Detection Categories and Computational Methods Available ...... 10 Table E 4 Sensor Locations for Vincent Thomas Bridge ...... 16

iii FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

APPENDIX E – BRIDGE HEALTH MONITORING

E 1.0 Introduction

Health Monitoring is gaining recognition among owners of major bridges. Major bridges may include long-span bridges (such as suspension, cable-stayed and arches) or structures comprised of short and medium spans that share standard principal features. This appendix addresses health monitoring of long-span bridges with emphasis given to new designs, although selected references are also made to existing bridges that have been retrofitted with monitoring systems when deemed relevant to new structures. This appendix also provides an overview of emerging bridge health monitoring technologies that are under development for bridge engineers.

Some key issues for long-span bridges that warrant the consideration of health monitoring systems include:

• Towers, arch ribs and cable elements of long-span bridges make the structure susceptible to the uncertainties of wind and its aerodynamic effects. • Complex or proprietary devices used on long-span bridges often require monitoring for optimal performance. They are often used to mitigate between competing demands on flexible long-span structures, i.e. the need to make the structural system rigid for dynamic effects such as wind, and flexible for others, such as seismic effects and differential settlements. High performance devices, such as lock-up devices, isolation bearings, tuned mass dampers, etc., are often needed to mitigate extraneous construction costs. • Current design trends include consideration of the high cost of total bridge replacement, and the trend to extend design life through the use of durable high performance materials in conjunction with replaceable/sacrificial elements. • Currently, the cost of inspection and maintenance of long-span bridges is relatively high. • Consideration should be given to the nonlinearity and variability of parameters associated with the design of long-span bridges.

Monitoring a long-span bridge is primarily used preventively to assess structural behavior that may change gradually over time, as well as to detect the structure’s response to sudden and extreme events. As non-destructive sensing is becoming more commonplace, particularly in new long- span bridge design, automated analysis of collected data is being considered and developed for future health monitoring practice. Automatic long-term monitoring may seem expensive on an annualized basis at the present, but holds promise to become an effective means to avoid or prevent problems before they grow too large and expensive to understand, analyze, or handle in the budget planning process. The general objectives of long-span bridge structural health monitoring may, therefore, be any or all of the following: . Design Verification:

o To provide data on structural dynamic response to verify design assumptions and/or performance of the structure during wind, seismic, vessel impact, storm surges, and other special or extreme events.

1 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

o To provide data to monitor longer-term non-extreme events such as gradient of temperature changes, creep and shrinkage of concrete, soil-structure interaction, scour, and water currents.

o To collect data for improving existing design guidelines or developing future design guidelines.

o To develop a reliable health monitoring system that checks itself and identifies anomalies in the system itself. . Structural Maintenance:

o To provide data for analyzing and evaluating the healthy behavior of the difficult-to- replace components (due to component cost, access, or lane outage availability) or non-replaceable bridge components. This may be desirable regardless of the age of components.

o To provide data to aid in assessing structural deterioration and performance degradation of replaceable structural elements such as bearings, dampers, and lock- up devices. . Traffic Management:

o To provide data that would help make assessments on whether traffic restriction is necessary after seismic or extreme wind events.

o To provide data for assessing post-earthquake or post-hurricane structural reliability and the ability of the structure to accept live traffic loads or to divert traffic automatically while the assessment is being carried out. . Inspection

o To reduce the cost of inspection, since all bridges are currently inspected with the same minimum frequency, regardless of condition.

o To provide real-time data to detect potential problems between intermittent inspections. Although it is possible to retrofit a bridge with a structural health monitoring system, a comprehensive system is probably most cost-effective if considered during the design phase for a new bridge and installed during its original construction. One methodology for designing a structural health monitoring system involves a modular concept. Six integrated modules can be described in terms of their respective functional performance requirements: (1) the sensory system; (2) the data acquisition and transmission system; (3) the data processing and control system; (4) the structural health evaluation system; (5) the structural health data management system; and (6) the inspection and maintenance system. A structural health monitoring system for long-span bridges should be able to monitor the loading and structural parameters set by the bridge Designer so that the bridge performance under current and future loading conditions can be evaluated20, 26. Monitoring typically involves tracking two parameters: load effects and structural response. Loads may be due to wind, seismic event, temperature (including the effects of fire), impact from water/ice/debris flow, and vehicular or rail traffic. Structural responses may be forces, displacements, rotations, accelerations, stresses, depths (of scour) and strains. Loads are typically measured by automated sensors that may include (earthquake forces); anemometers (wind speed and direction); triaxial accelerometers (dynamic structural behavior in three orthogonal directions); velocity gauges (vibration response due to wind and earthquake loads); GPS (global location and position); thermometers (temperature of structural steel, concrete,

2 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

steel cables); and displacement gauges (displacement of girders, joints, or other structural members). Much work and research to-date on structural health monitoring systems resembles existing bridge management systems, which focus on processing collected data. The robustness of a structural health monitoring system is often defined in terms of levels15 Level I: Determine whether damage has occurred Level II: Determine that damage has occurred and determine the location of the damage Level III: Determine that damage has occurred, locate the damage, and estimate its severity Level IV: Determine that damage has occurred, locate the damage, estimate its severity, and evaluate the impact of the damage on the structure or estimate the remaining useful life of the structure Monitoring can be performed automatically using sensors planned for during design and installed during construction of newer bridges, or retrofitted on existing bridges built before advances in technology allowed for fiber optic or wireless transfer of data from automated sensing devices. Alternatively, data can be gathered manually using technicians with expertise in the particular monitoring technique of interest. Often, a combination of both scenarios is optimal. A typical system for automatic sensor data collection on a simple span bridge is given below for illustration purposes25. Systems for long-span bridges are understandably more complex. While automated monitoring gives streaming real-time data under all operating conditions and extreme events, it is a tool for aiding in pinpointing potential trouble spots to be targeted for further in-depth investigation (usually visual inspection). Manual monitoring can give an in-depth instantaneous data point or “snapshot in time”. Comparison with previous history of bridge behavior must therefore rely on the previous gathering and availability of additional data in the past. A progressive alternative to automated monitoring is “smart automated monitoring”, where digitized streaming of real-time data is filtered for outlying data points, or frequency of data stored is altered based on polling the relative magnitude of recent data points that are collected on a regular basis. Weigh-in-Motion technology, for example, utilizes this feature in filtering many wheel load data points to save and record extreme wheel load events for use in stress frequency and fatigue computations. As another example, temperature sensors can sense and save temperature highs and lows on a daily thermal cycle, storing or replacing data points with the next outlying point as new data are received. Similarly, sensors can be designed and programmed to take more or less frequent automated readings, based on an automatic assessment of past readings. For example, due to the development of a hurricane or seismic event, an accelerometer may detect unusual movements, and the frequency of readings may be programmed to increase automatically due to the unusual nature of previous readings. Localized temperature and GPS sensors at suspension bridge midspan locations (where the main cables are close to roadway level and vulnerable to impact or fire damage) can be programmed to automatically increase the frequency of data collection on impact or increase in temperature due to fire. The increase in frequency of any bridge sensor reading could be gradual or very rapid, depending on a comparison of the most recent data polling and the magnitude of change. For example, GPS sensor data from protective dolphins or sacrificial armoring of piers or cable low points can detect movements due to ship or vehicle impact, and the frequency of data can be programmed to increase rapidly on impact. This principle is used for more rapid data gathering due to earthquake forces in the structural health monitoring system used on the Rion-Antiron Bridge (Greece)14. Hydraulic cylinders, in lock-up devices for example, that allow gradual relative movement between the superstructure and substructure due to thermal changes (at expansion joints at towers or anchoring points on some cable-supported bridges) but contain orifice plates to lock relative

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movements due to seismic events, can have their output pressure or fluid flow used as input for determining the frequency of data-gathering by accelerometers. Similar hydraulic information can be used from tuned mass dampers that are installed in the tower tops of some cable-supported bridges. These are examples of smart automated monitoring techniques for long-span bridges

TYPICAL SYSTEM FOR AUTOMATIC SENSOR DATA

COLLECTION Figure E 1 Typical System for Automatic Sensor Data where remote sensors can be programmed to revise the frequency of data collection due to outlying data from an event at another remote location. Programming may include self-checks to correct for outlying data, or trigger alarms for examination of the situation by the bridge owner and manual resetting if necessary (see Figure E125). Many devices such as tuned mass dampers that use hydraulic fluid or water or other mechanical parts may include automatic sensing devices as part of their operation. Automatic sensing devices

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that use fluids may include internal sensors for fluid levels or flow, or fluid pressure or leakage. If these types of devices are not included in the individual device it may be a consideration for inclusion by the design engineer in order to ensure that the device operates properly, ensure that the device captures crucial data in extreme events, and maintenance is handled on a timely basis. In designing a structural health monitoring system for new and existing long-span bridge structures, a hazard and operability study can be used as a useful tool to ask “what if” questions. By tracing a “what if” question, step-by-step, to all of its possible outcomes and their probabilities of occurrence, the design of a new bridge (or retrofit design for an existing bridge) may be revised for the desired outcome. Alternatively, the bridge health monitoring systems may be adjusted and/or installed to mitigate anticipated risks where other cost-effective design alternatives are not available. E 2.0 The Development of a Structural Health Monitoring Program

E 2.1 Overall Automated Bridge Health Monitoring Systems for Cable-Supported Bridges

An example of a typical monitoring system and its components for a new suspension bridge follows in Table E11: Suspension Bridge Typical Structural Health Monitoring System

Table E 1 Suspension Bridge Typical Structural Health Monitoring System

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An example of a typical monitoring system and its components for a cable-stayed bridge follows in Table E24:

Cable-Stayed Bridge Typical Structural Health Monitoring System

Table E 2 Cable-Stayed Bridge Typical Structural Health Monitoring System

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The number and placement location of sensors is important to the overall effectiveness of an integrated bridge health monitoring system, as are the frequency of data collection and filtering of data. A carefully planned system should be developed to achieve its goal without the collection and processing of too much information. The frequency of occurrence and importance of the parameters are weighted to determine the comprehensive bridge monitoring system to be implemented4. Some examples of the placement of sensors on some newer cable-supported long- span bridges are given in Figures E21, E318, E49 and E51 below1, 9, 18.

