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Gibraltar Point Erosion Control Final Design

August 21, 2015 11503.101

Innovation, Excellence & Service www.baird.com

Gibraltar Point Erosion Control Final Design

Prepared for

Toronto and Region Conservation Authority

City of Toronto

Prepared by

W.F. Baird & Associates Coastal Engineers Ltd.

For further information please contact Mark Kolberg, P.Eng. at (905) 845‐5385

11503.100

Revision Date Status Comments Reviewed by Approved by 0 21 Aug 2015 Draft For review MK/MD

This report was prepared by W.F. Baird & Associates Coastal Engineers Ltd. for Toronto Region Conservation Authority and City of Toronto. The material in it reflects the judgment of Baird & Associates in light of the information available to them at the time of preparation. Any use which a Third Party makes of this report, or any reliance on decisions to be made based on it, are the responsibility of such Third Parties. Baird & Associates accepts no responsibility for damages, if any, suffered by any Third Party as a result of decisions made or actions based on this report.

Innovation, Excellence & Service DRAFT Baird & Associates

TABLE OF CONTENTS

1.0 INTRODUCTION ...... 1 1.1 Background ...... 1 1.2 Scope of Work for Final Design ...... 3

2.0 COASTAL CONDITIONS ...... 4 2.1 Site Visit ...... 4 2.2 Bathymetry ...... 7 2.3 Lakebed Substrate ...... 7 2.3.1 Available Geotechnical Data ...... 7 2.3.1.1 Bedrock ...... 7 2.3.1.2 Existing Borehole Data ...... 7 2.3.2 Surficial Sediment Samples ...... 11 2.3.3 Lakebed Video ...... 12 2.4 Water Levels ...... 14 2.4.1 Monthly Mean Lake Levels ...... 14 2.4.2 Short-Term Lake Levels ...... 15 2.4.3 Water Level Extreme Value Analysis ...... 17 2.4.4 Lake Level Regulation ...... 18 2.4.5 Climate Change Impacts on Lake Levels and Storm Surge ...... 18 2.5 Wave Climate ...... 20 2.5.1 Offshore Wave Climate ...... 20 2.5.2 Nearshore Waves and Extreme Value Analysis ...... 23 2.6 Ice ...... 23 2.7 Shoreline Comparisons ...... 24 2.7.1 Historical Context ...... 24 2.7.2 Present Littoral Cell and Descriptive Model of Sediment Transport ...... 31 2.7.3 Contemporary Shoreline Evolution ...... 32 2.8 Bathymetry Comparisons ...... 37 2.9 Lakebed Lowering Associated with Shoreline Erosion ...... 44 2.10 Numerical Analysis of Existing Coastal Conditions ...... 47 2.10.1 Model Selection ...... 47 2.10.2 HYDROSED Modeling of Existing Conditions ...... 50 2.10.3 Long-term Analysis of HYDROSED Results ...... 54

Gibraltar Point Erosion Control Table of Contents Final Design 11503.101 DRAFT Baird & Associates

3.0 DETERMINATION OF PREFERRED OPTION FOR FINAL DESIGN ...... 57 3.1 Numerical Analysis of Structural Design Alternatives ...... 57 3.1.1 Submerged Sill ...... 57 3.1.2 Low-Crested Structures (LCS) ...... 60 3.1.3 Emerged Breakwaters ...... 62 3.1.4 Groynes ...... 65 3.2 Analysis of Selected Final Design Alternatives ...... 67 3.2.1 Alternative #1 – Stand-Alone Sand Management ...... 68 3.2.2 Alternative #2 – Offshore Breakwater, Groyne and Focused Sand Management ...... 72 3.2.3 Evaluation of Selected Alternatives ...... 77 3.3 Conclusions and Preferred Alternative for Final Design ...... 78

4.0 FINAL DESIGN ...... 80 4.1 Offshore Breakwater and Shore-Connected Groyne ...... 80 4.1.1 Summary ...... 80 4.1.2 Rubble Mound Design Guidelines ...... 84 4.1.3 Design Life ...... 84 4.1.4 Damage Levels ...... 84 4.1.5 Acceptable Level of Risk of Damage ...... 86 4.1.6 Return Period of Design Event ...... 88 4.1.7 Design Water Levels and Wave Conditions ...... 89 4.1.8 Geotechnical ...... 90 4.1.9 Armour, Filter, and Core Design ...... 90 4.1.9.1 Armour Stability ...... 91 4.1.9.2 Armour Stone Placement ...... 91 4.1.9.3 Armour Stone Quality ...... 92 4.1.9.4 Filter/Underlayer and Core ...... 92 4.1.10 Crest Height and Crest and Rear Slope Stability ...... 92 4.1.11 Bedding Layer (Filter) ...... 93 4.1.12 Toe Berm ...... 93 4.1.13 Scour Apron ...... 94 4.2 Navigation Aids ...... 94 4.3 Aquatic Habitat Enhancements ...... 95 4.4 Additional Sand Placement ...... 97 4.5 Estimated Quantities ...... 100

Gibraltar Point Erosion Control Table of Contents Final Design 11503.101 DRAFT Baird & Associates

5.0 FUTURE BREAKWATER MONITORING AND MAINTENANCE ...... 102 5.1 Breakwater Monitoring ...... 102 5.1.1 Introduction ...... 102 5.1.2 Frequency of Monitoring ...... 102 5.1.3 Routine Visual Inspection ...... 102 5.1.4 Detailed Inspections and Surveys ...... 103 5.1.5 Safety Considerations ...... 103 5.1.6 Profile Survey Procedures ...... 104 5.1.6.1 Conventional Survey ...... 104 5.1.6.2 Multibeam and Other Technologies ...... 104 5.1.6.3 Diving Surveys ...... 104 5.1.7 Monitoring Armour Stone Degradation ...... 104 5.2 Maintenance and Repairs ...... 105

6.0 REFERENCES ...... 107

APPENDIX A – SITE VISIT REPORT (DECEMBER, 2009) APPENDIX B – LAKEBED VIDEO (TRCA, 2014) APPENDIX C – NEARSHORE WAVES EXTREME VALUE ANALYSIS APPENDIX D – EVALUATION OF POTENTIAL OFFSHORE SAND SOURCES APPENDIX E – WESTERN BEACHES BREAKWATER PHYSICAL MODEL

Gibraltar Point Erosion Control Table of Contents Final Design 11503.101 DRAFT Baird & Associates

LIST OF TABLES

Table 2.1 Maximum and Minimum Hourly Water Levels at Toronto (1962-2001) ...... 16 Table 2.2 High Water Levels as a Function of Return Period ...... 17 Table 2.3 Low Water Levels as a Function of Return Period ...... 17 Table 2.4 Summary of Wave Heights for Given Return Periods ...... 23 Table 2.5 Calculated Average Annual Shoreline Recession Rates Along Transect A ...... 36 Table 2.6 Historic Lakebed Lowering Rates at -3.5 m CD ...... 44 Table 2.7 Modelled Wave Conditions (Height, Period, Direction) at Domain Boundary ...... 54 Table 3.1 Net Present Value Cost of Alternative #1, Stand-Alone Sand Management ...... 71 Table 3.2 Net Present Value Cost of Alternative #2, Breakwater, Groyne and Focused Sand Management ...... 75 Table 3.3 Summary Evaluation of Selected Alternatives ...... 77 Table 4.1 Damage Level, D for Two-Layer Rock Armour on Slope 1:2 to 1:3 (USACE, 2006) ...... 85 Table 4.2 Damage Level S for Two-Layer Rock Front Breakwater Slope (1:1.5) (Vidal et al., 1992) ...... 86 Table 4.3 Risk, Project Design Life and Design Event Return Period for Rock Armour ...... 88 Table 4.4 Design Water Levels and Wave Heights ...... 89 Table 4.5 Offshore Breakwater Crest Height Considerations ...... 93 Table 4.6 Estimated Quantities of Final Design ...... 101 Table 5.1 Breakwater Monitoring Frequency ...... 102

LIST OF FIGURES

Figure 1.1 Oblique aerial view of Gibraltar Point (TRCA photo, August 2014) ...... 1 Figure 1.2 Preferred Concept (“Concept 6”) from Environmental Study Report (TRCA, 2008) ...... 2 Figure 2.1 Oblique aerial view of Gibraltar Point to Hanlan’s Beach (TRCA photo, November, 2008) ...... 4 Figure 2.2 Site visit, December 2009 ...... 5 Figure 2.3 Oblique aerial view of revetment protection at washroom building at Gibraltar Point ...... 6 (TRCA photo, 2013) ...... 6 Figure 2.4 Existing shoreline protection works at Gibraltar Point (TRCA, 2008) ...... 6 Figure 2.5 Bathymetry at Gibraltar Point (TRCA 2009 survey) ...... 8 Figure 2.6 Bedrock contours (metres below CD) (from Lewis and Sly, 1971) ...... 9 Figure 2.7 Surficial grain size and prior borehole locations ...... 10 Figure 2.8 Bottom sediment types, Humber Bay (from Coakley and Poulton, 1991)...... 12 Figure 2.9 Lakebed video transects (adapted from TRCA mapping, 2014) ...... 13 Figure 2.10 Sandy lakebed at Transect 1 (from TRCA video, August 2014) ...... 13 Figure 2.11 Lake Ontario monthly mean lake levels from 1918 to 2010 ...... 14 Figure 2.12 Annual water level variation (hourly time series), Toronto, Lake Ontario (1962-2012) . 15 Figure 2.13 Exceedence statistics for hourly water levels (all months) at Toronto (1962-2001)...... 16 Figure 2.14 Projected Lake Ontario water levels due to climate change (ref. http://www.glerl.noaa.gov/data/now/wlevels/dbd/ accessed June 21, 2014) ...... 19 Figure 2.15 Offshore wave height rose (Point 2416) ...... 21 Figure 2.16 Offshore wave height scatter plot (point rose) (Point 2416)...... 21 Figure 2.17 Offshore wave height distribution by direction ...... 22 Figure 2.18 Distribution of wave height and wave energy by direction ...... 22 Figure 2.19 Location of gabion protection at Gibraltar Point (THC, 1983) ...... 26

Gibraltar Point Erosion Control Table of Contents Final Design 11503.101 DRAFT Baird & Associates

Figure 2.20 Location of gabion basket protection at Gibraltar Point (1978) ...... 27 Figure 2.21 Concrete blocks placed near Hanlan’s Point (photo December 2009) ...... 27 Figure 2.22 Shoreline changes to west side of Toronto Islands ...... 29 Figure 2.23 Exposure of steel sheet pile above lake bed at intake pipes in June 1993 (Baird, 1994a) ...... 30 Figure 2.24 Downdrift (west) side of groyne at east intake (December 2009) ...... 30 Figure 2.25 Dredged trench for Deep Lake Water Cooling Pipes east of Gibraltar Point (August 2003) ...... 31 Figure 2.26 Baird (1994b) descriptive model of sediment transport at Gibraltar Point ...... 32 Figure 2.27 Historic shoreline comparison, 1913 to 2009 ...... 33 Figure 2.28 Comparison between 1980 and 2009 shorelines with estimated erosion and accretion areas ...... 34 Figure 2.29 Comparison between 1999 and 2009 shorelines with estimated erosion and accretion areas ...... 34 Figure 2.30 Comparison of 1982 airphoto (top panel) and 1990 airphoto (bottom panel) on 1980 base mapping (blue line) (from mapping by TRCA) ...... 35 Figure 2.31 Historic shorelines and Transect A in the area between Gibraltar Point and Hanlan’s Point ...... 36 Figure 2.32 Estimated future shorelines based on extrapolation of the average historic recession rate ...... 37 Figure 2.33 Hydrographic survey lines for 1981, 1993, 2005 and 2009 surveys and profile locations ...... 38 Figure 2.34 Portion of 1962 THC bathymetric survey plan ...... 38 Figure 2.35 Comparison between historic bathymetric surveys along Profile E1 ...... 40 Figure 2.36 Comparison between historic bathymetric surveys along Profile F1 ...... 40 Figure 2.37 Comparison between historic bathymetric surveys along Profile F2 ...... 41 Figure 2.38 Comparison between historic bathymetric surveys along Profile H2 ...... 41 Figure 2.39 Comparison between historic bathymetric surveys along Profile D1 ...... 42 Figure 2.40 Lakebed surface comparison of accretion and erosion between 1993 and 2009 surveys ...... 42 Figure 2.41 Distribution of accretion along Hanlan’s Beach between 1993 and 2009 ...... 43 Figure 2.42 Lakebed lowering at Profile E1 ...... 45 Figure 2.43 Lakebed lowering at Profile F1 ...... 46 Figure 2.44 Lakebed lowering at Profile F2 ...... 46 Figure 2.45 Comparison of wave height vectors from MIKE21 and HYDROSED for southwest storm ...... 48 Figure 2.46 Comparison of wave height vectors from MIKE21 and HYDROSED for east storm .... 48 Figure 2.47 Comparison of nearshore currents from MIKE21 and HYDROSED for southwest storm ...... 49 Figure 2.48 Comparison of nearshore currents from MIKE21 and HYDROSED for east storm ...... 49 Figure 2.49 HYDROSED model domains ...... 50 Figure 2.50 Modelled wave height and direction for a southwesterly storm ...... 51 Figure 2.51 Modelled nearshore currents for a southwesterly storm ...... 52 Figure 2.52 Modelled wave height and direction for an easterly storm ...... 53 Figure 2.53 Modelled nearshore currents for an easterly storm ...... 53 Figure 2.54 Modelled long-term average potential longshore sand transport (LST) rates ...... 55 Figure 2.55 Schematic summary of erosion problem at Gibraltar Point ...... 56 Figure 3.1 Comparison of nearshore currents when a 750 m submerged sill is in place (black arrows) with existing conditions (red arrows) ...... 58 Figure 3.2 Wave height and direction when a 33% sill-67% emerged breakwater is in place ...... 59

Gibraltar Point Erosion Control Table of Contents Final Design 11503.101 DRAFT Baird & Associates