LAYOUT OF SENSORS ON THE AKASHI KAIKYO BRIDGE (JAPAN)

Figure E 2 Layout of Sensors on the Akashi Kaikyo Bridge (Japan)

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A short summary of the sensor placement of the Akashi-Kaikyo Bridge is given below: The sensor layout system on the Akashi-Kaikyo Bridge, depicted in Figure E2, was installed primarily for design verification. Seismometers were installed in the anchorages. The was installed on one anchorage at a distance of approximately 100 meters from the bridge axis, in order to avoid response vibration influence of the foundation to the original seismic motion. The anemometer was placed at midspan in order to measure wind speed in the longitudinal and transverse directions. A three-component accelerometer was installed in at least one location on each foundation to verify the structural behavior due to seismic forces. Velocity gauges were installed on the girders and towers to assess their behavior due to wind and seismic forces. A GPS unit was installed at one anchorage as a fixed working point. Additional GPS units were installed at the tower tops and at midspan to measure displacements in three orthogonal directions. Girder edge displacement gauges were installed on the west and east edges of the center span, and on the west side of one side span. Tuned mass dampers were installed in each tower, a need confirmed by wind tunnel tests that showed that vortex oscillation would occur following a wind speed below the design wind speed. In order to compensate for temperature for the GPS displacement measurements, three cable thermometers were installed, and one atmospheric thermometer was installed at bridge midspan. Monitoring sensors were divided into blocks based on their location of installation, and terminals were established for each block. From each block, data are transferred via optical fiber to the information unit room located near one anchorage, where the data are collected and processed. Monitoring data are displayed in the information unit room.

LAYOUT OF SENSORS ON THE TSING MA BRIDGE (HONG KONG)

Figure E 3 Layout of Sensors on the Tsing Ma Bridge (Hong Kong)

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LAYOUT OF SENSORS AND DATA ACQUISITION UNITS ON THE SUTONG BRIDGE (CHINA) Figure E 4 Layout of Sensors and Data FIGUREAcquisition E4 Units on the Sutong Bridge (China)

SENSOR CONFIGURATION FOR TATARA BRIDGE (JAPAN) Figure E 5 Sensor Configuration for Tatara Bridge (Japan)

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The processing of sensor data to make structural maintenance and policy decisions is in the formative stages. Sensors placed on existing bridges largely have been used to verify finite element models developed during design, to monitor physical parameters over the medium to long term, or to develop baseline data of modal behavior for comparison with future data on modal behavior. Research on processing sensor data using computational methods is currently being performed. Several references discuss computational methods in this emerging field15, 16, 18, 20, 21, 22, 23. Table E3 summarizes some categories of damage detection and computational methods available15. Fully automated use of sensor data is still an emerging technology.

Damage Detection Categories and Computational Methods Available

Table E 3 Damage Detection Categories and Computational Methods Available

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E 3.0 Descriptions of Automated Detection Methods

E 3.1 Inert Gas Dehumidification

An inert gas can be used to either circulate or statically maintain a slight positive pressure in steel members to provide a constant, known environment. Research has shown that maintaining carbon steel in an environment with relative humidity below 50% significantly reduces its corrosion rate. Dehumidified air is often used in the cross section of wrapped and sealed main cables of suspension bridges (as well as in the anchorage chambers) in order to maintain humidity below 50% relative humidity. Dehumidified air is also used in the chambers of steel box members to provide an environment with a relative humidity generally below a set point 40%, in order to ensure that the actual relative humidity remains below 50%. The relative humidity of the dry air (being injected into the cable cross section or being sent to the anchorage strand enclosure) as well as the relative humidity of the exiting air (from the cable cross section or anchorage strand enclosure) are both monitored as an indirect means of assessing the water leakage into the cable or anchorage, condensation in the cable or anchorage due to thermal gradients, and the overall effectiveness of dehumidification efforts. Bridge health monitoring in the area of dehumidification can also involve monitoring the electrical usage of a dehumidifier over time. Unusually high electrical usage may indicate a breach in the cable wrapping or malfunctioning dehumidification equipment, or an unusually average low electrical usage may indicate faulty sensors or malfunctioning dehumidification equipment. Triggers or alarms can be set to alert maintenance personnel of unusual electrical usages from a dehumidification plant. Pressure drop through the system can also be monitored to indicate additional blockage (possibly from wire corrosion) or abnormally low pressure drop, possibly indicating system leaks or breaches. Triggers or alarms can be set to alert maintenance personnel of unusually high or low pressure drops in the system. One of the issues for consideration in installing main cable dehumidification on existing, older suspension bridges is the fact that oiling of the cables has been a preventive maintenance practice in the past, and may hinder the flow of air through the cable cross section. Also, the cable may need to be rewrapped to form a tight seal for air flow through the cable cross section (an added capital expense for project planning). While main cable dehumidification as a preventive maintenance measure has not yet been adopted in the United States for suspension bridges, it is currently being considered for the William Preston Lane, Jr. Memorial Bridges in Maryland and the Walt Whitman Bridge in Philadelphia, Pennsylvania. Nitrogen gas has been used in the cross section of the stay cables on the Penobscot River cable- stayed bridge in Maine to maintain a slight static over-pressure inside the stay-cable cross-section, with a relative humidity below 50%. This bridge has continuous cable stays through the tower that are cradled in specially-designed cable saddles. Nitrogen from compressed gas bottles is used to maintain the static over-pressure in the stay cables. Monitoring of nitrogen depletion and monitoring of the maintenance of a constant over-pressure are both used to assess the success of the dehumidified environment of the stay-cable strands. A loss of pressure indicates a breach in the initial corrosion barrier or a malfunction in the automated pressure maintenance valve. The system eliminates the need to protect the end of the stay anchorage with grease, and no cable stay grouting is used. The end of each anchor has a full diameter clear polycarbonate portal window for easy visual inspection, and a low-point drain to allow water to be removed5 . Current on-going research at Columbia University and City University of New York (both in New York City) involves exposing bridge main cable cross sections to various weather conditions with the intent to test various sensors for remotely monitoring main cable health. This is currently a topic of great interest in the United States, since many major suspension bridges in the U.S. are over 40 years old. Other researchers have studied the mechanism of initiation and propagation of

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pitting corrosion of galvanized steel, a particularly important topic of interest in steel bridge wires, since it can lead to brittle wire failures and affect main cable reliability. Under some conditions, corrosion products have been observed to be highly localized and can form a barrier. This can lead to accentuation of the difference in pH within and outside the pit, thus exacerbating and accelerating pitting corrosion. Thus, other researchers suggest that pitting is a highly localized condition that can be caused by large pH differences on a micro-scale rather than on a moderate- or macro-scale6. E 3.2 Vibration-Based Structural Health Monitoring

The main principle of vibration-based structural health monitoring relies on changes in structural characteristics such as mass, stiffness, and damping, which affect the global vibration response of the structure. By studying changes in measured vibration behavior, unknown changes in structural properties can be identified. The global nature of vibration characteristics provides advantages compared to other health monitoring techniques. Global vibration signatures such as natural frequencies and mode shapes lead to monitoring of the entire structural system, not individual structural components. Thus, a long-span bridge can be monitored effectively with a relatively small set of sensors and minimal equipment. The basic procedure of a vibration-based structural health monitoring system is summarized in five steps, as shown in Figure E616: 1. Measurement of structural dynamic response, e.g., in terms of accelerations or displacements; 2. Characterization of an initial bridge model through dynamic and static tests; 3. Continuous monitoring and damage localization of the structure; 4. Performing finite element model updating in order to obtain up-to-date structural model; 5. Evaluation of structural performance using the updated finite element model.

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ALGORITHM FOR A VIBRATION-BASED STRUCTURAL HEALTH MONITORING SYSTEM

Figure E 6 Algorithm for a Vibration-Based Structural Health Monitoring System

Vibration-based structural health monitoring has been performed on a short-term basis on the 18- span Watson Wash Bridge (California) and the Vincent Thomas suspension bridge (California) with good results16. For continuous monitoring of a long-span bridge, large quantities of dynamic data are collected and need to be filtered and processed. The instrumentation system must be designed to handle this filtering and throughput effectively and efficiently. Using data to provide an initial characterization of the structure serves as a baseline model that adequately predicts the structural behavior in its initial, or pristine, state. Changes in features and structural vibration properties can reveal information about structural health. Techniques have been developed to statistically discriminate and distinguish between noise and structural damage. E 3.2.1 Accelerometers Accelerometers can be used on long-span bridges to measure traffic-induced vibrations and modal frequencies of roadway decks as well as movements and structural monitoring due to seismic, tidal, temperature, or wind effects. Triaxial accelerometers are used to obtain vertical, horizontal (transverse and longitudinal) orthogonal accelerations. Accelerations can be obtained from deck, cables, and bridge towers on cable-supported bridges, and can be used in conjunction with GPS data to improve the frequency response of deflection measurements. They are useful in monitoring long-term deformations as well as instantaneous deflections. Accelerometers can achieve sampling rates of several hundred Hz or higher, which can be important for some bridge dynamics depending on the application needs. Triaxial sensors are also superior to other sensors in that they do not depend on the propagation of electromagnetic waves, and therefore avoid issues of signal refraction and line-of-sight problems with terrestrial or space objects, and are unhindered by adverse weather conditions. Drift caused by instrument bias and scale factor offsets make accelerometers inaccurate by

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themselves over long periods of time; accelerometers are useful for measuring high frequency vibrations and small movements, but not low frequency vibrations and large deformations. Hence it is useful to combine accelerometers with GPS positioning to form an optimal combination of position measurement. Current GPS technology and the availability of satellites in the proper positions are impediments to using GPS measurements alone for precise engineering work7. Accelerometers on cable-supported bridges are typically placed on the stay cables, towers, and deck to address dynamic effects. Accelerometers have been used on numerous bridges, including the Verrazano-Narrows Bridge in New York City, where the behavior of the bridge was monitored during the running of the New York City Marathon. The Vincent Thomas suspension bridge in Los Angeles, CA (USA) has been fitted with 26 accelerometers since 1980 as part of a seismic upgrade project (see Figures E716 and E816 and Table E4). The accelerometers installed on the bridge measure accelerations in three orthogonal directions: vertical, transverse horizontal and longitudinal horizontal directions16. The agreement of modal parameters between ambient vibration data and finite element-based modal parameters is good. Comparison to the 1987 Whittier earthquake and the 1994 Northridge earthquake are also good. Ambient vibration data have also been used successfully on the Alfred Zampa Memorial Bridge (New Carquinez Bridge) (USA)12.

PLAN AND ELEVATION VIEW OF VINCENT THOMAS BRIDGE, SHOWING SENSORS

Figure E 7 Plan and Elevation View of Vincent Thomas Bridge, Showing Sensors

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ISOMETRIC VIEW OF VINCENT THOMAS BRIDGE, SHOWING SENSOR LOCATIONS

Figure E 8 Isometric View of Vincent Thomas Bridge, Showing Sensor Locations

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Sensor Locations for Vincent Thomas Bridge

Table E 4 Sensor Locations for Vincent Thomas Bridge

For many long-span bridges the dominant structural modes due to ambient excitation are in the lower end of the frequency spectrum, usually less than 1 Hz. Traditional piezoelectric accelerometers have limited accuracy operating in such low frequency ranges. Capacitive or resistive accelerometers are more suitable for such applications. Modal vibration parameters can be determined by analysis, and from this, the type, number, and location of sensors in the sensing system can be determined, and the data acquisition, sampling rate, transmission, and archiving system can be properly designed. This can be used to update a finite element model of the bridge structure. The updating of the finite element model should occur on a regular frequency in order to detect global damage, be it gradual (e.g. corrosion over time) or instantaneous due to extreme events (e.g. earthquake, wind storm, or water/ice/snow-load damage).

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E 3.2.2 Laser Vibrometers Remote local techniques using laser readout of vibration are currently performed manually on bridge suspenders and stay cables on cable-stay bridges. Such methods are currently considered localized vibration NDE methods and have not yet been useful for overall structural health monitoring. The chief advantage of this method is its low cost and ease of use. A laser vibrometer can be used from a remote location relative to the measuring location and does not involve the expense of installing and maintaining accelerometers at remote locations on a bridge that are frequently difficult to access. This method has been used successfully to measure the vibrations and compute the corresponding tension forces in all 84 stay cables of the Sunshine Skyway Bridge. Calculation of cable forces based on vibration frequency for that bridge used a computation model that accounted for cable sag, bending stiffness, fixed boundary conditions, and intermediate springs and dampers. This method yielded more accurate results versus the over-simplified “taut string model” equation2. Successful NDE monitoring of new construction on the suspenders for the Cass Street Arch Bridge in La Crosse, WI has been carried out in 2004 using an impact mallet to artificially induce vibration, and adjusting suspender tension to balance loads7. The laser vibrometer method, while not currently implemented automatically in the US, holds promise as an easy method for measuring vibrations at remote locations.

E 3.3 Weigh-In-Motion Methods

Weigh-in-motion (WIM) sensing devices are useful in determining the true traffic patterns on long- span bridges, particularly traffic frequency of trucks and other heavily-laden vehicles. WIM sensors are more efficient than static weigh stations in that they do not require vehicles to stop in order to get weighed. Trucks can travel at highway speeds and their weight can be captured within a few percent tolerance of the actual weight, depending on the type of sensor used. It is important that WIM devices are placed in fixed locations that will not be adversely affected by vibration or electromagnetic disturbances. Static weigh stations have been an effective tool in monitoring truck weights and immediately apprehending overweight vehicles on roadways and bridges. While WIM devices have been used for enforcing truck size and weight, their main use is as diagnostic tools for bridge and pavement design, monitoring, and research; regulation development; fatigue analysis and remaining useful life of steel structures; and vital statistics and planning. WIM data is typically collected and filtered for outlying points, or filtered for all points above or below a set threshold limit. There are several commonly used WIM sensing devices currently in use: . Piezoelectric Sensor - This is the most common WIM sensor currently in use, and it is embedded in the pavement. It produces a charge that is equivalent to the deformation induced by the tire loads on the pavement’s surface. It is common to install two inductive loops and two piezoelectric sensors in each monitored lane. A properly installed and calibrated piezoelectric WIM system yields gross vehicle weights that are within 15% of the actual vehicle weight for 95% of the measured trucks. . Bending Plate - The bending scale consists of two steel platforms that are 2 ft x 6 ft, adjacently placed to cover a 12-ft lane. The plates are instrumented with strain gauges, which measure tire-load-induced plate strains. The measured strains are then analyzed to

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determine the tire load. A properly installed and calibrated bending plate WIM system yields gross vehicle weights that are within 10% of the actual vehicle weight for 95% of the measured trucks. . Single Load Cell - This device consists of two 6 ft x 6 ft platforms placed adjacently to cover the 12-ft monitored lane. A single hydraulic load cell is installed at the center of each platform to measure the tire-load-induced forces that are then transformed into tire loads. A properly installed and calibrated single load cell WIM system yields gross vehicle weights that are within 6% of the actual vehicle weight for 95% of the measured trucks. Advantages of WIM Sensing Systems over Static Scales: 1. High processing rate - Trucks can be weighed as they travel at highway speeds, resulting in a significantly greater number of counted vehicles in a short period of time compared to static weigh stations. 2. Safety - The minimization of static weighing significantly decreases vehicle accumulation at highway lanes leading to weight stations. 3. Continuous data processing - WIM can be performed continuously rather than static weighing, which uses samples from the traffic stream. Continuous sampling eliminates inherent data bias in the sampling for static weighing. 4. Increased coverage and lower cost - More sites may be monitored with WIM at the same cost. 5. Minimized scale avoidance - WIM can monitor truck traffic without alerting truck drivers. This results in more truthful data as overweight trucks are less likely to avoid weigh stations. 6. Dynamic loading data - Unlike static weigh stations, WIM sensing devices can record dynamic axle load information (impact), which can be significantly greater than static load information. Disadvantages of WIM Sensing Systems over Static Scales: 1. Less accurate - WIM systems are less accurate than static scales. According to the National Bureau of Standards, wheel load scales are required to have an accuracy of ±1% when tested for certification and must be maintained thereafter at ±2%. The best accuracy obtained with the most expensive commonly-used WIM devices is 6% of actual vehicle weights for 95% of measured trucks. 2. Reduced information - Truck information that is easily collected at static weight stations such as fuel type, state of registry, model year, loaded or unloaded status, origin, and destination cannot be obtained with typical WIM systems. 3. Susceptibility to damage from electromagnetic transients - WIM systems are sensitive to electromagnetic disturbances caused mostly by lightning strikes in the vicinity of the equipment. Particular use can be made of WIM systems for long-span bridges. WIM data can pinpoint truck loading that may favor a particular bridge lane due to geometric idiosyncrasies such as lane width, roadway curvature, superelevation, or presence/absence of a median barrier. This is important when assessing loading and impact damage to individual bridge members. Live-streaming WIM

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data can be used to give the frequency of heavy loads at all times and days of the week, and can be used in estimates of fatigue life. Site-specific WIM data can be used to justify a reduction in design live load for a bridge crossing in lieu of using standard live load lane reductions from design guidelines, and may also affect bridge load rating computations due to more realistic loading and impact data.

E 3.4 Strain Gauges

Strain gauges can be used for a multitude of functions on steel and concrete long-span bridges. They are used to measure local static strain, curvature, concrete creep and shrinkage. The most common kind of strain gauge consists of an insulating flexible backing which supports a metallic foil pattern. For smaller measurements of strain, a semiconductor, or piezoresistor can be used. Semiconductor gauges are normally more temperature sensitive, and more fragile than foil-type strain gauges. Both types are normally affixed to the structure using a cyanoacrylate adhesive. Both of the common types of electronic strain gauges, electrical resistance and vibrating wire, have flaws: electrical resistance strain gauges can measure dynamic strain but have low zero-stability, which results in drift over time. Vibrating wire strain gauges have high zero-stability but can only be used for quasi-static measurement. Mechanical strain gauges are often used to measure larger movements of foundations or crack growth. The two halves of the gauge (one opaque half and the other half of clear material with an embossed measurement grid) are affixed to opposite sides of an existing or suspected crack or location of movement. Relative vertical and horizontal movement can be measured over time. Sensitivity of strain gauges will vary by the material used, temperature range, and the amount of strain measured. Constantan alloy is the oldest and most widely used material in strain gauges since it has a relatively low temperature-induced strain in the range of -30°C to +193°C (-20°F to +380°F). It is thus considered to have self-temperature compensation. Its almost constant sensitivity across a wide range of strain makes it particularly suitable for a wide range of applications. However, the resistance of Constantan drifts continuously at temperatures above +65°C (+150°F), which can become troublesome over a long period of time or at high temperature. Constantan’s sensitivity is higher than average, at 2.1 (sensitivity is proportional to approximately 1+ 2 times Poisson’s ratio, which is in the range of 0.25 to 0.35 for most metals encountered on bridges, so a normal sensitivity range would be 1.5 to 1.7). Isoelastic alloy is widely used for dynamic strain measurements for vibration and impact. Its sensitivity is 3.6, higher than that of Constantan, which gives very good signal to noise ratio. Its resistance is generally around 350 Ω compared to approximately 20 Ω for Constantan, which also increases strain sensitivity. It also has better fatigue properties than many strain gauge materials. Unfortunately, it is normally unsuitable for long-term monitoring with temperature fluctuations due to the fact that, unlike Constantan and Karma (see below), it does not possess self-temperature compensation properties. Karma alloy has similar overall properties to Constantan, including self-temperature compensation in the range of -73°C to +260°C (-100°F to +500°F), but it has a higher cyclic strain resistance than Constantan. Strain gauge sensitivity factors are usually provided by the manufacturer, but bridge design engineers need to choose the right gauge wire for their particular application. Strain gauges can be used on fatigue-sensitive details on existing or new long-span bridges to measure stress/strain

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for fatigue cycles, or out-of-plane bending. Strains greater than a pre-set maximum value can trigger an alarm, or flag for investigation or response, by the bridge owner. Strain gauges are known for their relatively short durability in field applications. More recent applications for strain measurement have included fiber optic sensors as a potentially more durable method for long-term structural health monitoring applications. E 3.5 Global Positioning System (GPS)

Using global positioning system (GPS) data to monitor bridge health can be useful for recording bridge dynamic behavior during normal traffic flows or during extreme events. It can be used in conjunction with accelerometers to provide a more comprehensive measuring system of bridge movements. Conventional sensors become impractical for long-span bridges for long-term monitoring of absolute displacement. Traditional displacement transducers are useful for relative displacement only, whereas laser transducers and total stations have been proven unreliable over the long term for monitoring purposes9. Currently, GPS monitoring is used for this application. GPS can be used to monitor vertical and horizontal displacement of tower tops on cable supported bridges, or midspan deck vertical deflections for many types of long-span bridges. It can also be used for monitoring relative horizontal displacements across large expansion joints for long-span bridges, and verifying creep/shrinkage and movements due to temperature changes. However, satellite positioning is crucial to accurate measurements, and the current GPS satellite distribution across the sky is uneven in the mid- and high latitude regions. Due to this uneven distribution of GPS satellites, the precision of position measurements varies with the latitude position of the satellites11, 13. Therefore, it is important to note that the inherent issues with measurement accuracy and current satellite distribution are hindrances for a GPS system. However, these issues are known and accepted and can be further augmented with earth-based systems to meet bridge structural health monitoring needs. Thus, an earth-based pseudolite or Galileo system could be used to augment a GPS/accelerometer measuring system7.