Figure 3.3 Comparison of nearshore currents when a 33% sill-67% emerged breakwater is in place (black arrows) with existing conditions (red arrows) ...... 59 Figure 3.4 Wave height and direction when a pair of low crested structures are in place ...... 61 Figure 3.5 Comparison of nearshore currents when a pair of low crested structures are in place (black arrows) with existing conditions (red arrows) ...... 61 Figure 3.6 Beach profile change and wave setup for various LCS options ...... 62 Figure 3.7 Wave height and direction when a 900 m long breakwater is in place ...... 63 Figure 3.8 Wave height and direction under SW waves with the 550 m breakwater and groyne option ...... 64 Figure 3.9 Comparison of nearshore currents under SW waves when a 550 m breakwater with a groyne are in place (black arrows) with existing conditions (red arrows) ...... 64 Figure 3.10 Comparison of nearshore currents under E waves when a 550 m breakwater with a groyne are in place (black arrows) with existing conditions (red arrows) ...... 65 Figure 3.11 Wave height and direction under SW waves when a two groynes are in place...... 66 Figure 3.12 Comparison of nearshore currents under SW waves when two groynes are in place (black arrows) with existing conditions (red arrows) ...... 66 Figure 3.13 Estimated future shorelines with two groynes if the shoreline protection at Gibraltar Point is maintained (green line) or removed (red line) ...... 67 Figure 3.14 Sand placement area for Alternative #1, Stand-Alone Sand Management ...... 68 Figure 3.15 Variation of annual net LST rate at Gibraltar Point ...... 69 Figure 3.16 Examples of beach nourishment activities ...... 70 Figure 3.17 Alternative #2, 550 m long offshore breakwater and 100 m long shore-connected groyne ...... 72 Figure 3.18 Groynes at Eastern Beaches ...... 73 Figure 3.19 Annual variation of sand volume loss from Hanlan’s Beach ...... 74 Figure 3.20 Sand placement area for focused sand management as part of Alternative #2 ...... 76 Figure 3.21 Offshore breakwater and groyne with focused sand management (Alternative #2) option selected for final design ...... 79 Figure 4.1 Final design plan view ...... 81 Figure 4.2 Final design cross-sections of breakwater ...... 82 Figure 4.3 Final design cross-sections of groyne ...... 83 Figure 4.4 Schematic illustration of eroded cross-sectional area, Ae ...... 85 Figure 4.5 Graphical relationship of design event return period, risk and design life...... 88 Figure 4.6 Typical navigation aid ...... 95 Figure 4.7 Aquatic habitat enhancements ...... 96 Figure 4.8 Additional sand placement plan ...... 98 Figure 4.9 Predicted movement of sand placed in lee of breakwater following one-time placement ...... 99 Figure 4.10 Additional beach sand showing projected final beach configuration following multiple placements ...... 100 Figure 5.1 Typical marine plant required for repair work (top -barge mounted crane; bottom – backhoe on barge and tugs) ...... 106

Gibraltar Point Erosion Control Table of Contents Final Design 11503.101 DRAFT Baird & Associates

1.0 INTRODUCTION

1.1 Background

Gibraltar Point is located at the southwesterly tip of the Toronto Islands, on Lake Ontario. Figure 1.1 shows Gibraltar Point and the adjacent Hanlan’s Point and Hanlan’s Beach. The Toronto and Region Conservation Authority (TRCA) completed an Environmental Study Report (ESR), in accordance with Conservation Ontario’s Class Environmental Assessment for Remedial Flood and Erosion Control Projects to develop a long‐term solution to address the shoreline erosion around Gibraltar Point (TRCA, 2008). The background and history of the erosion, the purpose of the proposed remedial action (the “undertaking”) and the rationale for the undertaking are detailed in the ESR (TRCA, 2008).

Figure 1.1 Oblique aerial view of Gibraltar Point (TRCA photo, August 2014)

Based on the outcome of the Class EA process for the project, the proposed remedial action was a sand management plan, which recognized that some form of offshore structure would likely be required to make the sand management plan technically and economically feasible. The decision‐ making approach used in selecting the preferred remedial action is provided in the ESR (TRCA, 2008).

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Figure 1.2 presents the sand management plan concept (i.e., “Concept 6”) identified in the ESR (TRCA, 2008) as the preferred concept. The ESR did not provide details of the preferred concept but concluded that the final design would determine how to balance the preservation of the supply and transport of sand to Hanlan’s Beach and the requirement for some structural offshore protection of Gibraltar Point. The ESR anticipated that the sand supply would be produced from a backpassing operation where sand would initially be removed from the depositional area offshore of Hanlan’s Point Beach and the Western Gap and placed updrift of Gibraltar Point in the vicinity of Centre Island Beach. The ESR noted the possibility that sand could be supplied from other sources, such as dredging at Ashbridge’s Bay and the Eastern Channel but details were not provided.

Figure 1.2 Preferred Concept (“Concept 6”) from Environmental Study Report (TRCA, 2008)

The ESR (TRCA, 2008) determined that an offshore protection structure would likely be required to reduce the volume and frequency of sand required to be placed updrift of Gibraltar Point as part of the sand management plan (i.e., the more protection constructed, the less sand placement required). Five possible concept options for implementing the preferred alternative were identified in the ESR. Four of the options consisted of offshore protection structures with different lengths of emergent (surface piercing) segmented breakwaters connected with submerged breakwaters (sills), in conjunction with different levels of sand management. The fourth option was an emergent breakwater with little to no sand management. The ESR considered that an overall length of 930 m for the offshore structure was “a reasonable approximation of the longest length likely to be required in order to produce conservative cost estimates. The fifth option was a sand management plan without offshore protection. Essentially the “preferred concept” ranged from “all breakwaters and no sand” to “no breakwaters and all sand”.

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1.2 Scope of Work for Final Design

The purpose of this study is to complete the coastal engineering analysis and final design for the Gibraltar Point Erosion Control Project to prevent further loss of recreational beaches, parkland and unique habitats and protect existing infrastructure. The objectives are to control erosion and maintain a dynamic beach shore through a long‐term, sustainable sand management approach with the minimum structural intervention that is economically feasible. The objectives are to be achieved through a detailed assessment of the five “preferred concept” alternatives developed through the Class Environmental Assessment (TRCA, 2008), identification of the final preferred alternative and preparation of the final design and construction drawings and specifications.

The detail design study approach is summarized as follows:

1. Quantify the key erosion, deposition and sediment transport processes at Gibraltar Point, including the localized effects of the intakes. 2. Evaluate the five ESR‐generated alternatives and determine the preferred sand management approach that minimizes, to the greatest extent that is feasible, structural intervention and provides the most practical, sustainable, long‐term solution to controlling the erosion at Gibraltar Point. 3. Prepare the final design, construction plans and specifications and monitoring requirements of the selected sand management approach including the dredging and placement of borrow sand and coastal structures.

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2.0 COASTAL CONDITIONS

The existing coastal conditions at the site, including lakebed substrate and surficial sediments, bathymetry, water levels, waves, ice are described in this section. This data is used to quantify the key erosion, deposition and sediment transport processes at Gibraltar Point and the design loads to support the final design.

2.1 Site Visit

A detailed site visit was undertaken by two coastal engineers from Baird at the outset of the project in December 2009. The purpose of the site visit was to observe the shoreline conditions at ground level in the vicinity of the project (see Figure 2.1). The site visit is detailed in Appendix A. Figure 2.2 shows the study area visited and provides a key plan for the place names used in this report. Figure 2.3 provides a close‐up oblique aerial view of the revetment structure that was constructed in 2004 at Gibraltar Point to protect the washroom facilities. Figure 2.4 shows a plan of the existing conditions around the washroom area.

Gibraltar Point

Hanlan’s Beach Hanlan’s Point

Figure 2.1 Oblique aerial view of Gibraltar Point to Hanlan’s Beach (TRCA photo, November, 2008)

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Figure 2.2 Site visit, December 2009

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Figure 2.3 Oblique aerial view of revetment protection at washroom building at Gibraltar Point (TRCA photo, 2013)

Figure 2.4 Existing shoreline protection works at Gibraltar Point (TRCA, 2008)

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2.2 Bathymetry

Bathymetry data was collected in 2009 by TRCA. The 2009 bathymetry is shown in Figure 2.5 and was used for the detailed analysis and design. The contours are relatively uniform around the point, except in the vicinity of the Deep Lake Water Cooling intake pipes to the east of the proposed protection works. The TRCA undertook another bathymetric survey in 2014. No notable changes that would affect the detail analysis and design were noted between the 2009 and 2014 conditions.

2.3 Lakebed Substrate

The lakebed is expected to be fine sand based on the nature of the formation of the Toronto Islands, the existing geotechnical data, the collected surficial sediment samples and an underwater video of the lakebed. The formation of the Islands is described in the ESR (TRCA, 2008) and in Section 2.7.1. The available geotechnical data, the surficial sediments and the underwater video are presented in this section.

2.3.1 Available Geotechnical Data

A geotechnical investigation was not part of the scope of the detail design work. Existing reports and information were assembled and reviewed to provide a basis for the preparation of the final design. It is recommended that boreholes be drilled as part of the construction program to confirm the substrate conditions.

2.3.1.1 Bedrock

Available geotechnical information indicates that bedrock is well below the lakebed in the project area and it is not expected to influence the project design. Bedrock topographic contours from Lewis and Sly (1971) are shown in Figure 2.6 (with depths given in metres below chart datum (CD), 74.2 m IGLD85). In the vicinity of the proposed structure, bedrock is greater than 40 m below CD.

2.3.1.2 Existing Borehole Data

Geotechnical boreholes were not part of the present scope of work due to the high cost of undertaking a borehole drilling program offshore. It was determined that sufficient information was available to advance the detail design. It is recommended that boreholes be drilled as part of the construction program, when the offshore barge is mobilized, to confirm the substrate conditions prior to placement of the structures.

Toronto Port Authority (formerly Toronto Harbour Commission, THC) records provided four boreholes in the nearshore in the vicinity of the proposed work. Referenced as “1962 Geocon”, the THC location plan of the four boreholes was superimposed over the base mapping for the present project. The accuracy of the initial mapping is uncertain but considered sufficient as collaborative evidence. The location of the four 1962 Geocon boreholes are shown in Figure 2.7.

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Figure 2.5 Bathymetry at Gibraltar Point (TRCA 2009 survey) (proposed breakwater and groyne shown in grey)

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Figure 2.6 Bedrock contours (metres below CD) (from Lewis and Sly, 1971)

The THC plan reports the 1962 Geocon boreholes consisted of an upper layer, several metres thick, of compact to dense grey fine sand over a lower layer of sandy silt. No details of the physical properties of the sand and silt materials were provided. The borehole profiles are presented in Section 2.8. From the borehole data, the elevation of the sand‐silt interface appears to trend upwards in the northwesterly direction. At borehole number 3 (BH3), roughly at the location of the proposed breakwater, the elevation of the interface between the upper fine sand layer and the underlying silt layer was reported to be approximately ‐10 m CD. From Figure 2.7, it can be seen that there is no information at westerly half of proposed breakwater, although it can be expected that the lakebed material is fairly homogeneous.

The locations of boreholes undertaken in 2000 by Thurber (pers. comm.) for the Deep Lake Water Cooling project are shown in Figure 2.7. This data was provided to Baird for information purposes only and is not intended for final design. Borehole (BH) 9 at the shore, about 500 m to the east of Gibraltar Point, consisted of 1.3 m of fill onshore, over “SAND, fine to medium grained, trace gravel, compact” to ‐8.4 m CD, then “SILT, sandy, occasional grey silt lenses, compact” to ‐60 m CD, over “CLAY, silty, trace to some sand, hard” to shale bedrock at ‐67 m CD. BH7, located offshore east of Gibraltar Point at ‐8.6 m CD, consisted of layer of “SAND, fine grained, trace silt, loose, grey” to depth ‐14.7 m CD, over “SILT, sandy, loose, grey, occasional seams of organics and clay” to the end of the borehole at ‐19.8 m CD. Sieve testing of a sand sample from BH7 revealed a D50 of 0.16 mm.

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Figure 2.7 Surficial grain size and prior borehole locations

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Interpolating between Thurber BH9 and BH7 suggests the elevation of the sand‐silt interface may be approximately ‐10 m CD at the ‐3.5 m CD lakebed contour. This is consistent with the elevation of the sand‐silt interface in 1962 Geocon BH3 (see Section 2.8, Profile F1). The bedrock elevation in BH9 is consistent with the bedrock contour elevation in Figure 2.6.

During excavation through the nearshore for trenches for the Deep Lake Water Cooling intake pipes just to the east of the present project site, McNally Construction (J. Starbuck, pers. comm.) encountered “dense, compacted sands ‐ no tills or boulders were encountered”. Reportedly, McNally used both a hydraulic suction dredge and a clam dredge with a heavy digging bucket.

2.3.2 Surficial Sediment Samples

Surficial sediment samples from the lakebed were collected by TRCA in November 2009. TRCA provided the results of the dry sieve analysis to Baird. The results of sediment sampling undertaken by TRCA in May 2002 in the area of the Western Channel near the east side of Ontario Place were also provided to Baird. Figure 2.7 shows the locations of the samples along with the D10, D50 and D90 grain sizes.

The nearshore surficial sediment samples at Gibraltar Point are predominantly fine to medium sand based on the D50. This is consistent with previous reports from the borehole data (Section 2.3.1.2). Along Hanlan’s Beach, the sediment samples are fine sand with D50 values typically 0.2 mm and D10 values of about 0.13 mm. The grain size distribution of the beach sand at Hanlan’s Beach is an important consideration when assessing sources of sand for the sand management options as any imported sand should be compatible (e.g., grain size of imported sand should be the same or larger to reduce losses of sand to the offshore).

Figure 2.8 shows bottom sediment types in Humber Bay and the vicinity of Gibraltar Point as characterized by Coakley and Poulton (1991). Nearshore at Gibraltar Point the sediment is described as sand and gravel. At Humber Bay, modern sediments in the offshore are comprised of muddy sands and sandy muds. Areas in Humber Bay where mud deposition exceeds 1 m in thickness were found to occur primarily within an east‐west band around the 10‐15 m contour. Glacial deposits (glaciolacustrine sediments or till) primarily outcrop in the western and central areas of Humber Bay. Bedrock outcrops are found offshore of the north end of Hanlan’s Beach and to the east and west of Ontario Place. As discussed in Section 2.3.1.1, bedrock is not expected to influence the final design of the protection works at Gibraltar Point.