E 3.6 Acoustic Monitoring

Acoustic monitoring is a continuous non-intrusive monitoring method that is used to detect and locate failures in steel wire, strand and cable. The response of the structure is measured, caused by energy released when a bridge wire fails or another event of interest occurs. Its uses in long- span bridges are chiefly in the main cables of suspension bridges, stay cables on cable-stayed bridges, pre-stressing strands in concrete structures, and reinforcement strands in post-tensioned concrete and steel structures. It currently complements or augments the recommended frequency of visual main cable examinations on suspension bridges in NCHRP Report 534 - Guidelines for Inspection and Strength Evaluation of Suspension Bridge Parallel-Wire Cables, and is mainly used when corrosion or embrittlement problems are found during an in-depth investigation of the wires of the main cable. For other strands that are not able to be visually inspected, acoustic monitoring sensors are an important tool in bridge health monitoring. Sensors are placed at strategic points on the bridge cables or strands, and one or more wires may be cut sacrificially (and spliced with repair wire, if accessible) to simulate wire breaks and to distinguish breaks from the sound of other simulated noises such as knocking or hammering. Insignificant noises are filtered using screening software, and wire failures or other events of interest can be identified by date, time, and location/coordinates

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on the suspension bridge cable, stay cable, or pre-stressing or post-tensioning strand. Data acquisition systems are often located on or near the bridge site. Automatic notification of failure events is available by bridge owners through secure websites. Acoustic monitoring has been performed on a number of suspension bridges, including the Bear Mountain Bridge in New York. Some acoustic monitoring methods are patented technologies.

E 3.7 Peak Displacement Sensor

Displacement gauges were developed specifically for the Akashi Kaikyo Bridge (Japan) based on needs determined during the design phase of this record-breaking suspension bridge. One channel of the dual-output gauge measures relative displacement to a high degree of precision, and the other channel records peak displacement and passively retains the data value for future interrogation. Thus, displacements during heavy vehicular loads or extreme wind/seismic events can be recorded and peak displacement values during the event can be called up. Sensors require no electrical power except for reading stored displacement values1. An example of a peak displacement application is shown in Figure E91 below.

APPLICATION OF PEAK DISPLACEMENT SENSOR TO MONITOR PEAK DISPLACEMENT OF BRIDGE PEDESTAL DAMPER Figure E 9 Application of Peak Displacement Sensor to Monitor Peak Displacement of Bridge Pedestal Damper

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E 3.8 Elasto-Magnetic Stress Sensor

Stress sensors have been developed that can reliably monitor the true stress in tendons and cables, particularly applied to stay cables. The measuring principle is based on the permeability of ferromagnetic materials being a function of magnetic history and applied field (stress and temperature). This type of sensor was developed and applied to the stay cables on the Tatara Bridge (Japan)1. An example of an elasto-magnetic sensor application is shown in Figure E101. Three elasto-magnetic sensors were installed at the anchors of the cables on the Penobscot River Bridge (Maine, USA) to monitor cable forces during and after construction. Elasto-magnetic sensors have been installed on the Stonecutters Bridge (Hong Kong) to monitor the tensions of post-tension concrete girder, at the interface between concrete girder and steel girder24.

APPLICATION OF ELASTO-MAGNETIC STRESS SENSOR FOR MONITORING TRUE STRESS IN BRIDGE CABLES

F igure E 10 Application of Elasto -Magnetic Stress Sensor for Monitoring True Stress in Bridge Cables

E 3.9 Fiber Optic Sensors

Fiber optic sensors can be used for their capabilities in monitoring geometric conformity and for sensing of a variety of perturbations. Applications include strain, crack development and growth monitoring in bridge substructures, deformations, accelerations, and cable dynamics. Most current sensor designs are based on alterations in light wavelength or frequency, including fiber Bragg gratings (FBG), and interfermometric sensors based on Fabry-Perot, white light, and others19. The attraction of fiber optic sensors is their capability of distributed sensing and measurement, in parallel or in series, which can result in elaborate structural health monitoring of long-span bridges. Another attractive feature of fiber optic sensors is their ability to be embedded within bridge cables for both temperature and strain measurement19.

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Protection of the fibers against damage is a key issue for the design of a structural health monitoring system, with soft PVC protective materials for the fibers being the current state of the art. Fiber optic sensors have been used to monitor cable strains and deformations for the pedestrian Lingotto Bridge (Italy), built for the 2005 Winter Olympics in Turin (see Figure E1119). This cable stay bridge with a main span of 150 meters is supported by a single arch. The cable sensors were constructed using an individual FBG that was pre-tensioned and protected inside a PVDF tube. Details of this application are shown in Figures E1219, E1319 and E159 below.

VIEW OF THE TURIN, ITALY LINGOTTO PEDESTRIAN CABLE- STAYED BRIDGE

Figure E 11 Vie w of the Turin, Italy Lingotto Pedestrian Cable-Stayed Bridge

FIBER OPTIC SENSOR ON A CABLE OF LINGOTTO BRIDGE

Figure E 12 Fiber Optic Sensor on a Cable of Lingotto Bridge

23 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

SCHEMATIC OF THE FBG CABLE SENSOR ON LINGOTTO BRIDGE

Figure E 13 Schematic of the FBG Cable Sensor on Lingotto Bridge

Fiber optic sensors have also been installed along the length of the deck on the Jiangyin suspension bridge (China) for both strain and temperature measurement (see Figure E149 below).

LAYOUT OF FIBER OPTIC SENSORS ON THE JIANGYIN BRIDGE (CHINA) Figure E 14 Layout of Fiber Optic Sensors on the Jiangyin Bridge (China)

24 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

FIBER OPTIC SENSORS EMBEDDED WITHIN A BRIDGE CABLE CROSS SECTION

Figure E 15 Fiber Optic Sensors Embedded Within a Bridge Cable Cross Section

Fiber optic crack sensors have also been successfully tested for monitoring cracks on test material subjected to shake table tests, as shown below. Fiber optic crack sensors have been installed to monitor cracks in masonry vaults and on a bridge pier plastic hinge zones, as shown in Figures E1619 and E1719.

A FIBER OPTIC CRACK SENSOR. NOTE THE PVC PROTECTIVE MATERIAL AROUND PARABOLIC CURVE OF THE SENSOR

Figure E 16 A Fiber Optic Crack Sensor

25 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

CRACK SENSORS USED TO MONITOR A BRIDGE PIER

Figure E 17 A Crack Sensor Used to Monitor a Bridge Pier

E 3.10 Weather Sensors: Temperature Sensors, Anemometers, Barometers, Rainfall Gauges, and Hygrometers

E 3.10.1 Temperature Sensors Temperature sensors can be used to measure ambient temperature, wet-bulb and dry-bulb temperature for determining relative humidity, and determining temperature of steel and concrete. Temperature sensors are mounted in the pavement on approaches to long-span bridges as well as in the pavement of the long-span bridge (often at midspan), to help assess roadway freezing temperatures in order to optimize the use of roadway salt for deicing in Northern climates. Temperature extremes can be used to automatically notify the bridge owner’s staff of possible issues with expansion joint movements, or possible kinks in roadways at pinned joints on cable-supported bridges. Temperature sensors can be used to measure ambient temperature, wet-bulb and dry-bulb temperature for determining relative humidity, and determining temperature of steel and concrete. Temperature sensors are mounted in the pavement on approaches to long-span bridges as well as in the pavement of the long-span bridge (often at midspan), to help assess roadway freezing temperatures in order to optimize the use of roadway salt for deicing in Northern climates. Temperature extremes can be used to automatically notify the bridge owner’s staff of possible issues with expansion joint movements, or possible kinks in roadways at pinned joints on cable-supported bridges.

26 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

E 3.10.2 Anemometers Anemometers measure wind speed. Anemometers have been developed based on two measuring principles: wind velocity and pressure, which can be related to speed. Anemometers that measure wind speed comprise various types: cup; windmill; hot-wire; laser Doppler; and sonic. Pressure measurement anemometers consist of the following types: plate; and tube. The sonic anemometer, which measures wind speed, lacks moving parts, which makes it well- suited for long-term use in exposed weather stations or weather buoys, where accuracy and reliability are important. One key concern for this type of anemometer is that the air flow is distorted by the structure supporting the transducer, which requires a correction based on wind tunnel tests. ISO norm 16622 – Meteorology - Sonic anemometers/thermometers – Acceptance test methods for mean wind measurements are generally used for this correction. Two-dimensional sonic anemometers (wind speed and direction) are used in weather stations and weather buoys, ship navigation, and aviation. Plate anemometers measure the pressure exerted by the wind force on a plate by use of the compressive force on a spring with a given spring constant. Plate anemometers have been used for high wind alarms on bridges, although they do not respond well to light winds, can be inaccurate for high wind readings, and are sometimes slow at responding to variable-speed winds. Anemometers have been placed on numerous bridges as part of weather station monitoring; they have also been integrated into the data collection algorithm on numerous cable-supported long-span bridges, including the Akashi Kaikyo suspension bridge (Japan), the Tatara cable- stayed bridge (Japan), and the Donghai cable-stayed bridge (China).

E 3.10.3 Barometers

A barometer measures atmospheric pressure. Changes in atmospheric pressure can forecast short term changes in the weather, that may be useful for long-span bridge structural health monitoring. Barometers, as well as anemometers, temperature readouts, hygrometers, and other weather sensors may be used as input to revise the frequency of data collection on other areas of the bridge (such as accelerometers or displacement transducers) due to changing weather patterns.

Electronic barometers offer higher accuracy and greater recording length than the older barograph (a mechanically-based barometer with analog recording of pressure data) and the ability to perform further data analysis on the captured data, including automated use of the data to forecast the weather. High-quality electronic barometers are hermetically sealed with all stainless steel construction which makes them ideally suitable for harsh environments. Their high operating temperature, broad compensated range and excellent temperature compensation allow for stable readings in bridge applications.

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E 3.10.4 Pluviometers (Rainfall Gauges)

Electronic rain gauges typically operate on the “tipping bucket” principle and meet national Weather Service specifications for statistical accuracy. Every time the bucket tips, a count is transmitted to the display and the gauge empties. Long-life batteries for electronic rain gauge units typically last 6 months to a year. Some issues with this type of rain gauge include the need for heaters to prevent freezing, and spikes to prevent bird nesting, since the measurement system is open to the atmosphere.

Other types of rain gauges sense water hitting the outside surface of a hemisphere using beams of infrared light. These types of sensors are normally totally enclosed and contain no mechanical parts, reducing the possibility of infiltration by insects. These sensors have a heater for cold weather operation, and compensate for surface grime or other degradation (although they need to be protected during maintenance painting), so maintenance is normally not needed. There are no exposed conductors to corrode. This type of sensor can be programmed to detect the first raindrop.

Rain gauges are often installed as part of a bridge’s weather station monitoring system, such as on the Stonecutter’s Bridge (China) and HSWC Bridge (China). Wind and rain can induce vibrations in the stays of cable-stayed bridges. E 3.10.5 Hygrometers

Hygrometers measure relative humidity, which can be calculated by measuring dry bulb and wet bulb temperature and consulting a psychrometric chart. Modern electronic devices use temperature of condensation, changes in electrical resistance, and changes in electrical capacitance to measure humidity changes.

Cooled-mirror dewpoint hygrometers are among the most precise instruments available. Dewpoint is the temperature at which a sample of moist air (or any other water vapor) at constant pressure reaches water vapor saturation. At this saturation temperature, water condenses with further cooling. Cooled mirror dewpoint hygrometers use a chilled mirror and an optoelectronic mechanism to detect condensation on the mirror surface. The temperature of the mirror is closely controlled by electronic feedback to maintain a dynamic equilibrium between evaporation and condensation on the mirror, thus accurately measuring the dewpoint temperature.

Other modern instruments record information using electronic means. The two most common electronic sensors are capacitive or resistive. Capacitive sensors sense water by applying an AC signal between two plates and measuring the change in capacitance caused by the amount of water present. The resistive sensors use a polymer membrane which changes conductivity according to absorbed water. Recently, an unbalanced AC Bridge approach was adapted for low power/energy operation and has shown to provide better measurement performance over a wide operating range. To further increase accuracy in this same device, which combines a sensor with a data logging instrument, a calibration method utilizing a large memory array was developed to maximize performance. In most instruments, resistive sensors can be read by common meters or data acquisition boards. Temperature must also be measured, as it affects the calibration of all of these sensors.

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E 3.11 Digital Video Cameras

Digital video cameras have multiple capabilities for use on long-span suspension bridges for monitoring structural health as well as security and traffic flow. Cameras in use today are color, high resolution devices, with pan/tilt/zoom features as standard requirements. Current uses for video cameras are for monitoring traffic flow in traffic centers, and for monitoring security of sensitive areas on long-span bridges. Video cameras have also been integrated with Weigh-in- Motion sensors to take short-term video, to match heavy axle loads to corresponding strain gauge data. Digital cameras have the ability to be accessed and controlled manually via secure internet websites.

The design of a comprehensive structural health monitoring system can incorporate digital video cameras to document bridge movements over time or under extreme seismic and weather-related events.

E 3.12 Corrosion Cells

Corrosion of steel is an electrochemical process, and electrochemical instrumentation can be used to monitor the corrosion process. Many electrochemical techniques have been developed over the years especially for the measurement of corrosion processes. Potentiostats/galvanostats are the main type used, due to the low noise level required to sense very low currents. Potentiostat/galvanostat models vary by sensitivity, with those measuring higher currents more useful for higher corrosion rates. Anode ladder-type sensors have been used in the Storebaelt Bridge (Denmark) 22.

Detecting the corrosion of steel rebar in concrete roadway decks is of most concern, particularly in the Northern US where chlorides from roadway deicing salts can infiltrate concrete. Chlorides reduce the pH of the concrete over time and thereby can destroy the passivation layer between steel rebar and concrete if the localized pH dips below 11.5.

Corrosion sensors have been placed at the base and top of the towers of the Sutong Bridge (China), at the base of the main towers of the Stonecutters Bridge (Hong Kong), and at the tower base of the HSWC Bridge (Hong Kong)9, 23, 20 . Corrosion sensors have also been installed in the deck of the new St. Anthony Falls Bridge (Minneapolis, MN, USA)10, 21. Since deck replacement is a key issue on cable-supported bridges and long-span bridges in general, there should be numerous additional applications for corrosion cells on the concrete or composite decks of long- span bridges.

E 3.13 Tiltmeters

Tiltmeters can be used on long-span bridges to monitor angular deflection, which can be related to deflection, moment, and shear forces by computation; monitor bridge pier settlement; monitor the performance of expansion joints; and monitor overall bridge behavior during seismic and wind events. generate an artificial horizon and measure angular tilt with respect to this horizon. Tiltmeters can also be mounted to portions of a bridge structure under water. Tiltmeters can be specified for narrow-angle or wide-angle measurement, depending on the expected application. Typical narrow-angle tiltmeters have a measurement range on the order of +/- 40 arc minutes and can detect changes as small as one arc second (0.00028 degree). Typical wide-

29 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix E

angle tiltmeters measure tilt over a range of +/- 10 degrees and can detect changes as small as 0.0024 degree. One example of a tiltmeter application would be to measure angular deflection of cable bents on suspension bridges. This can be correlated to the inclination of the bent at the design temperature for monitoring bridge behavior. Similar applications are on the suspended superstructure of cable- supported bridges to monitor the horizontal and vertical effects of wind events; traffic live loads at expansion joints; or pinned joints in the bridge superstructure. Where the integrity of a bridge’s substructure is of concern, tiltmeters can be used on abutments or piers to monitor settlement of pier stems or footings, above or below the water line.

E 3.14 Infrared Thermography (IR)

Infrared thermography, developed for security and military applications, is also used as a non- destructive technique for bridges. While not widely automated in the US at this time, it holds potential for automation and is discussed here. Thermal radiation from objects is depicted on a color scale, so that warm objects stand out from their cooler environments, or vice versa. The technique can be used to detect differences in material properties caused by infrared thermal gradients. For structural applications, it can be used to find deteriorating mechanical or electronic components (due to higher temperature) prior to failure, and it can identify delaminated or cracked surface concrete due to thermal differences. Since delaminated or cracked concrete often fails at the rebar mat closest to the surface, if the rebar mat is embedded too deeply (often due to concrete or asphalt overlays) the method may fail to detect concrete failures in some applications. The method is only useful in detecting surface temperatures, has an accuracy of ±2% at best, and is not as accurate as contact temperature measurement methods. Some state transportation departments have experience using IR and ground penetrating radar for cost-effective concrete roadway deck assessments. A two-level strategy is used to first implement IR and GPR to assess the deck condition. IR thermography is used in a semi-automated process to survey the deck, one lane per pass of the survey vehicle, at a maximum speed of 5 mph. Data collection is typically between 10 a.m. and 3 p.m. to highlight temperature differentials due to debonding or delamination of concrete. GPR is used in the same survey vehicle in multiple passes of the deck with data lines spaced at 3-foot intervals. GPR requires dry pavement, but has no restrictions on temperature; GPR scanning is normally done at times and on days when the deck cannot be IR scanned. Occasionally this first-level assessment includes visual inspection of the deck underside for evidence of efflorescence, or concrete cores are taken if additional evidence of the deck condition is needed. The Wisconsin Department of Transportation (WisDOT) has used a combined program of IR/GPR imaging for bridge deck evaluation and had an average 83% success rate of identifying trouble spots, a 17% false alarm rate, and a 5% miss rate (results were verified by coring)3.

E 3.15 Bearing Sensors

Mechanical or rotating bridge bearings operate under arduous conditions - often for considerable periods of time between maintenance turnarounds. For bearings that rotate or slide on a frequent basis, a reliable indicator of bearing condition may be the localized temperature of the bearing metal beneath the shoe. Displacement sensors may be useful for extreme movements due to ambient temperature or earthquake. Displacement sensors may also be useful to measure increases or reductions in movement or rotation due to excessive wear or corrosion build-up.

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E 3.16 Scour Detection

Automated scour detection and measurement is still an emerging technology. Low-cost fathometers have been tested to measure scour, although their use is predicated on low sediment levels and no icing conditions, both of which may be issues for long-term reliable use. Buried or driven rods that use a sliding collar can be installed by driving, auguring, or excavating and burying the device adjacent to a bridge pier. Research needs to be undertaken to develop piezoelectric film, mercury tip switches, and magnetic switches in order to link these types of scour detection devices to data-logging equipment. Research has indicated that buried or driven devices do not enhance or reduce scour effects at a pier. Other avenues of research include buried devices tethered to a bridge pier with a motion-activated transmitter that would be activated once sufficient scour has released the sensor from its buried position27. More recent research involves using time domain reflectometry (TDR) for automated scour detection. An algorithm for interpreting scour signals was recently developed and a prototype tested, in order to validate the prototype’s performance in detecting scour and assessing the and density of sediments28. Recent work in Taiwan has combined vibration technology with several other sensor systems to form an intelligent sensor system that directly monitors structural stability of bridge substructures29, based on the principle that the static and dynamic stiffness of the substructure changes when scour, flood, or debris flow occurs.

E 4.0 Information Transfer - Fiber Optic Networks/Wireless Communication

Fiber optic networks are currently used to transfer bridge health monitoring data from bridge sites to centralized collection points or to data management centers. Health monitoring information relayed by fiber optic networks include real-time camera video images, strain gauge data, and weigh-in-motion data, etc. Some countries such as Japan have fiber optic networks laid along major national highways, and the incremental cost of sending data from a bridge site through the existing network is normally small, because the initial costly fiber optic infrastructure is already in place8. Fiber optic networks in major cities like New York are connected specifically from traffic cameras and other sensing devices on bridges to traffic management centers for monitoring traffic conditions so that variable message signs (VMS’s) can be updated with the latest traffic information. Other bridge authorities use antennae and wireless transfer of live-cam video streaming from security cameras, which can be accessed from any computer with internet access and the proper security passwords. The remote user can use the pan-zoom-tilt functions on the cameras using the wireless website connection. Some bridge authorities use wireless technology for motion lights around critical or secured areas. In some instances, the high frequency of nuisance alarms may make their use ineffective. A bridge authority in the northeastern U.S. decided to forgo the use of alarms on motion lights at the remote suspension bridge anchorages due to the concern over false alarms from gulls, falcons, and other wildlife frequenting the remote bridge anchorage areas. Emerging technologies for wireless communication include performing data filtering and sorting locally at the sensor site, to reduce remote battery usage, and for using the bulk of battery power for transmitting only essential data (which usually requires more energy than performing local data filtering tasks). Solar-powered and vibration-powered technologies for sensor batteries are currently being developed.