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Figure 2.8 Bottom sediment types, Humber Bay (from Coakley and Poulton, 1991)

2.3.3 Lakebed Video

TRCA undertook a video of the lakebed in the vicinity of the project in August 2014. The video transects are shown in Figure 2.9 along with location of the proposed breakwater and groyne. The video revealed that the lakebed surface is sandy along all the transects, particularly towards the location of the proposed breakwater. Figure 2.10 shows a typical image from the video of the sandy bottom. A limited amount of loose stone (boulders, armour stones, rip rap) was observed at number of locations. These materials are possibly the remains of earlier attempts at erosion protection. A section of pipe was observed. Observations from the lakebed video are summarized in Appendix B along with several more images from the video.

Figure 2.9 shows the locations of the former CKEY radio antenna towers based on a 1978 aerial photograph. The towers have been removed, but 2007 aerial imagery provided by TRCA shows the possible remains of the foundations.

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Figure 2.9 Lakebed video transects (adapted from TRCA mapping, 2014) (Location of former CKEY radio antenna towers shown)

Figure 2.10 Sandy lakebed at Transect 1 (from TRCA video, August 2014)

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2.4 Water Levels

Water levels are an important factor in the development of the final design of the crest elevation of the protection structures. The depth of water also influences the wave height, which in turn is a significant factor in establishing the stability of the armour protection of the structures.

Monthly mean water levels on Lake Ontario vary in the long term (i.e., years) and seasonally in response to climatic conditions over the Great Lakes drainage basin (principally precipitation and evaporation), as well as lake level regulation. Future mean levels may be affected by climate change. Water levels can also vary in the short term (i.e., hours) in response to storm surge events.

2.4.1 Monthly Mean Lake Levels

Figure 2.11 presents Lake Ontario monthly mean water levels from 1918 to 2010. Water levels are referenced to Lake Ontario’s chart datum (CD), which is 74.2 m International Great Lakes Datum (IGLD 1985) and is considered the elevation that the water level seldom falls below. Over this period, the maximum recorded monthly mean water level on Lake Ontario was 75.76 m IGLD’85 and the minimum was 73.74 m IGLD’85, a range of 2.0 m (http://www.waterlevels.gc.ca/c&a/back‐ arrieres_e.html; accessed 2014/07/07).

Figure 2.11 Lake Ontario monthly mean lake levels from 1918 to 2010

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Hourly lake level data for Toronto is available since 1962. A plot of the annual variations in lake levels from 1962 to 2012 is provided in Figure 2.12, which also highlights the years containing the maximum (1973) and minimum (1965) static lake levels since lake level regulation began in 1960. It can be seen in Figure 2.12 that water levels in 1973 persisted at a relatively high level over a period of almost three months through the summer. Figure 2.12 also highlights the long term monthly mean water level.

1986

1964

Figure 2.12 Annual water level variation (hourly time series), Toronto, Lake Ontario (1962‐2012) (red line is year with maximum level and green line is year with minimum level since 1962)

The typical seasonal variation in lake level (based on monthly mean averages) is approximately +0.5 m CD, with the average seasonal low occurring in December (+0.3 m CD) and the average seasonal high occurring in June (+0.8 m CD). Through the summer, from June to September, the long‐term average monthly mean water level varies from about +0.8 m CD to +0.5 m CD.

2.4.2 Short-Term Lake Levels

Figure 2.13 provides the frequency of occurrence and frequency of exceedence for hourly water levels throughout the year for Toronto.

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1962-2001 Toronto Hourly Water Levels (CHS gage 13320) 5.0% 100%

Occurrence 4.5% 90% Exceedance

4.0% 80%

3.5% 70%

3.0% 60%

2.5% 50%

2.0% 40% Frequency of Occurence (%) of Occurence Frequency Frequency of Exceedance (%) Frequency 1.5% 30%

1.0% 20%

0.5% 10%

0.0% 0% 73.62 73.73 73.84 73.95 74.06 74.17 74.28 74.39 74.50 74.61 74.72 74.82 74.93 75.04 75.15 75.26 75.37 75.48 75.59 75.70 Lake Level (m, IGLD '85)

Figure 2.13 Exceedence statistics for hourly water levels (all months) at Toronto (1962‐2001)

Table 2.1 summarizes the maximum and minimum hourly water levels at Toronto over the period of record of the hourly data (1962‐2001).

Table 2.1 Maximum and Minimum Hourly Water Levels at Toronto (1962‐2001) Water Level Water Level (IGLD 1985) Date

Maximum Hourly 75.81 m May 28, 1973

Minimum Hourly 73.62 m Feb. 4, 1965

Range 2.19 m ‐

The annual maximum monthly mean (“static”) water level was determined from the data. The hourly water level data were further analyzed in order to estimate extreme high water levels as a function of return period. Storm surge events were separated out from the hourly water level records and the peak surge events occurring over the period from 1962 to 2001 were identified. Storm surge is a short‐term (hours) increase in the level of the lake at a site. When strong winds continue to blow over the lake surface in one direction for a number of hours, an increase in the water level against the downwind shoreline is produced, referred to as “storm surge” or “wind setup”. When winds away from the site, storm setup occurs at the opposite end of the lake and storm set‐down occurs at the site.

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2.4.3 Water Level Extreme Value Analysis

A combined probability analysis was then performed on the monthly mean (static) and storm setup (surge) data (1962 – 2001) in order to estimate the combined water level (surge plus monthly mean) as a function of return period. Static levels and storm surge were approximated as independent events. The results are shown in Table 2.2 for the full year and summer boating season (May to October). The full year 100‐year combined water level is +1.8 m CD.

Table 2.2 High Water Levels as a Function of Return Period Period Water Level Return Period (years) 5 10 25 50 100 Static 75.34 75.47 75.62 75.73 75.84 (m IGLD85) Full Year Surge (m) 0.20 0.22 0.24 0.25 0.26 Combined 75.51 75.64 75.78 75.89 75.99 (m IGLD85) +1.31 m CD +1.44 m CD +1.58 m CD +1.69 m CD +1.79 m CD Static 75.34 75.46 75.59 75.68 75.76 Boating (m IGLD85) Season Surge (m) 0.15 0.17 0.19 0.20 0.21 (Summer) Combined 75.47 75.58 75.71 75.80 75.87 (m IGLD85) +1.27 m CD +1.38 m CD +1.51 m CD +1.60 m CD +1.67 m CD

A similar analysis was undertaken to estimate extreme low water levels (static, storm set‐down and combined) as a function of return period. The results are shown in Table 2.3 for the full year and summer boating season (May 1 to October 31). The full year combined low water level with a return period of 100 years is +1.8 m CD.

Table 2.3 Low Water Levels as a Function of Return Period

Period Water Level Return Period (years) 5 10 25 50 100 Static 74.34 74.24 74.09 73.97 73.85 (m IGLD85) Set‐down Full Year ‐0.23 ‐0.26 ‐0.30 ‐0.33 ‐0.36 (m) Combined 74.14 74.03 73.91 73.81 73.71 (m IGLD85) ‐0.06 m CD ‐0.17 m CD ‐0.29 m CD ‐0.39 m CD ‐0.49 m CD Static 74.52 74.47 74.41 74.38 74.35 (m IGLD85) +0.32 m CD +0.27 m CD +0.21 m CD +0.18 m CD +0.15 m CD Boating Set‐down Season ‐0.16 ‐0.18 ‐0.21 ‐0.22 ‐0.25 (m) (Summer) Combined 74.38 74.32 74.26 74.23 74.19 (m IGLD85)

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The sum of the 50‐year static low water level (‐ 0.2 m) and a 10‐year storm setup (surge) (0.22 m) is approximately +0.0 m CD.

2.4.4 Lake Level Regulation

The water levels of Lake Ontario have been regulated by the outflow of the Moses‐Saunders Power Dam located on the St. Lawrence River at Cornwall‐Massena since 1960. The operation plan for the dam attempts to balance the water needs for multiple stakeholders (e.g., riparian owners, natural habitat, shipping, hydroelectric power generation, and recreation) while keeping Lake Ontario water levels within a 1.22 m range, from 74.15 m to 75.37 m IGLD1985. Figure 2.11 shows the effect of regulation with a more compressed range of water levels in the post‐regulation period since 1960 than during the pre‐regulation period from 1918 to 1959.

The International Joint Commission developed an alternative regulation plan (“Plan 2014”) for Lake Ontario that would slightly increase the upper and lower limits of the present operating range. The governments of Canada and the United States are assessing the new regulation plan and the timing of implementation of the plan is not known.

2.4.5 Climate Change Impacts on Lake Levels and Storm Surge

There is considerable uncertainty regarding the impact of climate change on the water levels of the Great Lakes. Despite the uncertainty exhibited in the simulations, there is general consensus about probable reductions of lake levels as a result of climate change (Environment Canada, 2004).

Angel and Kunkel (2010) suggest that through to the period 2050 to 2060 the level of Lake Ontario may decline in the order 0.5 m, within a range of slightly less than 2 m. This estimated range is within the historic range of Lake Ontario levels over the last century (see Section 2.4.1). Figure 2.14 presents data on lake level projections due to climate change taken from a website sponsored by the Great Lakes Environmental Research Laboratory (GRERL), part of the National Oceanic and Atmospheric Administration (NOAA) and Great Lakes Restoration Initiative (GLRI) (Gronewold et al. 2013). The Angel and Kunkel (2010) projection is for the higher emission scenario (the range shown in Figure 2.14 is the bottom 25 percentile and the top 75 percentile). Hayhoe et al. (2010) projected a drop of about 0.5 m in the average Lake Ontario level by the end of the century under the higher emissions scenario. Under a lower emissions scenario (i.e., with less warming), little change in the average water level is projected.

For design purposes, it has been assumed that water levels over the next 50 year will be similar to the historical water levels. Lower water levels would result in lower wave heights at the structure, due to depth‐limited conditions. The structure armouring would be more stable and less overtopping would occur. With lower water levels, nearshore lowering would increase.

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Figure 2.14 Projected Lake Ontario water levels due to climate change (ref. http://www.glerl.noaa.gov/data/now/wlevels/dbd/ accessed June 21, 2014)

There also is concern that the frequency and intensity of severe storms will increase as a result of climate change (http://glisa.umich.edu/media/files/GLISA_climate_change_summary.pdf , accessed July 2, 2014). There is a high level of uncertainty associated with impacts of climate change on storm frequency and intensity. However, the uncertainty can be accommodated in detailed design by considering a range of plausible increases in storm surges and wave heights over the life of the project.

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2.5 Wave Climate

2.5.1 Offshore Wave Climate

Estimation of waves from winds (often called hindcasting) on the Great Lakes is typically done on a lake‐wide basis using modern, advanced numerical hindcast models. Baird’s hindcast model (Scott et al., 2004) has been used to generate wave statistics for Lake Ontario for the Wave Information Studies (WIS) database (USACE, 2010). WIS is a US Army Corps of Engineers (USACE) sponsored project that generates consistent, hourly, long‐term (20 plus years) wave climatologies along all US coastlines, including the Great Lakes and US island territories (USACE, 2010). The Lake Ontario WIS database was prepared by Baird as part of an International Joint Commission (IJC) Lake Ontario Wave Climate – St. Lawrence River Water Level Regulation Study (Baird, 2003). The hindcast uses the WAVAD model and is based on 40 years of wind and wave data covering the period from 1961 to 2000. The model results were validated against two multi‐year sets of wave buoy measurements, as well as data from various shorter‐term buoy deployments. It is a standard wave data set used on Lake Ontario.

For the detailed design at Gibraltar Point, Baird extracted 40 years (1961 to 2000) of hourly hindcast wave data for Point 2416, offshore of the project site at a depth of 72 m.

The offshore wave height rose for Toronto Islands based on Baird’s WAVAD hindcast is presented in Figure 2.15. Figure 2.16 shows the corresponding scatter plot (point rose) of wave height by direction. In general, the wave climate off Toronto Islands is dominated by waves arriving from the east and southwest. Figure 2.17 shows joint distribution of wave height and wave direction in tabular format. Figure 2.18 graphically shows the distribution of offshore wave height and energy by direction. The offshore wave energy is predominantly from the easterly (azimuth 80‐90°) and southwesterly (azimuth 210‐230°) directions. The hindcast indicates that waves arrive from the south (S) to west (W) window about 43% of the time with southwest (SW) being the predominant direction. Wave heights up to 4.5 m occur during severe storms from SW direction. Waves arrive from east‐southeast (ESE) to east‐northeast (ENE) window about 35% of the time with east (E) being the predominant direction. Wave heights up to 5.5 m occur during severe storms from the E direction. Wave height is less than 1 m and 2 m for 85% and 97% of the time, respectively. Wave height is larger than 1 m and 2 m for 15% and 2% of the time, respectively. Shoreline erosion events are expected to occur mostly during major storm events. Predicted wave periods range between 2 to 10 seconds (s) from the east and 2 to 8 s from the southwest.

As discussed in Section 2.4.5, there is uncertainty that climate change will increase the frequency and intensity of wind‐generated waves. However, the uncertainty can be accommodated in detailed design by considering a range of plausible increases in wind‐generated wave heights over the life of the project.

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Figure 2.15 Offshore wave height rose (Point 2416)

Figure 2.16 Offshore wave height scatter plot (point rose) (Point 2416)

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Figure 2.17 Offshore wave height distribution by direction

6 18

Maximum Wave Height 5 15 Wave Energy

4 12 (%) Energy Wave Total

3 9

Wave Height (m) Height Wave 2 6

1 3

0 0 0 60 120 180 240 300 360 Wave Direction (deg)

Figure 2.18 Distribution of wave height and wave energy by direction

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The Shoreplan (2007) wave hindcast was not used for the detailed design as it was based solely on Toronto Island winds that were “scaled up” by a wholesale factor of 20%. Shoreplan’s approach does not use a basin‐wide wind model and calibration and verification information was not provided.

2.5.2 Nearshore Waves and Extreme Value Analysis

The offshore hourly hindcast wave data were transformed to the nearshore at the project site using numerical modelling. The nearshore numerical modelling is described in Section 2.10.