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REFERENCES

1 Sumitro, Sunaryo. “Current and Future Trends in Long-Span Bridge Health Monitoring System in Japan.” A workshop Sponsored by the National Science Foundation on Health Monitoring of Long- span Bridges, University of California, Irvine Campus, March 9-10, 2001. 2 Mehrabi, A. B., and A. T. Ciolko. “Health monitoring and problem solving for cable supported bridges.” 3 Maser, K. R., and B. C. Miller. “Multi-Level Bridge Deck Evaluation Using Combined NDT Methods.” International Bridge Conference, Pittsburgh, PA, June 14-17, 2009. 4 Sun, Limin, Qiwei Zhang, Airong Chen, and Zhixing Lin. “Cable Vibration Control Countermeasures and Structural Health Monitoring System Design of Sutong Bridge.” 2nd International Conference on Bridge Maintenance, Safety, and Management, October 18-22, 2004, Kyoto, Japan. 5 Ciolko, Adrian, Varsha Singh, and Thomas Weinmann. “Spy in the Wire.” Bridge Design and Engineering, Fourth Quarter 2005, pp. 54-55. 6 Spence, J.W., and F.H. Haynie. “Pitting of Galvanized Steel in Controlled Clean Air Environments.” ASTM STP 576, American Society for Testing and Materials, 1976, pp. 132-146. 7 Meng, Xiaoling, Gethyn Wyn Roberts, Emily Cosser, Alan Henry Dodson. “Real-Time Bridge Vibration and Deflection Monitoring Using an Integrated GPS/Accelerometer/Pseudolite System.” Proceedings, 11th FIG Symposium on Deformation Measurements, Santorini, Greece, 2003. 8 Kobayashi, Yuusuki, and Chitoshi Miki. “Bridge Monitoring System with Fiber-Optic Communications Network.” The Eight East Asia-Pacific Conference on Structural Engineering and Construction, December 5-7, 2001, Nanyang Technological University, Singapore. Paper No. 1278. 9 Ko, J.M. and Y.Q. Ni. “Technology developments in structural health monitoring of large-scale bridges.” Engineering Structures 27 (2005) pages 1715-1725. 10 Anaudi, Daniele. “Overview of 40 Bridge Structural Health Monitoring Projects.” 11 “A Bridge Health Monitoring System Based on NI Hardware and Software.” 12 He, X., B. Moaveni, J. J. Conti, A. Elgamal, S, F. Masri, J. P. Caffrey, M. Wahbeh, F. Tasbihgoo, and D. H. Wang. “System Identification of New Carquinez Bridge Using Ambient Vibration Data.” International Conference on Experimental Vibration Analysis for Civil Engineering Structures, Bordeaux, France, October 26-28, 2005. 13 Larocca, Ana Paula C. “Using High-rate GPS Data to Monitor the Dynamic Behavior of a Cable- Stay Bridge.” 14 Vlamis-Stathopoulos, Gilles Hovhanessian, and Benoit Kroely. “Monitoring System Behavior During Earthquake.” 15 Karbhari, V.M. and L.S-W. Lee. “Vibration-based damage detection techniques for structural health monitoring of civil infrastructure systems.” Structural Health Monitoring of Civil Infrastructure Systems, Woodhead Publishing, Ltd. and CRC Press, 2009.

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16 Karbhari, V.M., H. Guan, and C. Sikorsky. “Operational modal analysis for vibration-based structural health monitoring of civil structures.” Structural Health Monitoring of Civil Infrastructure Systems, Woodhead Publishing, Ltd. and CRC Press, 2009 17 Mehrabi, Armin B. “Cass Street Bridge Hanger Force Verification.” Report to Wisconsin DOT by Construction Technology Laboratories, Inc., June, 2004. 18 Wong, Kai-Yuen. “Design of a structural health monitoring system for long-span bridges.” Structure and Infrastructure Engineering, Volume 3, Issue 2, June, 2007, Pages 169-185. 19 Ansari, F. “Fiber optic sensors for structural health monitoring of civil infrastructure systems.” Structural Health Monitoring of Civil Infrastructure Systems, Woodhead Publishing, Ltd. and CRC Press, 2009. 20 Wong, K.Y. and Y.Q. Ni. “Structural health monitoring of cable-supported brides in Hong Kong,” Structural Health Monitoring of Civil Infrastructure Systems, Woodhead Publishing, Ltd. and CRC Press, 2009. 21 Inaudi, D. “Structural health monitoring of bridges”, Structural Health Monitoring of Civil Infrastructure Systems, Woodhead Publishing, Ltd. and CRC Press, 2009. 22 Habel, W.R. “Structural health monitoring research in Europe: trends and applications,” Structural Health Monitoring of Civil Infrastructure Systems, Woodhead Publishing, Ltd. and CRC Press, 2009. 23 Ou, J. and H. Li. “Structural health monitoring research in China: trends and applications,” Structural Health Monitoring of Civil Infrastructure Systems, Woodhead Publishing, Ltd. and CRC Press, 2009. 24 Wang, M.L. “Magnetoelastic stress sensors for structural health monitoring of civil infrastructure systems,” Structural Health Monitoring of Civil Infrastructure Systems, Woodhead Publishing, Ltd. and CRC Press, 2009. 25 Aktan, A.E., A.J. Helmicki, and V.J. Hunt. “Instrumentation and Intelligence Issues in Bridge Health Monitoring,” SPIE, Volume 2446, pages 106-115. 26 Shaw, Donald E., James H. Garrett, Jr., Jacobo, Bielak, and Fernando Cerda. “A Holistic Approach to Structural Health Monitoring of Bridges”, International Bridge Conference 2009, Pittsburgh, PA. 27 Lagasse, P.F., E.V. Richardson, and J.D. Schall. “Fixed Instrumentation for Monitoring Scour at Bridges”, Transportation Research Record 1647, Paper No. 98-0057, NCHRP Project 21-3. 28 Yu, Xinbao. “Time Domain Reflectometry Automatic Bridge Scour Measurement System: Principles and Potentials”, Structural Health Monitoring Vol. 8, No. 6, 463-476 (2009). 29 Ko, Y.Y., W.F. Lee, W.K. Chang, H.T. Mei, C.H. Chen. “Scour Evaluation of Bridge Foundations Using Vibration Measurement”, “Scour and Erosion – Proceedings of the Fifth International Conference on Scour and Erosion”, Susan E. Burns, Shobka K. Bhatia, Catherine M.C. Avila, and Beatrice E. Hunt, editors, ASCE (2011), pages 884-893.

33 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

TABLE OF CONTENTS

F 1.0 General ...... 1 F 2.0 De-coupled Soil-Structure Interaction Analysis ...... 1 F 3.0 Fully Coupled Soil-Foundation Interaction Analysis ...... 4 F 4.0 Simplified Design Charts ...... 9 F 5.0 Pile / Pile Group Analysis ...... 18 F 6.0 Equivalent Stiffness of Shallow Foundations (Approximate Methods) ...... 20 F 7.0 Analysis of Interaction Between Component Force Effects (Axial, Shear and Bending) ...... 26

i FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

LIST OF FIGURES

Figure F 1 Complete System ...... 3 Figure F 2 Substructure No. 1 with Foundation Interaction Force ...... 3 Figure F 3 Foundation Impedance; Fb(ω) = Gb(ω) U(ω) ...... 3 sb Figure F 4 Scattered Foundation Motions; U bj (t), j = 1.4 ...... 4 Figure F 5 Conceptual Model for Substructure No. 1 with Foundation Driving Forces ...... 4 Figure F 6 Soil-Foundation-Structure Interaction: Fully-Coupled Gravity Caisson Modeling ...... 6 Figure F 7 Soil-Foundation-Structure Interaction: Example of Fully-Coupled Gravity Caisson Modeling ...... 7 Figure F 8 Soil-Foundation-Structure Interaction: Fully-Coupled Modeling of Piles or Drilled Shafts ...... 8 Figure F 9 Soil-Foundation-Structure Interaction: Example of Fully-Coupled Modeling of Drilled Shafts ...... 9 Figure F 10 Recommendations for Coefficient of Variation in Modulus with Depth for Sand ...... 10 Figure F 11 Recommendations for Coefficient of Variation in Subgrade Modulus with Depth for Clay ...... 11 Figure F 12 Lateral Stiffness of Free-Headed Pile ...... 12 Figure F 13 Coefficient for Lateral Pile Head Stiffness (Fixed Head Pile Lateral Stiffness)...... 13 Figure F 14 Coefficient for Pile Head Rotation ...... 14 Figure F 15 Coefficient for Cross-Coupling Term ...... 15 Figure F 16 Comparison of Fixed Pile Head Stiffness at Various Embedments ...... 16 Figure F 17 Comparison of the Rotational Stiffness Coefficient at Various Embedments ...... 17 Figure F 18 Comparison of the Cross-Coupling Stiffness Coefficient at Various Embedments ...... 18 Figure F 19 Procedure for Calculating Equivalent Radius of a Rectangular Footing ...... 22 Figure F 20 Shape Factor for Rectangular Footings ...... 23 Figure F 21 Embedment Factor for Rectangular Footings ...... 24

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LIST OF TABLES

Table F 1 “p-Multiplier” Values for Laterally Loaded Pile Group Analysis ...... 20 Table F 2 Equivalent Damping Ratios for Rigid Circular Footings ...... 25

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APPENDIX F – FOUNDATION MODELING AND SOIL-STRUCTURE INTERACTION

Topics selected are those deemed likely to be of immediate application for the readers of these Guidelines. Materials presented are available in the literature and provided to supplement the Guidelines as a convenient source of reference.

F 1.0 General

The following articles outline recommendations for modeling of soil-structure interaction (SSI) effects using the de-coupled and fully-coupled approaches, as well as other issues related to soil- foundation-structure interaction.

The emphasis in the following sections is on SSI effects under seismic loading. But the analytical approaches to foundation stiffness modeling described below can be (and often are) applied for analysis of non-seismic lateral loads as well, including wind load, vessel collision force, vehicular centrifugal force and deformation-based loads such as temperature, creep and shrinkage. Depending on the relative magnitude of non-seismic and seismic loads, nonlinear soil effects may be less pronounced for non-seismic loads.

Additional discussion and recommendations on soil-foundation-structure interaction analysis, both the de-coupled (substructure modeling) and fully-coupled approaches, and pile group analysis may be found in the literature, including the following:

• Ho, et. al., “Seismic Retrofitting Guidelines for Complex Steel truss Highway Bridges”, Chapter 4 - Structural Analysis (Ref. F11) • Parsons Brinckerhoff, Inc., Geotechnical Engineering Circular No. 3, Chapter 8 – Geotechnical Seismic Design for Transportation Structures and Soil-Foundation-Structure Interaction and Chapter 10 – Deep Foundations (Ref. F12) • Tseng, W-S., and Penzien, J., (2000), “Soil-Foundation-Structure Interaction”, Chapter 42, Bridge Engineering Handbook – Edited by W-F Chen & L. Duan, CRC Press (Ref. F17) • Lam and Law, “Soil Structure Interaction of Bridges for Seismic Analysis”, Report MCEER-00-0008 (F19)

F 2.0 De-coupled Soil-Structure Interaction Analysis

In the de-coupled soil-structure interaction analysis, the effect of foundation stiffness on structural response is examined by replacing the foundation in the structural model with a set of springs (or stiffness matrix). The effects of inertial loads from the superstructure and ground displacement loads on the foundation are approximated as an equivalent static load.

A foundation subject to dynamic excitation has six degrees of freedom (modes of motion): horizontal translation in two orthogonal directions; vertical motions; rocking about two orthogonal horizontal axes; and torsion (rotation) about the vertical axis. The stiffness of the foundation is first computed using a separate analysis using software such as FB-MultiPier or other commercially available software. The foundation model must account for the nonlinear effects of soils.

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The foundation stiffness is then reduced into a 6x6 stiffness matrix that correlates the forces (or moments) and deflections (or rotations) at the structure/foundation interface defined as follows.

where: Kxx = x-axis (longitudinal) translation stiffness coefficient Kyy = y-axis (transverse) translation stiffness coefficient Kzz = vertical translation stiffness coefficient Krx = x-axis rocking stiffness coefficient Kry = y-axis rocking stiffness coefficient Kt = torsional stiffness coefficient Kx-ry=Kry-x and Ky-rx=Krx-y are the cross coupling stiffness coefficients

Note that the above stiffness matrix shows that the foundation system is symmetric about two horizontal axes and the stiffness matrix is uncoupled in the horizontal and vertical directions.

For a grouped pile foundation with a rigid pile cap, the resistance to rocking is derived mainly by the axial forces (compression and tension) in the piles about the corresponding axis of rotation. The cross coupling between horizontal and rocking stiffnesses is typically small and may be neglected; i.e., Kx-ry=Kry-x=Ky-rx=Krx-y = 0. For single pile or single drilled shaft foundation, the cross coupling stiffnesses (Kx-ry, Kry-x, Ky-rx and Krx-y) are significant and should be considered in the analysis as the rocking is mainly resisted by the flexural stiffness of the piles.

It is important to develop the foundation stiffness matrix to include nonlinear effects of soils. This is typically the case for long span bridges, because most long span bridges can induce significant lateral deformations to soft marine clays or loose liquefiable river . One way to address this issue to perform several iterations between the foundation model analysis and the global (structural) model analysis, until force demands and displacements in the foundation model and global analytical model are consistent.

An illustrative example of how the de-coupled soil-foundation-structure interaction analysis approach (substructure modeling) can be incorporated in the global seismic analysis is presented in Figures F1 through F5 below.

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Figure F 1 Complete System (Ref. F1)

Figure F 2 Substructure No. 1 with Foundation Interaction Force (Ref. F1)

Figure F 3 Foundation Impedance; Fb(ω) = Gb(ω) U(ω) (Ref. F1)

3 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

sb Figure F 4 Scattered Foundation Motions; U bj (t), j = 1.4 (Ref. F1)

Figure F 5 Conceptual Model for Substructure No. 1 with Foundation Driving Forces (Ref. F1)

F 3.0 Fully Coupled Soil-Foundation Interaction Analysis

The coupled analysis examines the behavior of the entire soil-structure system simultaneously in a single, complex model, in which non-linear soil behavior is described by a continuum model and/or non-linear p-y springs. The figures below provide illustrative examples of how a fully- coupled soil-foundation-structure interaction analysis may be incorporated for caisson and pile (drilled shaft) foundations.

Seismic response of long span bridges is affected by the characteristic of the input ground motion which can be different among various support locations due to differing soil overburden conditions. Furthermore, ground motions can also vary at different elevations (depth) through the soil column at each support location. Therefore, care must be exercised in determining the effective depth for input ground motions (i.e. effective support motions). This is particularly important for deep foundations at soft soil sites. In the case of a fully-coupled SSI modeling approach using nonlinear time history analysis, the variation of ground motion inputs with depth can be handled directly by inputting different ground inputs (free-field motions) at each elevation (see Figure F6, below). Geotechnical Engineering Circular No. 3 (Ref. F12) provides a more detailed discussion on effective support motions.

4 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

The seismic response of a bridge subject to high seismic load demand is highly affected by damping characteristics of the soils and structures. For a fully-coupled dynamic analysis using a time history of input ground motions, damping is often represented as Rayleigh damping. Clough and Penzien (Ref. F18) present a detailed description of Rayleigh damping and its implementation in a global dynamic (seismic) analysis. The energy dissipation associated with highly nonlinear behavior of soils can be represented by the non-linear hysteretic loops of the p-y spring model used to characterize the soils.

It should be noted that the p-y curve methodology was developed based on pile load tests. Since lateral load test data on large gravity caissons are not available, care must be exercised in applying the free-field / p-y spring model to gravity caisson foundations. Extrapolation of the free-field / p-y curve model to gravity caisson foundations may require careful modification of both the free-field motions and the p-y springs.

Near-fault (near-field) directivity effects may be significant if a site is located sufficiently near to a fault and the earthquake fault ruptures toward the site, as a ground motion with a very large velocity pulse can occur. Because these kinds of pulses can be very damaging to long-period bridges, it is necessary to take special design measures in these situations. However, the methods used to account for these near-fault directivity effects vary from project to project, in part because the subject matter is still under research. In particular, the means for quantifying the velocity pulse and relating it to structural response is still a subject of research, although a time history analysis using appropriate time histories is one way this can be accomplished. The need to consider near fault directivity is, in general, limited to those states with well-defined shallow active faults (e.g. California, Washington, and Utah). Generally, fault directivity effects would need to be considered only for project sites within about 10 miles (15 kilometers) of the rupturing fault subject to relatively large magnitude earthquakes (M > 6.5).

Additional discussion and recommendations regarding consideration of free-field versus near- fault (near-field) effects can be found in the literature, including Geotechnical Engineering Circular No. 3, Chapter 3 – Seismic Hazard Analysis and Ground Motion Characterization (Ref. F12).

The fully-coupled approach to soil-foundation-structure interaction analysis is illustrated for a gravity caisson foundation in Figures F6 and F7, below.

5 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

Figure F 6 Soil-Foundation-Structure Interaction: Fully-Coupled Gravity Caisson Modeling

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Figure F 7 Soil-Foundation-Structure Interaction: Example of Fully-Coupled Gravity Caisson Modeling

For fully-coupled soil-foundation-structure interaction analysis in the case of pile (or drilled shaft) foundations with a limited number of piles (e.g. drilled shaft or pipe pile foundations), the pile cap and individual piles are explicitly modeled in the global model. SSI effects are included through the use of soil springs (linear or nonlinear) along the length of the pile. The stiffness of the soil springs is determined through analysis using programs such as COM624P or similar commercially available programs.

The COM624P program was developed for use in the analysis of stresses and deflections of piles or drilled shafts under lateral loads. For a detailed discussion on use of the COM624P program to calculate soil spring parameters for the piles or drilled shafts, refer to the literature, including the following:

• Ensoft, Inc., COM624P – Laterally Loaded Pile Analysis Program for the Microcomputer, Version 2.0 (Ref. F13) • Report No. FHWA/RD-85/106, "Behavior of Piles and Pile Groups under Lateral Loading" (Ref. F14) • Report No. FHWA-IP-84-11, "Handbook on Design of Piles and Drilled Shafts Under Lateral Load" (Ref. F15)

7 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

The fully-coupled approach to soil-foundation-structure interaction analysis for pile or drilled shaft foundations is illustrated in Figures F8 and F9, below.

Figure F 8 Soil-Foundation-Structure Interaction: Fully-Coupled Modeling of Piles or Drilled Shafts

8 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

Figure F 9 Soil-Foundation-Structure Interaction: Example of Fully-Coupled Modeling of Drilled Shafts

F 4.0 Simplified Design Charts

Simplified design charts are available which can be useful for determination of foundation stiffnesses during conceptual and preliminary design phases. For preliminary design or sensitivity evaluations, subgrade reaction model can be used to determine foundation stiffness coefficients. It assumes that support springs along the pile are elastic, independent of the pile diameter and vary linearly with depth. This linear representation of the soil stiffness can generally yield reasonable results provided that the soil conditions are not highly variable and the expected lateral deflection of the pile is within a reasonable range (i.e., between 0.25 and 1.0 inch).

It should be noted that most of these tools are not suitable for soil-structure analysis of long span bridges during the final design stage. For the final design phase, the determination of the lateral stiffness of individual piles, for instance, should be performed using COM624P or similar programs. For a detailed discussion of the use of the COM624P program to determine lateral pile stiffness, see references F13, F14 and F15.

Martin and Lam (Ref. F2) have developed a set of linear pile-head stiffness charts for use in soil- pile analysis under lateral loading. These design charts are presented in Figures F12 through

9 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

F18 as a function of the bending stiffness of the pile EI and the coefficient of variation, f, of soil modulus Es with depth. Coefficients of variation of subgrade modulus with depth for sand and for clay are presented in Figures F10 and F11, respectively. These design charts provide stiffness values for various pile-head embedment and boundary conditions. The pile-cap connection shall be properly accounted for when using these charts.

Similar simplified design charts and detailed explanations on their use (with examples) can be found in the literature, including the following:

• MCEER/ATC-49, Recommended LRFD Guidelines for the Seismic Design of Highway Bridges (Ref. F3) • Parsons Brinckerhoff, Inc., Geotechnical Engineering Circular No. 3 (Ref. F12)

Figure F 10 Recommendations for Coefficient of Variation in Subgrade Modulus with Depth for Sand

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Figure F 11 Recommendations for Coefficient of Variation in Subgrade Modulus with Depth for Clay

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Figure F 12 Lateral Stiffness of Free-Headed Pile

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Figure F 13 Coefficient for Lateral Pile Head Stiffness (Fixed Head Pile Lateral Stiffness)

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Figure F 14 Coefficient for Pile Head Rotation

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Figure F 15 Coefficient for Cross-Coupling Term

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Figure F 16 Comparison of Fixed Pile Head Stiffness at Various Embedments

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Figure F 17 Comparison of the Rotational Stiffness Coefficient at Various Embedments

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Figure F 18 Comparison of the Cross-Coupling Stiffness Coefficient at Various Embedments

F 5.0 Pile / Pile Group Analysis

For final design or detailed evaluations, stiffness coefficients should be obtained by performing laterally loaded pile or pile group analyses, using non-linear spring (i.e., p-y and t-z curves) distributed along the pile length.

18 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

A general procedure for developing pile foundation stiffness is outlined below:

(1) Develop site subsurface conditions, including stratigraphy, layer geometry, and soil properties.

(2) Select or derive appropriate soil parameters specifically used for laterally loaded pile analysis, including, but not limited to, the total and effective unit weights, undrained shear strength, Cu, and the strain at 50% peak stress level, ε50, of cohesive soils, friction angle of granular soils, Ф, and the initial soil modulus, Ki. Guidelines for selecting the required soil parameters are provided by Hannigan, et al. (Ref. F4).