A peak‐over‐threshold (POT) extreme value analysis (EVA) of the nearshore waves was undertaken to develop design wave heights for various return periods (e.g., 50‐year, 100‐year, 200‐year). The nearshore water depth was assumed to be 6 m (future depth of lakebed ‐4.4 m CD (i.e., ‐3.5 m CD with 0.9 m of lakebed lowering) plus water level of +1.6 m). Table 2.4 provides a summary of the wave heights for given return periods along with the upper and lower confidence levels based on the 95% confidence interval. At a water depth of 6 m, the 25‐year, 50‐year and 100‐year wave heights (Hm0) are 3.30 m, 3.67 m and 3.98 m respectively. Appendix B provides the extreme value analysis of the nearshore waves.

Table 2.4 Summary of Wave Heights for Given Return Periods

Wave Height, Hm0 (m) Largest Point 5‐Year 10‐Year 25‐Year 50‐Year 100‐Year Event in Record Deepwater 4.7 5.0 5.4 5.7 6.0 5.6 (Point 2416) 32 (at water depth 6 m) 2.89 3.0 3.4 3.7 4.08 4.01 Upper Confidence Level 3.0 3.4 3.9 4.3 4.7 ‐ Lower Confidence Level 2.6 2.7 2.9 3.0 3.2 ‐

2.6 Ice

The formation of ice during winter months affects the coastal climate at the project location in two ways. First, the formation of shorefast ice, in combination with an “ice foot”, protects the shoreline area from wave action even when the main body of the lake is relatively ice‐free. The second factor is that ice formed within the greater water body has the effect of reducing wave generation during winter months, thereby limiting the wave climate at the lakefill structure.

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Ice is also a parameter that must be considered in the design of the perimeter protection of the lakefill structure. Localized damage to the perimeter armour structure can occur as a result of ice effects (e.g., bulldozing, plucking) but ice piled on shore by wind and wave action does not, in general, cause serious damage to sloped rubble‐mound structures. Typically, the net effects of ice formation are beneficial, as spray from wind and waves freezes on the structures and covers them with a protective layer of ice. Accepted practice for exposed shorelines of the Great Lakes is to size the primary armour layer to resist wave forces and accept some level of risk that ice damage could occur and that repairs may be required.

On Lake Ontario ice usually originates (and is most prevalent) at the east end of the lake next to the entrance to the St. Lawrence River. However, in cold winters it is not uncommon for ice cover to extend west along the north shore of the Lake, where it may occasionally affect Gibraltar Point. Typical winter coverage in Lake Ontario peaks around 17%. Ice coverage data is the concentration of ice, that is, the fraction of a unit of lake surface area that is completely covered with ice. The Lake is divided into grid cells and each grid cell is coded with a number between 0 and 100, representing the percentage of that cell that is covered by ice. Mild winters may only have 10% coverage while severe winters reach 65% coverage. For example, four such extreme events occurred in the winters of 1973, 1979, 1994 and 2015; in 1979 there was near‐total ice coverage on the lake. However, when it does occasionally develop, the ice cover is not very thick and the ice foot is usually less than 2 m deep.

It is likely that ice cover will have a higher frequency of a “no ice” condition in the future if global warming continues (Lofgren et al., 2002).

2.7 Shoreline Comparisons

Investigation of the entire historic shoreline evolution of Toronto Island is beyond the scope of this study. However, it is important to appreciate the complexity of the natural system and the large‐ scale human changes to the system. In the present analysis we focus on recent history of shoreline change at Gibraltar Point since completion of major human alterations to provide context for the development of the preferred option.

2.7.1 Historical Context

A comparison of the shoreline position at Gibraltar Point over time must consider that the Toronto Islands are part of a complex, naturally evolving recurved sand spit that have also been subject to large‐scale human interventions for more than 100 years. The complex natural form of the Toronto Islands and the large‐scale human interventions complicates the analysis of the shoreline conditions to determine trends and project ultimate outcomes.

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Naturally Complex Feature

As outlined in Baird (1994a, 1994b) and the ESR (TRCA, 2008), this type of natural feature is characterized by a progressive extension in the direction of the dominant sediment transport (i.e., to the southwest at this site) in the late‐glacial and post glacial periods by sediment supplied to Lake Ontario by rivers and bluff erosion along the Scarborough Bluffs. Originally, the formation of the sand spit extended from Ashbridge’s Bay to the present Toronto Islands. The shoreward directed hooks at the southwest extremity would have been initiated by periods of dominant southwest wave attack, temporarily reversing the long term southwesterly directed sediment transport caused by easterly wave attack. The curved tip of the spit, Gibraltar Point was continually shifting in response to the underlying substrate, the wave climate and the sediment supply.

Loss of Sediment Supply

The ability of the spit to continue its southward growth depended on a continued supply of sediment, both to build the subaerial part of the feature and to extend the offshore shelf southwestwards. However, increasingly the Toronto Islands were starved of sediment supply through removal of sediment from the littoral system by maintenance dredging of the Eastern Channel, by the construction of shoreline protection along the sediment producing Scarborough Bluffs, and construction of the littoral sediment blocking Leslie Street Spit. THC (1983) provides a brief description of human interventions to the Islands following a severe storm that breached the peninsula in April 1858. By 1882 the gap was about 2 km wide. In 1892, both sides of the gap were cribbed and the width reduced to 122 m. The crib jetties initially prevented sand bypassing but with time the beach east of the gap filled and sediment bypassed into the channel. Dredging of the gap stated in the 1920’s and volumes varied from 10,500 m3 to 68,000 m3 with an average of 27,000 m3 annually. All the dredged material was dumped offshore in deep water, effectively removing it from the littoral system and significantly reducing the sediment supply to the Toronto Islands.

Construction of the Leslie Street Spit commenced in the 1960’s. As it was extended lakeward, it would progressively act to prevent further longshore transport to the southwest to supply the Toronto Islands. By the mid 1960’s the Leslie Street Spit would have been effectively cutting off the supply of sediment from the Bluffs to the Islands. Protection of Scarborough Bluffs further reduced the sediment supply. Sediment in the nearshore, west of the Leslie Street Spit, would still be available for transport to the southwest. As the Leslie Street Spit (the “East Headland”) was extended further lakeward through the late 1960’s and into the early 1970’s, it would increasingly alter the nearshore wave climate at easterly portion of the Island. The East Headland portion of the Spit reached its terminus by 1978.

For nearly a century the sediment supply to the Toronto Islands was increasingly reduced and for at least the last 50 years now, Gibraltar Point has been virtually starved of a supply of sediment.

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Response to Erosion of Islands

Attempts have been made since the late 19th century to address the erosion and recession of the Toronto Islands. THC (1983) summarized some key actions:

- In 1884, a 1950 m long cedar pile seawall was constructed west of the East Gap along the south shore of what was now effectively an “island”. The seawall was reconstructed c. 1930 as a steel sheet pile wall with a recurved concrete cap. - In 1929‐30, a 520 m long detached rock breakwater was constructed to the west of the seawall (see Appendix A: Site Visit Report). - This was followed by a 90 m long wood sheet pile seawall and groynes west of the breakwater. This was then followed a 90 m long broken concrete revetment west of the wood seawall. - Gabon baskets were installed along the shore at Gibraltar Point c. 1973‐74. Figure 2.19 shows the location of the gabions and Figure 2.20 shows the location of the gabions in relation to the proposed project. The gabions began to fail shortly after they were installed. - Ad hoc amounts of concrete rubble were dumped at the shore. - A revetment was constructed to protect the washroom in 2004 (see Figure 2.3). - Concrete blocks were placed near Hanlan’s Point following a storm in 2007 (see Figure 2.21).

Figure 2.19 Location of gabion protection at Gibraltar Point (THC, 1983)

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Figure 2.20 Location of gabion basket protection at Gibraltar Point (1978) (proposed breakwater and groyne are shown for context)

Figure 2.21 Concrete blocks placed near Hanlan’s Point (photo December 2009)

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Considerable fill was also added to the islands and Toronto Harbour commencing in the mid 1800’s and continuing through the 1950’s. In the 1920’s most of the marsh in Ashbridge’s Bay was filled to create industrial land.

At the western side of the Islands, the Western Channel (Western Gap) was previously located further to the north than the present channel. To accommodate larger vessels without dredging into bedrock, and because the former channel silted in so much, it was decided to relocate the channel to the south. The relocation, including the construction of long jetties took place between 1908 and 1913 (THC, 1983). The new west channel was cut into rock with a bottom elevation of about ‐6.7 m CD. Later, to accommodate the opening of the St. Lawrence Seaway, the channel depth was increased to ‐8 m CD, mostly into rock (THC, 1977).

The jetty at the Western Gap) was extended in 1916 by 165 m and again in 1937 by 195 m when the west shore was being filled to create the present day Billy Bishop Toronto City Airport (BBTCA). Figure 2.22 shows the shoreline conditions in 1912, including both the former and present locations of the Western Gap and the areas subsequently filled by 1939. Other interventions at the western side included a 365 m long “saw‐tooth” jetty constructed in 1939 on the south side of the Western Gap and a north‐south timber seawall along the west shore. The saw‐tooth jetty was removed and replaced by with the present steel sheet pile cell jetty c. 1959 and east‐west runway was added to the airport in 1962, extending the land a further 280 m westwards.

Two water intake pipes, constructed between the late 1950’s and the early 1960’s, are located east of Gibraltar Point (see Figure 2.23). The west intake (No. 1) extends about 550 m offshore to a depth of about 20 m. The east intake (No. 2) extends to a depth of 10 m. Each intake pipe is located between two parallel steel sheet pile walls. Measurements taken from a diving survey in June 1993 revealed that long sections of the steel sheet piles were exposed with between 0.3 m and 3 m of wall protruding above the lake bed (see Figure 2.23). Both intakes are visible in the 2007 aerial image provided by TRCA. The east intake also features a 14 m long groyne formed by steel sheet pile walls infilled with concrete (see Figure 2.24; also Figure 2.2). These intakes, where exposed, act as partial barriers to sediment moving along the lakebed and at the east intake, along the shore.

In 2003, trenching for the Deep Lake Water Cooling intake pipes was undertaken (see Figure 2.25) and then the pipes were submerged into place. Initially the trench was not fully backfilled. Spoils from the dredging in the nearshore were placed in the nearshore to the west of the trench and in deeper water to the east of the pipelines. Subsequently, the trench was backfilled to some extent.

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Shoreline conditions, 1912 (THC) Lakefill on west side since 1909 (THC, 1983)

Figure 2.22 Shoreline changes to west side of Toronto Islands

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Figure 2.23 Exposure of steel sheet pile above lake bed at intake pipes in June 1993 (Baird, 1994a)

Figure 2.24 Downdrift (west) side of groyne at east intake (December 2009)

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City of Toronto Open cut drinking water dredged trench intake pipe

Figure 2.25 Dredged trench for Deep Lake Water Cooling Pipes east of Gibraltar Point (August 2003)

2.7.2 Present Littoral Cell and Descriptive Model of Sediment Transport

Gibraltar Point is now located within a litoral cell extending from west side of Tommy Thompson Park / Leslie Street Spit at the east end of the Toronto Islands to the east side of Ontario Place at the Western Gap. A littoral cell is a coastal compartment that contains a complete cycle of sedimentation including sources, transport paths and sinks. The cell boundaries delineate the geographical area within which the budget of sediment is balanced and thus provides the limits for assessing impacts of an offshore breakwater on sediment transport.

East of Gibraltar Point is protected by an offshore breakwater and seawall. Further to the east is the Eastern Channel. Tommy Thompson Park/Leslie Street Spit forms a complete littoral barrier.

West of Gibraltar Point, some sediment may have been historically transported from the distal extension of the Toronto Shelf/Spit complex towards Humber Bay by easterly wave action. However, for at least the last several decades, there has been minimal to no actual longshore sediment transport at Humber Bay due to presence of almost continuous offshore breakwaters along the Western Beaches and the littoral barriers formed by Ontario Place and the Western Gap.

The ESR (TRCA, 2008) presented a descriptive model of sediment transport at Gibraltar Point first developed by Baird (1994b). The descriptive model from Baird (1994b) is presented in Figure 2.26. It estimated that erosion at Gibraltar Point would continue in the future. The sediment supply to the Point would be of the order of 3000 m3 countered by losses of 7000 m3 to the offshore and 23,000 m3 to the northwest to Hanlan’s Beach. Further analysis and numerical modelling was undertaken in the present study to complete a detailed analysis of the sediment transport at Gibraltar Point and further develop the descriptive model to support the final design of the preferred option.

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Figure 2.26 Baird (1994b) descriptive model of sediment transport at Gibraltar Point

2.7.3 Contemporary Shoreline Evolution

A historic 1913 shoreline, a series of historic aerial photographs from 1951, 1980, 1999, 2005, and 2007, and a 2009 shoreline traced from a Google satellite image were used for shoreline comparisons in GIS. Figure 2.27 shows a comparison of the above historic shorelines.

As described in Section 2.7.1, the area has undergone significant anthropogenic changes over the past century. Of particular interest is the position of the shoreline in 1913 (yellow line) which shows that filling along the western shore for the construction of the Island Airport has significantly changed the adjacent shoreline position and orientation.

Looking at the evolution of shoreline since 1980, the change in the shoreline has been limited to two areas, namely; 1) erosion of Gibraltar Point area, and 2) accretion on Hanlan’s Beach. The remaining shoreline east of Gibraltar Point has been relatively stable.

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Figure 2.27 Historic shoreline comparison, 1913 to 2009

Figure 2.28 shows a comparison between the 1980 and 2009 shorelines with estimates of accretion and erosion areas. Over the 29 years between 1980 and 2009, about 47,000 m2 of land was eroded from the Gibraltar Point area and about 83,000 m2 was added to Hanlan’s Beach. Figure 2.29 shows a similar comparison between the 1999 and 2009 shorelines. Over the 10 years between 1999 and 2009, about 24,000 m2 of land was eroded from the Gibraltar Point area and about 35,000 m2 was added to Hanlan’s Beach. Assuming a profile height of 7 m at Gibraltar Point and 5 m for Hanlan’s Beach (based on bathymetric survey profile comparisons), this translates to an average annual erosion of approximately 14,000 m3/year in the area of Gibraltar Point and an average annual accretion of about 16,000 m3/year at Hanlan’s Beach. The above values should be considered as approximate based on a simplified volume estimation method. Nevertheless, they provide an estimate of the magnitude of actual longshore sediment transport rate in the study area.

Figure 2.30 provides a comparison of the 1982 shoreline with the 1990 shoreline with 1980 base mapping. It can be seen that a significant amount of the erosion occurred between 1982 and 1990.