(3) Derive foundation stiffness for a single pile foundation:

• Compute horizontal (Kxx and Kyy), rocking (Krx and Kry) and coupled stiffness (Kx-ry and Ky-rx) versus deflection of the single pile foundation using the non-linear p-y spring method used in computer code COM624P or similar commercial programs. For a detailed discussion on the use of the COM624P program to determine pile deflections, see references F13, F4 and F15. • Compute vertical stiffness (Kzz) versus deflection using the non-linear t-z spring method or other suitable methods. Ignore torsional stiffness (Kt) of a single pile foundation.

(4) Derive foundation stiffness for a pile group foundation.

• Compute horizontal stiffness (Kxx,1 and Kyy,1) versus deflection for a single pile in the group using the non-linear p-y spring method. • Compute vertical stiffness (Kzz,1) versus deflection for a single pile in the group using the t-z spring method. • Determine the pile group reduction factors (i.e., p-multiplier) for horizontal stiffness using the recommended values presented in Table F-1, below. • Calculate the horizontal (Kxx and Kyy) and vertical stiffness (Kzz) of the pile group using the non-linear p-y and t-z spring methods. The p-multipliers derived above should be applied to the p-y spring in the analysis. • Calculate the rocking (Krx and Kry) stiffness of the pile group using the vertical stiffness of single piles. Calculate the torsional stiffness (Kt) using the horizontal stiffness of single piles. • Calculate the lateral stiffness of pile cap using limit equilibrium method described in Article 15.4.5 of these Guidelines. • Combine the pile group and pile cap stiffness to obtain the total foundations stiffness versus deflection.

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Table F 1 “p-Multiplier” Values for Laterally Loaded Pile Group Analysis (Ref. F5 and F6)

D=pile diameter

It should be noted that the p-multiplier presented in Table F-1 can be used for small to moderate pile group foundations. For very large pile groups, such as those that might be used for the main span foundations for long span bridges, the group effect may not be properly represented by the p-multipliers shown in the Table F-1. Due to the large number of piles (hundreds) used in these foundations, the soil-pile interaction may be changed into a pile-reinforced mass system. To account for this potentially very important effect, analyses by continuum finite element method should be considered.

F 6.0 Equivalent Stiffness of Shallow Foundations (Approximate Methods)

As discussed in the previous section, for de-coupled SSI analysis, the foundation stiffness is expressed by a 6x6 stiffness matrix so that it can be easily incorporated into the dynamic analyses of the structure.

While the final design and analysis should be based on a detailed SSI evaluation, these procedures may be useful for conceptual/preliminary analysis and design of some shallow foundations for long span bridges.

The FHWA publications (Ref. F7) and (Ref. F8) and Lam, Martin and Imbsen, 1991 (Ref. F9) recommended the following procedure in deriving stiffness coefficients for shallow foundations. This procedure assumes that the footing is rigid and founded on a semi-infinite elastic half space.

(1) The stiffness matrix, K, of a non-circular shaped and/or embedded footing can be evaluated by the following general equation:

K = αβKECF

where KECF is the stiffness matrix of an equivalent circular surface footing, α is the foundation shape correction factor, and β is the foundation embedment factor. (2) Calculate the equivalent radius of the rectangular footing for the various modes of displacement as illustrated in Figure F19.

(3) Determine the shape factors, αx, αy, αz, αrx, αry, αt, for the various modes of displacement according to Figure F20. As shown in this figure, the shape factors are a function of the ratio of the length and width of the footing (L/B). The shape factors modify the results calculated for a circular footing to represent the stiffnesses of a rectangular footing with the same equivalent radius.

(4) Determine the embedment factors, βx, βy, βz, βrx, βry, βt using data presented in Figure F21. The embedment factors account for the increased stiffnesses due to footing

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embedment. For practical purposes the embedment, D, can be assumed to be equal to the thickness (or height) of the footing (Ref. F7). (5) Derive the stiffness coefficient values for the various modes of displacement defined in Step (1) above, as follows:

where G and γ are the dynamic shear modulus and Poisson’s ratio of the elastic half space. In practice, the average G and γ values of the soils within the depth of 2B (twice the width of the footing) below the footing are used in the calculations.

Similar procedures for approximating the stiffness of shallow foundations based on a rigid plate on semi-infinite homogenous elastic half-space can be found in MCEER/ATC-49 (Ref. F10).

It should be noted that shear modulus is a function of the earthquake-induced shear strain in the soils. To properly account for the stiffness degradation effect, shear modulus values should be derived from the free-field site response analyses.

The format of the damping matrix is the same as the format of the stiffness matrix. The damping ratio for a depends upon the mass (or inertial) ratio of the footing. Table F-2 lists the mass ratios, damping coefficients and damping ratios for the various degrees of freedom of a footing.

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Figure F 19 Procedure for Calculating Equivalent Radius of a Rectangular Footing

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Figure F 20 Shape Factor for Rectangular Footings

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Figure F 21 Embedment Factor for Rectangular Footings

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Table F 2 Equivalent Damping Ratios for Rigid Circular Footings

Notes: m = mass of the foundation c = damping coefficient (cz, cx, cψ, cθ) I = moment of inertia of the foundation ρ = mass density of foundation soil ro = equivalent radius (Rx, Rz, Rψ) 2B = width of the foundation (along axis of rotation for rocking) 2L = length of the foundation (in the plane of rotation for rocking) G = shear modulus of the soil v = Poisson’s ratio of the soil D = damping ratio (Dz, Dx, Dψ, Dθ)

25 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

F 7.0 Analysis of Interaction Between Component Force Effects (Axial, Shear and Bending)

It is recommended that concurrent or corresponding forces, moments and shears be used in the evaluation and design of components that are subject to simultaneous interaction between different force/moment components.

An example of a component that is subjected to the simultaneous action of multiple force components is a reinforced concrete column which is subjected to axial force and biaxial moments. For the design of “typical” highway bridges, a common (and conservative) approach is to combine the maximum/minimum axial force Fx with the maximum/minimum moments My and Mz. However, often times the maximum moment My is not concurrent with the maximum moment Mz or with the maximum//minimum axial force Fx. Superposition of non-concurrent force components Fx(max./min.), My(max.) and Mz(max.) may be overly conservative for long span bridges and may adversely affect the design.

Therefore, it is recommended that the analysis track concurrent (or corresponding) force and moment components as follows:

Some commercially available analysis programs do this internally. The Designer may also develop post-processing tools (spreadsheets or programs) to aid tracking these numbers. 26 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

REFERENCES

F1. Parsons Brinckerhoff Quade & Douglas, Inc./Imbsen Consulting Engineer, “Seismic Investigation of the Bronx-Whitestone Bridge, Task 7 Report – Development of Bridge Model and Calculation of Vibration Characteristics”, February 1996.

F2. Martin, G. R. and Lam, I. P., Seismic Design of Pile Foundations: Structural and Geotechnical Issues, State-of-the-Art (SOA4), Proceedings 3rd International Conference on Recent Advances in Geotechnical Earthquake Engineering and Soil Dynamics, St. Louis, Missouri, Vol. 3, 1995.

F3. MCEER/ATC-49, Recommended LRFD Guidelines for the Seismic Design of Highway Bridges, Applied Technology Council and the Multidisciplinary Center for Earthquake Engineering Research, 2003.

F4. Hannigan, P. J., Goble, G. G., Thendean, G., Likins, G. E. and Raushe, F, Design and Construction of Driven Pile Foundations, Workshop Manual, NHI Course Nos. 13221 and 13222, Publication Report No. FHWA HI-97-013, 1997.

F5. Brown, D. and Bollmann, H. T., Pile Group Design for Lateral Loading Using COM624P, Proceedings of The Design of Bridges for Extreme Events, Atlanta, Georgia, 1996.

F6. Hannigan, P. J., Goble, G. G., Thendean, G., Likins, G. E. and Raushe, F, Design and Construction of Driven Pile Foundations, Workshop Manual, NHI Course Nos. 13221 and 13222, Publication Report No. FHWA HI-97-013, 1997.

F7. Lam, I. P. and Martin, G. R., Seismic Design of Highway Bridge Foundations – Vol. II, Design Procedures and Guidelines, Report No. FHWA-RD-86-102, 1986.

F8. Parsons Brinckerhoff Quade and Douglas, Inc. and GeoSyntec, Inc., Geotechnical Earthquake Engineering, Reference Manual, Training Course In Geotechnical and Foundation Engineering, NHI Course No. 13239 – Module 9, Publication No. FHWA HI- 99-012, 1998.

F9. Lam, I.P., Martin G.R. and Imbsen, Roy, Modeling Bridge Foundations for Seismic Design and Retrofitting, submitted to the Third Bridge Engineering Conference at Denver, Colorado, March 10-13, 1991.

F10. MCEER/ATC-49, Recommended LRFD Guidelines for the Seismic Design of Highway Bridges, Applied Technology Council and the Multidisciplinary Center for Earthquake Engineering Research, 2003.

F11. Ho, T., Donikian, R., Ingham, T. Seim, C. and Pan, A., “Seismic Retrofitting Guidelines for Complex Steel Truss Highway Bridges”, Report MCEER-06-SP05, August 1, 2006.

F12. Parsons Brinckerhoff, Inc., Geotechnical Engineering Circular No. 3, Reference Manual, NHI Course No. 130094 – LRFD Seismic Analysis and Design of Transportation Geotechnical Features and Structural Foundations, Publication No. FHWA-NHI-11-032, Rev. 1, August 2011.

27 FHWA – Design Guidelines for Arch and Cable-Supported Signature Bridges Appendix F

F13. Ensoft, Inc., COM624P – Laterally Loaded Pile Analysis Program for the Microcomputer, Version 2.0, Report No. FHWA-SA-91-048, June 1993.

F14. Federal Highway Administration, "Behavior of Piles and Pile Groups under Lateral Loading", Report No. FHWA/RD-85/106, March 1986.

F15. Federal Highway Administration, "Handbook on Design of Piles and Drilled Shafts Under Lateral Load", Report No. FHWA-IP-84-11, July 1984.

F16. Brown, Dan A., Turner, John P. and Castelli, Raymond J., Geotechnical Circular No. 10 – Drilled Shafts: Construction Procedures and LRFD Design Methods, NHI Course No. 132014, Publication No. FHWA-NHI-10-016, May 2010. F17. Tseng, W-S., and Penzien, J., “Soil-Foundation-Structure Interaction”, Chapter 42, Bridge Engineering Handbook – Edited by W-F. Chen and L. Duan, CRC Press, 1999.

F18. Clough, R.W. and Penzien, J., Dynamics of Structures, Second Edition, McGraw-Hill Inc., 1993.

F19. Lam, I.P. and Law, H., “Soil Structure Interaction of Bridges for Seismic Analysis”, Technical Report MCEER-00-0008, Multidisciplinary Center for Earthquake Engineering Research, University at Buffalo, Buffalo, NY, September 25, 2000.

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