The influence of the shoreline protection works can be seen in Figures 2.29 and 2.30 by the relatively small change in the shoreline position at the washroom area at Gibraltar Point and to the east.

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Figure 2.28 Comparison between 1980 and 2009 shorelines with estimated erosion and accretion areas

Figure 2.29 Comparison between 1999 and 2009 shorelines with estimated erosion and accretion areas

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Shoreline 1990

Figure 2.30 Comparison of 1982 airphoto (top panel) and 1990 airphoto (bottom panel) on 1980 base mapping (blue line) (from mapping by TRCA)

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Figures 2.28 and 2.29 indicate that most of the recent shoreline recession has occurred in the area between Gibraltar Point and Hanlan’s Point. To provide an estimate of the shoreline recession rate in this area, recession distances for various time periods were measured along Transect A in Figure 2.31. The corresponding shoreline recession rates are shown in Table 2.5. Recession rates vary through time depending on lake level at the time of the imagery as well as the accuracy of the historic lines. Focusing on the recent history of Gibraltar Point, shoreline retreat rate between 1980 and 2009 is presented in the bottom row of Table 2.5. The average annual recession rate was 4.5 m/year in this period.

Transect A

Figure 2.31 Historic shorelines and Transect A in the area between Gibraltar Point and Hanlan’s Point

Table 2.5 Calculated Average Annual Shoreline Recession Rates Along Transect A

Time Period Recession Length Years Annual Rate m/yr (m) 1951 to 1980 71 29 2.4 1980 to 1999 87 19 4.6 1999 to 2005 11 6 1.8 2005 to 2007 22 2 11 2007 to 2009 10 2 5.0 1980 to 2009 130 29 4.5

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Using an average retreat rate of 4.5 m/year, future shorelines along Transect A after 25, 50 and 100 years were estimated as shown with yellow, orange and red dots, respectively, in Figure 2.32. The location of the 25‐year point indicates that there will likely be a breach in the land separating the wetland and the lake in about 25 years. The corresponding future shorelines in the area between Gibraltar Point and the Island Airport are also shown in Figure 2.32. These future shorelines were estimated assuming that the intake and airport structures located at the two ends of predicted shorelines will be maintained and thus will stay in place in the future. It was assumed that the armour stone, concrete block and concrete rubble protection at the shore would be ineffective in the longer term. The predicted shorelines indicate that the growth of Hanlan’s Beach against the existing airport structure will likely end in about 25 years or earlier. As the shoreline reaches the tip of this structure, there will be more and more bypassing of sediment into the Western Gap and the lakeward growth of Hanlan’s Beach will slow down until it eventually ceases to grow.

Figure 2.32 Estimated future shorelines based on extrapolation of the average historic recession rate

2.8 Bathymetry Comparisons

At the time of the analysis for this study, hydrographic surveys around Toronto Island were available from 1981, 1993, 2005 and 2009. Figure 2.33 shows the corresponding survey lines. The offshore extent of each survey was different. The 2009 survey had the largest offshore extent, while the 1981 survey had the shortest extent. Several profiles in the initial 1981 survey have been reoccupied by subsequent surveys providing the opportunity to evaluate lakebed changes along individual profiles. An additional bathymetric survey was completed by TRCA in 2014. A comparison between the 2014 and the 2009 survey did not reveal any significant changes.

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A 1962 bathymetric survey is also available from the Toronto Harbour Commissioners. However, the unusual “wavy” pattern of lakebed contours, as seen in Figure 2.34, raises concerns about the reliability of the survey for detailed analysis. The bathymetry from the 1962 survey is included on the profile comparisons for information purposes and for an indicative comparison with the lakebed elevation from the 1962 Geocon boreholes (refer to Section 2.3.1.2) .

Figure 2.33 Hydrographic survey lines for 1981, 1993, 2005 and 2009 surveys and profile locations

Figure 2.34 Portion of 1962 THC bathymetric survey plan

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Figures 2.35, 2.36, 2.37 and 2.38 show the lakebed profile comparisons along profile lines E1, F1, F2 and H2. The locations of the profile lines are shown in Figure 2.33. These comparisons confirm beach profile erosion at E1, F1 F2 and sediment accumulation at H2.

The comparisons for profiles E1 (Figure 2.35) and F1 (Figure 2.36) show the profiles of the 1962 Geocon boreholes as well as the measured lakebed elevations for the CKEY radio towers. It can be seen that the elevation of the interface between the upper sand and underlying silt layers will be below the future lakebed based on estimates provided in Section 2.9. The radio towers have been removed, but remains of the foundations are thought to be visible in the airphotos.

In Profile D1 (Figure 2.39) the dredged trench for the Deep Lake Water Cooling intake pipes can be seen along with possibly the trench spoils that were dumped to west of the intake pipes.

A comparison between 1993 and 2009 bathymetry surveys was completed. In general, hydrographic surveys are subject to inaccuracies resulting from equipment limitations. Therefore, bathymetry comparisons become more meaningful for longer comparison periods where survey inaccuracies become less significant compared to actual changes that might have occurred in bottom elevations. The 1981 survey could provide a longer comparison period but had a limited offshore extent; therefore the period between 1993 and 2009 was selected for the comparison.

Figure 2.40 shows the comparison results. The comparison area generally extends to ‐6 m CD and is divided into 8 blocks (A to H). The total erosion volume from blocks D, E, F, and G at Gibraltar Point and Hanlan’s Point is about 234,500 m3 and includes DLWC dredged trench and dredge spoil areas. A bathymetry comparison does not include erosion of the beach above water level. This volume was estimated using an average erosion rate of 2000 m2/year obtained from shoreline comparisons (see Section 2.7). Assuming a 2 m height for the profile above water level, this translates into 64,000 m3 of erosion over 16 years resulting in a total erosion volume of approximately 299,000 m3. The total accretion volume in block H is about 121,700 m3. This volume includes above water level accretion because the 2009 survey extends to back of the beach. However, it does not cover the area around airport structure. This volume was estimated to be around 40,000 m3 (see Figure 2.41). The estimates of erosion and accretion rates from this are: Erosion at Gibraltar Point and Hanlan’s Point: 18,700 m3/year Accretion at Hanlan’s Beach: 10,100 m3/year.

Factors behind the difference between estimated accretion and erosion volumes may be summarized as:  Effect of survey inaccuracies over a large area resulting in overestimation of erosion  Neglect of windblown (aeolian) sand transport resulting in underestimation of accretion  Sediment loss into the Western Channel (approximately 2000 m3/year)  Possible offshore loss of fine material.

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1962, 1981, 1993, 2009, 2014 Comparison at Profile E1 2.0 1.0 E1-2014 E1-2009 0.0 E1-1993

-1.0 CKEY 2 E1-1981 -2.0 E1-1962 BH2 CKEY Tower 2 1969-82 -3.0 CKEY 3 CKEY Tower 3 1969-82 CKEY 4 -4.0 CKEY Tower 4 1969-82 -5.0 1962 Geocon Borehole 2 -6.0 COMPACT TO DENSE -7.0 GREY FINE SAND WITH OCCASIONAL FINE -8.0 GRAVEL -9.0 -10.0 Bottom Elevation (m) Elevation Bottom -11.0 -12.0 COMPACT TO DENSE -13.0 GREY SILTY FINE SAND WITH TRACE OF -14.0 ORGANICS -15.0 -16.0 0 100 200 300 400 500 600 700 800 Distance (m)

Figure 2.35 Comparison between historic bathymetric surveys along Profile E1

1962, 1981, 1993, 2009, 2014 Comparison at Profile F1 2.0 1.0 F1-2014 0.0 BH4 F1-2009 F1-1993 -1.0 BH3 F1-1981 -2.0 F1-1962 1962 Geocon Borehole 3 -3.0 1962 Geocon Borehole 4 -4.0 DENSE GREY FINE SAND WITH OCCASIONAL FINE -5.0 GRAVEL -6.0 COMPACT TO DENSE GREY FINE SAND -7.0 -8.0 COMPACT GREY SANDY -9.0 SILT WITH OCCASIONAL ORGANIC SEAMS -10.0 Bottom Elevation (m) Elevation Bottom -11.0 COMPACT GREY FINE SILTY SAND WITH ? -12.0 OCCASIONAL ORGANIC SEAMS -13.0 -14.0 -15.0 -16.0 0 100 200 300 400 500 600 700 800 Distance (m)

Figure 2.36 Comparison between historic bathymetric surveys along Profile F1

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1962, 1981, 1993, 2005, 2009, 2014 Comparison at Profile F2 2.0 1.0

0.0 F2-2014 -1.0 F2-2009 -2.0 F2-2005 -3.0 F2-1993 -4.0 F2-1981 F2-1962 -5.0 -6.0 -7.0 -8.0 -9.0 -10.0 Bottom Elevation (m) Elevation Bottom -11.0 -12.0 -13.0 -14.0 -15.0 -16.0 0 100 200 300 400 500 600 700 800 Distance (m)

Figure 2.37 Comparison between historic bathymetric surveys along Profile F2

1981, 1993, 2005 and 2009 Comparison at Profile H2 2.0

H2-2009 0.0 H2-2005

-2.0 H2-1993

H2-1981 -4.0

-6.0

-8.0

-10.0 Bottom Elevation (m) Elevation Bottom

-12.0

-14.0

-16.0 0 100 200 300 400 500 600 700 800 Distance (m)

Figure 2.38 Comparison between historic bathymetric surveys along Profile H2

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1962, 1981, 1993, 2005, 2009 Comparison at Profile D1 2.0 1.0 D1-2009 0.0 D1-2005 D1-1993 -1.0 D1-1981 -2.0 D1-1962 -3.0 -4.0 -5.0 -6.0 -7.0 -8.0 -9.0

Bottom Elevation (m) Elevation Bottom -10.0 -11.0 -12.0 -13.0 -14.0 -15.0 -16.0 0 100 200 300 400 500 600 700 800 Distance (m)

Figure 2.39 Comparison between historic bathymetric surveys along Profile D1

Figure 2.40 Lakebed surface comparison of accretion and erosion between 1993 and 2009 surveys

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Figure 2.41 Distribution of accretion along Hanlan’s Beach between 1993 and 2009

In summary, it may be concluded that erosion at Gibraltar Point (shore and lakebed) provides on average about 14,000 m3 to 18,000 m3 of sand to the littoral system each year, most of it is transported to northwest and deposited along Hanlan’s Beach.

THC (1983) estimated that along the south side of Toronto Islands (2800 m from Algonquin Park to Gibraltar Point) over a 300 m wide nearshore zone there was a net loss of 401,000 m3 in period between 1972 to 1981, or about 5 cm/yr of lakebed lowering. The area of Gibraltar Point (Sta. 3.8 km to 4.4 km, see Figure 2.19) experienced a loss of 194,000 m3 over the nearshore zone in the same period, or approximately 21,500 m3/yr. THC (1983) estimated that 487,000 m3 accumulated along the west side between Gibraltar Point and the south jetty at the Western Gap from 1972 to 1981.

The steel sheet pile lined water intake pipes to the east of Gibraltar Point act as a partial barrier to sediment transport. The DLWC intake pipes are buried in the nearshore. The material dredged in the nearshore for the trenches was placed back over the pipes and also in spoil piles in the nearshore to the west of the pipes and therefore has likely had minimal influence on the erosion at Gibraltar Point. The material dredged for the trenches further offshore was deposited to the east of the pipes in depths roughly corresponding to the depths from which it was dredged. Some infilling occurred, but this material was likely not “lost” from where it originally was in the littoral system. The trench in the offshore is visible in Profile D1 (Figure 2.39).

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2.9 Lakebed Lowering Associated with Shoreline Erosion

Shoreline erosion on a sandy shore (cohesionless or non‐cohesive shore) is a manifestation of the overall beach profile erosion. With respect to its horizontal position, the shoreline at Gibraltar Point is experiencing transgression (i.e., the shoreline is moving landward). This is accompanied by transgression of the nearshore profile which in effect is lowers the nearshore profile. The amount of lakebed lowering associated with beach profile erosion decreases with increasing depth along the profile. As the profile erosion continues, the rate of lakebed lowering at a given point decreases with time because of the increase in depth at that point and the reduced effect of the waves.

The lakebed lowering rate at the location of the proposed breakwater at ‐ 3.5 m CD was estimated by comparison with the same depth in the older surveys. Table 2.6 summarizes the results at the E1, F1 and F2 profiles (see Figures 2.42, 2.43 and 2.44 respectively; note exaggerated vertical scale) for the periods 1981to 1993, 1993 to 2014 and overall from 1981to 2014. This period is considered to be representative of the present coastal conditions, as described in Section 2.7.

In the period 1981 to 1993, lakebed lowering at ‐3.5 m CD ranged from 2.3 to 4.3 cm/yr. From 1993 to 2014, the lowering rates ranged from 1.2 to 1.9 cm/yr. The overall rate from 1981 to 2014 varied from 1.8 cm/yr to 2.7 cm/yr.

Table 2.6 Historic Lakebed Lowering Rates at ‐3.5 m CD Lakebed Lowering Rate at ‐3.5 m CD (cm/year)

Time Period Profile E1 Profile F1 Profile F2

1981 ‐ 1993 2.3 4.4 4.3

1993 ‐ 2014 1.4 1.9 1.2

1981 ‐ 2014 1.8 2.7 2.1

The results in Table 2.6 indicate that lakebed lowering rate appears to have decreased in the 1993‐ 2014 period. The reason for the decrease is not clear. Although the location of ‐3.5 m CD along a profile in 1993 was different from its corresponding location in 1981, the two locations are only about 60 m apart and it is unlikely that the observed differences are due to non‐homogeneity of lakebed material. The accuracy of earlier surveys may be a factor, however this cannot be verified. Changes to the sediment supply and wave climate, as described in Section 2.7.1, may also be factors.

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For the design of the proposed breakwater, it is reasonable to assume an average lakebed lowering rate of 2.2 cm/year over 33 years which translates to an overall lowering of 0.7 m. Over the remaining 17 years of a 50 year structure design life (see Section 4), the lakebed lowering rate would be less than the initial 33‐year period because it would be in deeper water. From Profile F2 (Figure 2.44) at a starting depth of ‐4.2 m CD, the lowering rate between 1993 and 2104 was 1 cm/yr; over 17 years. This would result in an additional 0.2 m of lakebed lowering. Therefore, the total lowering over 50 years is assumed to be 0.9 m (i.e., 0.7 m + 0.2 m). It should be noted that erosion of the Gibraltar Point shoreline will be controlled by the proposed breakwater and the corresponding nearshore profile lowering will thus be reduced and likely less than 0.9 m over the next 50 years. Generally, design of coastal structures includes local toe scour protection for a scour depth comparable to this lakebed lowering estimate.

It is expected that lakebed lowering over the 50‐year design life of the structure will take place in the fine sand upper layer.

Additional local scour caused by wave‐structure interaction will be addressed through design of the toe scour protection as presented in Section 4. Geotechnical investigations (boreholes) prior to construction of the breakwater are recommended to confirm the scour protection design.

1981, 1993, 2009, 2014 Comparison at Profile E1 2.0

E1-2014 1.0

E1-2009 0.0 E1-1993

-1.0 E1-1981

-2.0

-3.0 Bottom Elevation (m) Elevation Bottom

-4.0

-5.0

-6.0 200 250 300 350 400 450 Distance (m)

Figure 2.42 Lakebed lowering at Profile E1

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1981, 1993, 2009, 2014 Comparison at Profile F1 2.0

F1-2014 1.0 F1-2009

F1-1993 0.0 F1-1981

-1.0

-2.0

-3.0 Bottom Elevation (m) Elevation Bottom

-4.0

-5.0

-6.0 200 250 300 350 400 450 Distance (m)

Figure 2.43 Lakebed lowering at Profile F1

1981, 1993, 2005, 2009, 2014 Comparison at Profile F2 2.0

1.0 F2-2014 F2-2009 0.0 F2-2005 F2-1993

-1.0 F2-1981

-2.0

-3.0 Bottom Elevation (m) Elevation Bottom

-4.0

-5.0

-6.0 200 250 300 350 400 450 Distance (m)

Figure 2.44 Lakebed lowering at Profile F2

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2.10 Numerical Analysis of Existing Coastal Conditions

Numerical modeling using a two‐dimensional hydrodynamic (2DH) model was undertaken to calculate wave transformation and resulting nearshore currents that occur in Gibraltar Point area. The objective of the simulation was to 1) obtain an overall understanding of waves, nearshore currents and their corresponding sediment transport in the study area, and 2) evaluate various proposed design alternatives to control the erosion of Gibraltar Point through comparisons with existing conditions. Existing conditions are discussed in the present sections. Design alternatives are examined in the Section 3.

2.10.1 Model Selection

Two candidate models were considered for the present study, namely: 1) the MIKE21 package developed by Danish Hydraulic Institute (DHI), and 2) HYDROSED, a Baird in‐house model.

HYDROSED is a 2DH hydrodynamic and sediment transport state of the art model for coastal areas. It consists of a spectral wave transformation model (where the wave field is calculated by the spectral energy conservation equation of Karlsson 1969, with the breaking dissipation term of Isobe, 1987), a hydrodynamic model (Nishimura, 1988) to describe wave generated nearshore currents and circulations (driven by radiation stresses predicted with the spectral wave transformation model) and a sediment transport model presented by Dibajnia et al (2001). The sediment transport model considers the influence of non‐linear orbital velocities and undertow and is based on the sheet flow transport formula of Dibajnia and Watanabe (1992), which was extended by Dibajnia (1995) to consider suspended transport over ripples as well as the bedload transport. Dibajnia et al (2001) also conducted a sensitivity test of their model and showed that the model response under various actual nearshore wave environments is satisfactory. For a given wave condition, HYDROSED can provide a full spatial description of nearshore currents and sand transport over the calculation domain, for example around a harbor. The model has been verified through laboratory experiments as well as field measurements and has been extensively applied to several projects by Baird in the past few years.

Nearshore wave transformation and corresponding nearshore current calculations were completed for two storm events, one from the southwest and the other from the east, with both models. Comparisons of wave height vectors for the southwesterly and easterly storms are shown in Figures 2.45 and 2.46, respectively. Results from both models are very similar. Corresponding calculated nearshore currents are compared in Figures 2.47 and 2.48, respectively, for the SW and E events. Predicted nearshore currents by the two models have similar patterns and velocity in the nearshore area. It was therefore, concluded that both models provide acceptable results. HYDROSED model was selected for this project due to its flexibility in running a large number of wave conditions and post‐processing the results.

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Figure 2.45 Comparison of wave height vectors from MIKE21 and HYDROSED for southwest storm

Figure 2.46 Comparison of wave height vectors from MIKE21 and HYDROSED for east storm

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Figure 2.47 Comparison of nearshore currents from MIKE21 and HYDROSED for southwest storm

Figure 2.48 Comparison of nearshore currents from MIKE21 and HYDROSED for east storm

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2.10.2 HYDROSED Modeling of Existing Conditions

The existing conditions model bathymetry was composed of the 2009 hydrographic survey completed by TRCA, a 2008 Monteith and Sutherland bathymetric survey of the DLWC and East intakes, as well as Geodas bathymetry outside of the 2009 and 2008 survey datasets. As the waves arrive at this site from two completely different offshore directions, two separate grids were considered to handle easterly and southwesterly waves. Their corresponding calculation domains are shown in Figure 2.49. The SE grid included Leslie Street Spit to incorporate its sheltering effect on the study area against easterly waves. A 1042 × 840 mesh (7.8 km × 6.3 km) with grid size of 7.5 m was applied for the SE grid. The SW grid was extended westward to cover Humber Bay to avoid inappropriate waves coming through the western lateral boundary. A 989 × 1147 mesh (7.4 km × 8.6 km) with grid size of 7.5 m was applied for the SW grid. The depth at the offshore boundary of the model calculation domain was approximately 65 m.

Figure 2.49 HYDROSED model domains

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Example model results corresponding to 3.5 m wave height, 8 s wave period from the SW direction are presented in Figures 2.50 and 2.51, showing modelled wave height and direction and modelled nearshore currents respectively. This corresponds to a severe southwesterly storm condition occurring approximately once every 5 to 10 years. Calculations were completed for a lake level of +1.0 m CD to incorporate storm surge effect on water level. Figure 2.50 shows the distribution of modelled wave height (background colour shading) and direction (arrows). Waves entering the study area around Gibraltar Point undergo refraction and focusing resulting in an increase in wave energy around the Point. Figure 2.51 shows the corresponding modelled nearshore currents around Gibraltar Point for the above wave condition. Background colour shading in this figure shows the bathymetry. There is a nearshore current system initiated around Gibraltar Point that divides into two components; one easterly and the second northerly. The northerly component continues along Hanlan’s Beach and can potentially carry sediment from Gibraltar Point to this beach.

Figure 2.50 Modelled wave height and direction for a southwesterly storm

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Figure 2.51 Modelled nearshore currents for a southwesterly storm

Example model results corresponding to 5 m wave height, 10 s wave period from E direction are presented in Figures 2.52 and 2.53. This corresponds to a severe easterly storm condition occurring about once every 5 to 10 years. Figure 2.52 shows the distribution of wave height (background colour shading) and direction (arrows). Easterly waves undergo considerable refraction before reaching the study area. Highest wave conditions occur in an area offshore of the Island Pier. Figure 2.53 shows the corresponding nearshore currents around Gibraltar Point for the above wave condition. Background shading in this figure shows the bathymetry. There is a well established alongshore current system around Gibraltar Point that flows towards Hanlan’s Beach. It is therefore concluded that under both easterly and southwesterly storm events, longshore currents flow from Gibraltar Point towards Hanlan’s Beach. These currents are believed to be the main mechanism that carries sediment to this beach. The wave height around Gibraltar Point is large during southwesterly events. It is, therefore, expected that Gibraltar Point is eroded mainly by southwesterly storms and the eroded material get transported towards Hanlan’s Beach under both easterly and southwesterly events.

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Figure 2.52 Modelled wave height and direction for an easterly storm

Figure 2.53 Modelled nearshore currents for an easterly storm

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2.10.3 Long-term Analysis of HYDROSED Results

In this section, HYDROSED model results are utilized to establish long‐term average potential transport rates in the study area. Although the HYDROSED model can provide valuable insight into hydrodynamics and sediment transport patterns and rates, it is computationally time consuming and, at present, it cannot be applied to an hourly time series of wave conditions spanning more than several days. To obtain long‐term average annual longshore transport rates, the 40‐year hindcast period was represented by a selection of wave conditions and the model was run for those representative wave conditions shown in Table 2.7. In total, 110 easterly wave conditions and 75 southwesterly wave conditions were simulated. An average lake level of +0.6 m CD was assumed.

Model results were tabulated as lookup tables for a simple numerical interpolation program that stepped through the 40‐year time series of hourly wave data to determine the long‐term average sediment transport rate across several transects (cross‐sections) in the calculation domain.

Table 2.7 Modelled Wave Conditions (Height, Period, Direction) at Domain Boundary Wave Direction – SE Grid Wave Height 67.5° 90.0° 112.5° 135.0° 157.5° 0.5 m 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 1.0 m 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 2.0 m 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 4,6,8,10 s 3.0 m 6,8,10 s 6,8,10 s 6,8,10 s 6,8,10 s 6,8,10 s 4.0 m 6,8,10 s 6,8,10 s 6,8,10 s 6,8,10 s 6,8,10 s 5.0 m 8,10 s 8,10 s 8,10 s 8,10 s 8,10 s 6.0 m 8,10 s 8,10 s 8,10 s 8,10 s 8,10 s Wave Direction – SW Grid 180.0° 202.5° 225.0° 247.5° 270.0° 0.5 m 4,6,8 s 4,6,8 s 4,6,8 s 4,6,8 s 4,6,8 s 1.0 m 4,6,8 s 4,6,8 s 4,6,8 s 4,6,8 s 4,6,8 s 2.0 m 4,6,8 s 4,6,8 s 4,6,8 s 4,6,8 s 4,6,8 s 3.0 m 6,8 s 6,8 s 6,8 s 6,8 s 6,8 s 4.0 m 6,8 s 6,8 s 6,8 s 6,8 s 6,8 s 5.0 m 8 s 8 s 8 s 8 s 8 s 6.0 m 8 s 8 s 8 s 8 s 8 s

Long‐term average Longshore Sand Transport (LST) rates in the study area around Gibraltar Point predicted by HYDROSED are summarized in Figure 2.54. At Gibraltar Point the predicted annual average LST rate is 30,000 m3/year towards the northwest. It is important to note that the numerical model estimates represent a potential transport rate and it would only be realized if that quantity of sand was available for transport towards the northwest. Part of the shoreline is protected and there is some stone on the lakebed which limits the sand available for transport. The actual LST rate is thus less than the above predicted potential value. Moving towards the northwest, the predicted

Gibraltar Point Erosion Control Page 54 Final Design 11503.101 DRAFT Baird & Associates potential LST rate decreases to 18,000 m3/year and then to 11,000 m3/year along Hanlan’s Beach. This is a zone of decreasing LST rate which characterizes an accretion zone. The model also predicts that about 3,000 m3/year can potentially bypass the airport structure and enter the Western Channel. The predicted accretion rate on Hanlan’s Beach is equal to the difference between the predicted 18,000 m3/year moving to Hanlan’s Beach and the 3,000 m3/year bypassing to Western Channel. This translates to about 15,000 m3/year which is consistent with the accretion rates obtained through the shoreline and bathymetry comparisons (see Sections 2.7.3 and 2.8 respectively).

The erosion issue at Gibraltar Point is summarized in Figure 2.55.

Figure 2.54 Modelled long‐term average potential longshore sand transport (LST) rates

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Alongshore Transport by E & SW Storms

Erosion by SW Storms

Figure 2.55 Schematic summary of erosion problem at Gibraltar Point

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3.0 DETERMINATION OF PREFERRED OPTION FOR FINAL DESIGN

The ESR (TRCA, 2008) recommended five design options to be explored during the detailed design process. The recommended options include four offshore protection alternatives that consist of different lengths of surface piercing segmented breakwaters connected with submerged breakwaters (sills), and a pure sand management option with no structures. The entire offshore protection would have an overall length of 750 m to 900 m. Section 3.1 presents the results of the numerical analysis of the structural design alternatives. The preferred structural alternative is then compared to the stand‐alone management alternative in Section 3.2.

3.1 Numerical Analysis of Structural Design Alternatives

This study examined a number of alternative design options including those proposed in the ESR to evaluate their effectiveness to control the erosion of Gibraltar Point. The evaluation procedure was carried out through comparisons of resulting waves and nearshore currents with existing conditions. Existing conditions were discussed in the previous section. In total, 10 different structural protection design options were investigated involving combinations of submerged, low‐ crested, and emerged structures and a groyne option. These alternatives are divided into 4 groups and discussed in this section. Calculations were completed for a lake level of +1.0 m CD to incorporate storm surge effects.

3.1.1 Submerged Sill

A 750 m long submerged sill with sill crest at ‐2 m CD, as recommended in the ESR (TRCA, 2008), was examined. It was found that such a sill would have insignificant effect on incoming storm waves and their corresponding nearshore currents so that the wave and current field would be almost the same as those in the existing conditions. Figure 3.1 shows a comparison of nearshore currents when the proposed sill is in place with those under existing conditions for a SW storm. The two current fields are very similar. It was concluded that a sill is not capable of mitigating coastal processes at the site and would therefore be ineffective if applied to control the erosion at Gibraltar Point.

Similarly, combinations of 50% sill‒50% emerged breakwater and 33% sill‒67% emerged breakwater were modelled. It was found that with the 50% sill‒50% emerged breakwater, there is still considerable wave transmission over the sill and there is not much reduction in the intensity of nearshore currents compared to the existing conditions. The results for the 33% sill‒67% emerged breakwater option are shown in Figures 3.2 and 3.3. Figure 3.2 shows the distribution of wave height and direction. It indicates that waves are blocked by the emerged segments of the breakwater and there is limited wave transmission over the sill sections.

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Figure 3.1 Comparison of nearshore currents when a 750 m submerged sill is in place (black arrows) with existing conditions (red arrows)

Figure 3.3 indicates that the nearshore current field is mitigated to a certain degree compared to the existing conditions. Therefore, this alternative is likely to provide Gibraltar Point with partial protection against erosion. However, it will be shown in following sections that an emerged breakwater with a length roughly equal to total emerged length of the 33% sill‒67% emerged breakwater alternative is capable of providing a much better level of protection for Gibraltar Point against erosion.

It is generally concluded that a combination of sill and emerged breakwater is unlikely to provide sufficient (and efficient) protection for Gibraltar Point.

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Figure 3.2 Wave height and direction when a 33% sill‐67% emerged breakwater is in place

Figure 3.3 Comparison of nearshore currents when a 33% sill‐67% emerged breakwater is in place (black arrows) with existing conditions (red arrows)

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3.1.2 Low-Crested Structures (LCS)

A Low‐Crested Structure (LCS) is a breakwater that is submerged most of the time but may emerge during very low lake level periods. An LCS has the advantage of not obstructing the lake view and promoting fish habitat, but may be a significant navigation hazard. The crest width should be wide enough to cause sufficient dissipation following the initial wave breaking.

Two different configurations of LCS were examined, one with 3 structures with 2 gaps in between, and the other with 2 structures with one gap in between. In general, it was found that an LCS may generate additional nearshore currents, thus increasing the potential of sediment transport away from the GP area.

Distributions of wave height and nearshore currents for the 2 LCS option are shown in Figures 3.4 and 3.5, respectively. From Figure 3.4, it can be seen that waves are forced to break by the submerged breakwaters and the transmitted wave height is about 50% of the incoming wave height. There is more wave transmission through the gap section. The nearshore current figure (Figure 3.5) indicates that the intense breaking over the structures results in a flux of water resulting in additional/stronger currents compared to the existing conditions. There is also an outgoing current through the gap between the structures that could promote offshore sand losses.

In addition, COSMOS cross‐shore profile modeling of LCS (Figure 3.6) indicated that an LCS could result in higher wave setup at the shoreline compared to existing conditions. The increased setup may allow more wave energy to reach the shoreline thus reducing the effect of wave energy dissipation over the submerged structure.

It was therefore concluded that a submerged breakwater solution may not effectively protect the Gibraltar Point and is likely to create undesired sediment transport issues.

Visibility of the breakwater was also a serious concern for boater safety. It was determined that the crest should be +2.5 m CD which is 1 m above the 50‐yr static water level. This level is consistent with the crest elevation of the existing breakwater at Centre Island and the Western Beaches breakwater. The breakwater will be marked with caution lights, similar to the lights used at the Western Beaches breakwater constructed in 2005‐06.

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Figure 3.4 Wave height and direction when a pair of low crested structures are in place

Figure 3.5 Comparison of nearshore currents when a pair of low crested structures are in place (black arrows) with existing conditions (red arrows)

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Wave Set‐Up and Profile Change after 24 Hours, H = 3.5m, T = 8s, D50 = 0.20mm, SWL = 1.1m 3 1.1 Initial Profile 2 1 Breakwater 1

1 Breakwater 2 0.9

0 Breakwater 3 0.8

Final Profile (No ‐1 0.7 Breakwaters) Wave Set‐Up 1 ‐2 0.6 (m) Wave Set‐Up 2 ‐3 0.5 Wave Set‐Up 3 Setup(m)

Elevation

‐4 0.4 Wave Set‐Up (No

Breakwaters) Wave ‐5 0.3 Bottom

‐6 0.2

‐7 0.1

‐8 0

‐9 ‐0.1

‐10 ‐0.2 400 450 500 550 600 650 700 750 800 850 900 950 1000 1050 1100 1150 Distance from Offshore (m)

Figure 3.6 Beach profile change and wave setup for various LCS options

3.1.3 Emerged Breakwaters

An emerged offshore breakwater (also called detached breakwater) is a surface piercing breakwater structure that is not connected to land. Its function is to protect the shoreline by simply blocking the incoming waves. A 900 m long emerged breakwater option and another 750 m long emerged breakwater with a groyne option were initially considered in the ESR (TRCA, 2008). These options are examined in this section. In addition, 500 m, 550 m, and 600 m long breakwaters with an auxiliary groyne options as well as a two breakwater option were also investigated.

Figure 3.7 shows distribution of wave height and direction for the 900 m long breakwater option. In addition to the area between Gibraltar Point and Hanlan’s Point, this option also protects almost half of the beach between Gibraltar Point and the intake structure. It was found that with the excessive sheltering provided by this long breakwater structure, it is likely that any sediment carried from the east by westward longshore currents would be trapped behind this structure. The trapped sediment, however, could not be transported back to the east under SW wave conditions. This may eventually result in closure of the gap between the structure and the island potentially causing water quality issues.

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Figure 3.7 Wave height and direction when a 900 m long breakwater is in place

Several alternative breakwater lengths were examined. It was found that the combination of a 550 m long offshore breakwater and a groyne could effectively protect the shoreline at Gibraltar Point. The groyne is required to stabilize the shoreline to the east of the GP and mitigate the possibility of sediment being trapped behind the breakwater.

Distributions of SW wave height and nearshore currents for 550 m long offshore breakwater and groyne option are shown in Figures 3.8 and 3.9, respectively. From Figure 3.8, waves are blocked by the breakwater effectively reducing the wave height in the area between Gibraltar Point and Hanlan’s Point. The nearshore current figure (Figure 3.9) indicates that nearshore currents are considerably mitigated by the breakwater. At the north end of the structure adjacent to Hanlan’s Point a circulation pattern is formed that in the nearshore flows southward towards Gibraltar Point. Figure 3.10 shows the resulting nearshore currents under easterly storm conditions. Although alongshore currents are considerably mitigated, there remains some flow through the gap between the breakwater and the groyne to promote circulation/exchange of the water behind the breakwater.

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Figure 3.8 Wave height and direction under SW waves with the 550 m breakwater and groyne option

Figure 3.9 Comparison of nearshore currents under SW waves when a 550 m breakwater with a groyne are in place (black arrows) with existing conditions (red arrows)

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Figure 3.10 Comparison of nearshore currents under E waves when a 550 m breakwater with a groyne are in place (black arrows) with existing conditions (red arrows)

3.1.4 Groynes

The use of only groynes on their own was not recommended in the ESR (TRCA, 2008). A system of two large groynes was examined here for the sake of completeness of our evaluation of design alternatives. Groynes are generally an effective and least expensive option to stabilize a shoreline where the littoral system is dominated by alongshore transport of sediment and cross‐shore transport is relatively weak. On the Great Lakes in general, there often is a strong cross‐shore component to the transport that typically makes groynes an undesirable option.

Distributions of wave height and nearshore currents for the two‐groyne option are shown in Figures 3.11 and 3.12, respectively. From Figure 3.11, waves can arrive at the shoreline without any major interruptions. The nearshore current figure (Figure 3.12) indicates that nearshore currents are deflected by the groynes to some extent, but preserve their overall strength. It is expected that after construction of the groynes, the shoreline will continue to erode until it stabilizes at an equilibrium planform. Figure 3.13 shows the estimated future shoreline for this option. Two possible future shorelines are shown in this figure depending on whether the existing shore protection at Gibraltar Point is maintained (green line) or removed (red line). The groyne solution could be combined with sand management or be applied as a stand‐alone option by adjusting the length of the groynes.

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Figure 3.11 Wave height and direction under SW waves when a two groynes are in place

Figure 3.12 Comparison of nearshore currents under SW waves when two groynes are in place (black arrows) with existing conditions (red arrows)

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Figure 3.13 Estimated future shorelines with two groynes if the shoreline protection at Gibraltar Point is maintained (green line) or removed (red line)

3.2 Analysis of Selected Final Design Alternatives

As previously mentioned, the design objectives of the present project are to 1) control erosion of Gibraltar Point, and 2) mitigate the impacts on natural processes at Hanlan’s Beach. Based on the discussion in previous sections, two design alternatives were selected to be considered in more detail as a solution to control erosion at Gibraltar Point and maintain natural processes at Hanlan’s Beach:

1. Stand‐alone sand management approach. This alternative requires an annual placement of sand at Gibraltar Point to offset shoreline erosion; and

2. Structural protection works with a sand management program. This option requires a 550 m breakwater, a 130 m long partially submerged groyne with a reactive sand management program. In this scenario, the breakwater prevents wave attack of the shoreline around Gibraltar Point while the groyne helps to stabilize the beach to the east of Gibraltar Point. A limited sand management is also required to preserve the dynamics of Hanlan’s Beach and its associated dunes. The sand management program is reactive, meaning that sand is placed only when required, and in much smaller volumes than the stand‐alone sand management approach.

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3.2.1 Alternative #1 – Stand-Alone Sand Management

Alternative #1 focuses only on sand management that would replenish the sand lost through erosion by placing “new” sand in the eroding area. The sand placement area is shown in Figure 3.14. Sand management would be required proactively on an ongoing basis (i.e., every year) to be prepared for upcoming storm events and their sequencing in order to protect Gibraltar Point from further recession.

Figure 3.14 Sand placement area for Alternative #1, Stand‐Alone Sand Management

Using the numerical model results discussed in Section 2.10.3 for longshore sediment transport (LST), the time history of annual potential longshore sediment transport rate at Gibraltar Point was calculated. Potential transport rate assumes loose unconsolidated sands are available to be transported by the wave energy across the entire nearshore profile. Figure 3.15 shows the modelled variation of the annual net LST at Gibraltar Point. It is estimated that on average approximately 30,000 m3 of sand would be required at Gibraltar Point every year to maintain a sufficient reserve of sand to protect against further recession of the present shoreline position. However, Figure 3.15 also indicates that this value could vary between 15,000 m3 and 60,000 m3 depending on the wave climate in individual years.

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Variation of Annual Net LST at Gibraltar Point 80000

Average LST: 30000 m3/year

70000

60000

50000

40000 /year, positive westward) 3

30000

20000 Net LST (m Net Annual

10000

0

1960 1965 1970 1975 1980 1985 1990 1995 2000 Time (year)

Figure 3.15 Variation of annual net LST rate at Gibraltar Point

Activities typically involved with sand management (beach nourishment) include:

 Dredging suitable sand material from a local source using a hydraulic or mechanical dredge  Transporting the dredged sand material to shore to be nourished by pumping directly by pipeline (sand‐water slurry mixture) or by loading and offloading barges  Depositing the sand slurry onto the shore directly by spaying or outletting the pipe into a temporary bermed settling pond on the beach  Grading the placed sand to a finished profile.

Figure 3.16 illustrates the typical activities involved with sand management (i.e., “beach nourishment”). The activities at the beach will disrupt the beach use every year for the duration of the sand placement. The production rate will depend on a number of factors, including equipment used, the spatial extent and thickness of the sand source, the distance of the source from the placement site, wave conditions (seasonal weather) and environmental restrictions (e.g., fisheries windows). For illustrative purposes, it is estimated that it would take approximately 15 days (working 24 hrs/day) to 40 days (working 8 hrs/day) to dredge and place 30,000 m3 of sand from a local site. Public use of the beach and the shore would be restricted during the beach nourishment process.

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Hydraulic dredging of sand source Mechanical dredging of sand source

Pumping sand slurry through pipe from hydraulic Spraying sand slurry onto beach dredge to beach

Settling pond on beach Grading placed sand

Figure 3.16 Examples of beach nourishment activities

An option of placing a permanent sediment pumping station near the Western Channel that would replenish the sediment lost at Gibraltar Point was considered. Such a system requires a steady supply of sand and a fixed station is not flexible. This approach was not considered viable because only 3,000 m3 per year is deposited at the Western Channel.

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Potential Sand Sources

An evaluation was undertaken of potential sources of sand for the sand management option. The evaluation is provided in Appendix D. In summary, local nearshore or offshore sand sources do not have sufficient volume or suitable grain size and/or quality for ongoing, sustainable sand management at Gibraltar Point. Imported sand is required for the pure sand management option. There is currently about 60,000 m3 of sand available at the Western Channel. However, this only represents about 2 years of supply for the full sand management approach while replenishment into the Channel is only in the order approximately 3000 m3 per year. There are sand sources at Ashbridge’s Bay and Coatsworth Cut and offshore of Eastern Beaches. However, the cost of retrieving these sands and transporting it to Gibraltar Point would be greater than locally sourced sand. Also, there are expectations that sand materials in these areas should be used for local bypassing/renourishment efforts of Eastern Beaches (i.e., not removing sand from one littoral cell to nourish another littoral cell).

Currently, sediment originating from the Don River is dredged from the Keating Channel at a rate of approximately 35,000 m3 per year. These sediments are predominantly very fine sand, silt and clay and are not representative of the grain size distributions that exist off of Gibraltar Point. Also, using current technologies, the sediment does not meet Ministry of Environment open lakefill guidelines for contaminants (i.e., it exceeds lowest effect level threshold). Technologies may be developed for the Don Mouth Naturalization and the Port Lands Flood Protection and Enabling Infrastructure Project that might allow for the reuse of some of those materials in the future.

Opinion of Probable Cost of Alternative #1 ‐ Stand‐Alone Sand Management

The estimated net present value (NPV) cost of the stand‐alone sand management option (Alternative #1) is $21 million using a discount rate of 3% over a time horizon of 50 years (assuming unit rate of $25/m3 (from ESR (TRCA, 2008)) plus $65,000 for each mobilization/demobilization and $5000 survey costs). Annual costs are estimated to be $820,000. These costs are based on the assumption that all sand can be obtained locally using a hydraulic or mechanical dredge system. These costs increase if sand requires barging from other locations along the waterfront, river mouths or excavated from aggregate sources and trucked to the site. The sensitivity of the NPV to the discount rate is shown in Table 3.1.

Table 3.1 Net Present Value Cost of Alternative #1, Stand‐Alone Sand Management Discount Rate Net Present Value (NPV) Cost

2% $25.8 M 3% $21.1 M 4% $17.6 M

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If no sand management is practiced, the shoreline will gradually erode before becoming stable in approximately 50 years. Hanlan’s Beach would continue to exist, though with a smaller footprint and coarser grain sizes.

3.2.2 Alternative #2 – Offshore Breakwater, Groyne and Focused Sand Management

Alternative #2 relies on an engineered structural feature to control erosion of Gibraltar Point, as envisioned by the ESR (TRCA, 2008). This includes the construction of an emerged 550 m long detached breakwater, with a 100 m long, shore‐connected groyne (see Figure 3.17) and a reactive, focused sand management program. The numerical modelling of Alternative #2 was described in Section 3.1.3. The breakwater will protect the shore from waves that erode the shoreline and the groyne will stabilize and promote deposition of sediment at the beach located east of Gibraltar Point. Focused sand management is then also required to replenish the small average annual loss of beach material that would occur at Hanlan’s Beach northwest of the breakwater.

550 m Detached Offshore Breakwater and Groyne

Alternative #2 will effectively protect the shoreline between Gibraltar Point and Hanlan’s Point. The groyne will stabilize the shoreline to the east of Gibraltar Point (i.e., the shoreline between Gibraltar Point and the water intake groyne/structure). Once the Gibraltar Point shoreline is stabilized, any future change in Hanlan’s Beach shoreline could be addressed through focused sand management (“reactive sand management”).

Figure 3.17 Alternative #2, 550 m long offshore breakwater and 100 m long shore‐connected groyne

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The offshore breakwater will be 550 m long. It will be curved to follow the existing lakebed contours at a depth of about 3.5 m below chart datum (CD). The crest width of the main trunk section will be 5.5 m and the crest elevation will be about +2.5 m CD. As such, the top of the breakwater would be at an elevation of about 76.7 m. With the breakwater located 250 m or more offshore, its relatively low profile will not present a significant visual impact from shore (i.e., similar to offshore breakwater to the east of Gibraltar Point and the offshore breakwater at the Western Beaches. The breakwater will be constructed from armour stone and core stone material, similar to the Western Beaches breakwater constructed in 2005 and the stone breakwater to the east of Gibraltar Point.

The groyne will extend out from the shore to ‐2.5 m CD and will be about 100 m long. The groyne will be constructed from armour stone and core stone material, similar to the groyne headlands at the east end of the Eastern Beaches (Figure 3.18).

Figure 3.18 Groynes at Eastern Beaches

Focused Sand Management for Hanlan’s Beach

Using COSMOS, numerical computer modelling of sediment transport on Hanlan’s beach was performed. The annual variation of the sand material lost and gained at Hanlan’s Beach is presented in Figure 3.19 and ranges from 1000 to 6000 m3. The simulated results indicate that Hanlan’s Beach will lose, on average, approximately 3000 m3 per year of sand to the Western Channel and will be required to be replenished if the breakwater is placed at Gibraltar Point. Figure 3.20 shows the proposed sand placement area. From that area, the southwest storms will move the material northwards along Hanlan’s Beach.

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Annual Variation of Volume Lost/Gained 0.0

‐1.0 /m) 3 (m

‐2.0 ) ‐ (

‐3.0

‐4.0 Gained(+)/Lost

‐5.0 3 Volume Average = ‐2.3 m /m ‐6.0

‐7.0 1961 1966 1971 1976 1981 1986 1991 1996 2001 Time (year)

Figure 3.19 Annual variation of sand volume loss from Hanlan’s Beach

Due to the relatively small average annual volume required, the sand could be replenished once every 3 to 5 years (9000 m3 to 15,000 m3) before the storm season (i.e., in early September), or following particularly large storm events. This would reduce mobilization and demobilization costs. Placing the sand would act to maintain Hanlan’s Beach in its current configuration, and will allow for dynamic sand dune processes to continue. The present growth of Hanlan’s Beach would not continue, as there would no longer be an excess of sediment supply derived from the ongoing erosion at Gibraltar Point.

If the beach sand is not replenished, as the fine sediments are lost through wind and currents, the beach will consist of coarser materials. The dunes will not disappear, but will stabilize with new plant communities growing on them. Effective beach management practices will be critical to enhancing and maintaining beach‐dune complex at Hanlan’s Beach. These practices are independent of the proposed undertaking for the Gibraltar Point Project.

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Opinion of Probable Cost of Alternative #2 – Breakwater, Groyne and Focused Sand Management

Using a discount rate of 3% over a time horizon of 50 years, the net present value cost of the proposed breakwater and groyne with focused sand management is approximately $17.6 million (assumes 15,000 m3 of sand placed on the beach every 5 years and annual maintenance cost of 0.5% per annum for the breakwater and groyne). The initial cost of the breakwater and groyne is estimated to be $12.9 million plus an additional $0.8 million for aquatic habitat enhancements, with the remaining $3.9 million (NPV) for the continued focused sand management and structure maintenance over the design life of the project. The sensitivity of the NPV to the discount rate is shown in Table 3.2.

Table 3.2 Net Present Value Cost of Alternative #2, Breakwater, Groyne and Focused Sand Management Discount Rate Net Present Value (NPV) Cost

2% $18.6 million 3% $17.6 million

4% $16.9 million

Given that only approximately 3000 m3 of sand would be required on an annual basis, and given that approximately 3000 m3 of sand is available at the Western Channel every year, sand for the focused sand management could be derived entirely from local sources on a long‐term basis. This would reduce the cost risk factor of having to source and transport suitable sand material from other sources.

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Figure 3.20 Sand placement area for focused sand management as part of Alternative #2

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3.2.3 Evaluation of Selected Alternatives

Table 3.3 summarizes the evaluation of the selected alternatives.

Table 3.3 Summary Evaluation of Selected Alternatives Breakwater & Groyne with Stand‐Alone Sand Management Criteria Focused Sand Management (Alternative #1) (Alternative #2) Protect Gibraltar Point from Yes Yes erosion Minimize impact on natural Yes Yes processes at Hanlan’s Beach Cost (NPV @ 3% over 50 years) $21.1 million $17.6 million Risk* Higher Lower Sand source Multiple, Non‐local Western Channel Protection of existing Yes Yes infrastructure Stabilize the beach east of GP Potentially Yes

Aesthetics of lake view Maintained Diminished Potential for creation of new No Yes park space at Gibraltar Point Modest requirement for Requires significant annual Operation and Maintenance dredging and back‐passing sand expenditure to maintain project from Western Channel *See discussion on Risk Components:

Risk Components Stand‐Alone Sand Management (Alternative #1): - Timing must be pro‐active): Average requirement for 30,000 m3/year of sand, but impossible to predict the exact annual requirements; may require up to 60,000 m3/year to be placed immediately following storm to maintain protective beach at Gibraltar Point in state of readiness. - Greater uncertainty of securing future budget commitments ($820,000 per year) - Greater sensitivity to cost escalation due to increases in sand costs (due to diminished supply, increased demand for other construction activities in GTA; environmental permitting)

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- Multiple, non‐local sand sources required due to larger volume (10 times the volume of Alternative #2) - Length of time required for beach nourishment and available operational windows - Less design certainty - Environmental impact of dredging sand at source; annual permitting requirements - Possible negative public perception regarding continued loss of placed sand - Potential for introduction of non‐native species.

Breakwater and Groyne with Focused Sand Management (Alternative #2): - Less certainty regarding adverse physical impacts (on local scale) - Impact on habitat (direct loss from breakwater) and requirement for compensation measures.

3.3 Conclusions and Preferred Alternative for Final Design

Erosion of the shore and lakebed at Gibraltar Point provides on average about 14,000 m3 to 18,000 m3 of sand to the littoral system. Most of this sand is transported northward and deposited at Hanlan’s Beach.

A stand‐alone sand management approach (Alternative #1) requires having to source and place approximately 30,000 m3 of sand annually on average. Actual erosion events, however, are episodic making it impossible to predict the exact annual requirements, which may be as much as 60,000 m3/year. Alternative #1 would cost approximately $21.1 million (net present value) over 50 years.

A 550 m long detached breakwater and a shore‐connected groyne together with focused sand management (3000 m3/year) at Hanlan’s Beach (Alternative #2) would protect the shoreline at Gibraltar Point, stabilize the shoreline to the east of Gibraltar Point, and minimize impacts on the natural processes at Hanlan’s Beach. Alternative #2 would cost approximately $17.6 million (net present value) over 50 years.

The stand‐alone sand management alternative (Alternative #1) is estimated to cost 20% more on a net present value basis than the breakwater and groyne with focused sand management alternative (Alternative #2). Alternative #1 has a higher risk as it requires significant annual budgeting commitments in the future (i.e., $820,000 annually) and is more sensitive to escalating costs of imported sand. For example, if the unit rate for dredging/supplying sand is $40/m3 (similar to 2014 price for Bluffer’s Park dredging; pers. comm., TRCA), the net present value cost of the pure sand management option escalates to $32 million, while the offshore breakwater, groyne and focused sand management option net present value cost increases only to $18.7 million.

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Alternative #2 was selected as the preferred alternative for completing the final design. Figure 3.21 presents a view of the preferred alternative, including an offshore, detached breakwater, a shore‐ tied groyne and periodic, focused sand placement. “Additional sand” placement is also shown in the lee of the breakwater. This additional sand is separate from the focused sand management and is discussed in Section 4.4.

Shore‐tied Groyne

Periodic Sand Additional Sand Placement Placement (future configuration) Offshore Breakwater

Figure 3.21 Offshore breakwater and groyne with focused sand management (Alternative #2) option selected for final design (Additional sand placement shown is an additional feature and not part of original scope; see Section 4.4)

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4.0 FINAL DESIGN

This section describes the final design of the offshore breakwater and the shore‐connected groyne along with the aquatic habitat enhancements and the “additional sand placement” component of the project. The additional sand placement is an added feature that was not part of the original scope of work and is discussed in Section 4.3.

4.1 Offshore Breakwater and Shore-Connected Groyne

4.1.1 Summary

The final design for the offshore breakwater and the shore‐connected groyne are presented in Figures 4.1, 4.2 and 4.3.

Figure 4.1 shows the breakwater length measured along the curved crest as 525 m; with the addition of the roundheads at each end, the total length is about 550 m. The centre‐line of the breakwater generally follows the existing ‐3.5 m CD contour. The groyne is connected to the existing shoreline at the easterly end of the stone revetment protection at the washroom. The groyne is 94 m long measured along the crest and extends out to the ‐2.5 m CD contour. The breakwater and groyne are located as per the numerical modelling described in Section 3.

The cross‐section design of the rubble‐mound offshore breakwater is shown in Figure 4.2 and consists of a primary armour layer over a stone core. The offshore breakwater will be similar to the Western Beaches breakwater. All materials are natural, quarried stone. The breakwater is to be built on a stone bedding layer placed directly on the existing lakebed, with no excavation. The primary armour layer of the lakeward face of the breakwater trunk will consist of 3 layers of 2‐6 tonne armour stone at a slope of 2:1 (horizontal:vertical). The crest and rear slope of the breakwater will consist of 2 layers of 2‐4 tonne armour stone. The crest elevation is +2.5 m CD. The breakwater roundheads consist of 4 layers of 2‐6 tonne armour stone at a slope of 2:1. The core is 300‐600 mm diameter riprap stone. An intermediate filter layer is not used. A toe berm and a scour apron are provided.

Similarly, the cross‐section design of the rubble‐mound groyne consists of a primary armour layer of quarried stone over a stone core as shown in Figure 4.3. The armour layer varies from a single layer of 1‐3 tonne armour, to a single layer of 2‐6 tonne armour, to a double layer of 2‐6 tonne stone. The crest elevation varies from +2.2 m CD to +2.5 m CD. The core is 300‐600 mm diameter riprap stone. An intermediate filter layer is not used. The groyne is to be built on a stone bedding layer. The inner portion of the groyne will be excavated into the lakebed to elevation ‐1.5 m CD. The outer portion of the groyne will be on the existing lakebed. A toe berm and a scour apron are provided. The shore‐connected groyne will be similar to the groyne/headlands constructed at the east end of the Eastern Beaches.

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Figure 4.1 Final design plan view

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Figure 4.2 Final design cross‐sections of breakwater

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Figure 4.3 Final design cross‐sections of groyne

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4.1.2 Rubble Mound Design Guidelines

There are no standard marine codes in Canada for the design of rubble‐mound structures. Generally accepted guidelines and practices for design life, level of damage, and acceptable risk were used as references for the design of rubble‐mound structures.

4.1.3 Design Life

The design life of a structure is the period of time during which it is expected to function properly. There are no standard codes for selecting the design life. This life expectancy of the structure is typically selected by the Owner along with the level of accepted risk. Both of these parameters help dictate the design conditions. It is important to recognize that the design life is not equivalent to the return period of the design conditions.

For works and installations of local auxiliary interest, such as defense and coastal regeneration works, works in minor ports or marinas with a small risk of loss of human life or environmental damage in the event of failure, the minimum design life should be 25 years (PIANC, 2003). TRCA selected a design life of 50 years. This is compatible with accepted practice (e.g., BS 6349‐7:1991; ISO 21650:2007; Lamberti, 1992; PIANC, 2003; PIANC, 1992; ROM 0.2‐90; Pullen et al., 2007).

4.1.4 Damage Levels

Rubble‐mound stone damage is generally progressive, with increasing number of stones removed and increasing volume of eroded profile as the wave height increases and/or the duration of the storm increases. Rubble‐mound structures are routinely designed with an acceptable level of risk to survive severe storms with some moderate damage to the armour layers while accommodating ‘tolerable’ damage as a result of more extreme events (e.g., USACE, 2006 and PIANC, 1992). This is considered acceptable practice and designs approaching ‘absolute safety’ are not required (PIANC, 1992). To be considered ‘tolerable’, damage to the armour layers should be progressive and repairable with a sufficient cover layer of armour over the underlayer (e.g., typically at least one layer of armour) such that the underlayer material is not exposed. Exposure of the underlayer as a result of the loss of armour units from the outer layer is considered a failure of the protection.

Quantification of damage to stone armouring is, to a large extent, subjective and qualitative. Damage to the armour layer can be characterized by assessment of the number of armour units displaced (USACE, 2006).

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Damage Level, D

The damage level, D, of the armour layer of the exposed front slope can be defined as the percentage volume of armour stones eroded relative to the total volume of stones in the “active zone” (USACE, 1984; 2006). The active zone is the area directly exposed to wave action, and for the Gibraltar Point breakwater this zone extends from the middle of the crest to the toe. The damage level for two‐layer rock armoured slopes is given in Table 4.1.

Table 4.1 Damage Level, D for Two‐Layer Rock Armour on Slope 1:2 to 1:3 (USACE, 2006) Initial Damage Intermediate Failure Damage 0‐5% 5‐10% >20%

The “zero‐damage” criteria given in the Shore Protection Manual (USACE, 1984) corresponds to a D value less than or equal to 5% using the eroded volume method for two‐layer armour. For two‐ layer structures, the Shore Protection Manual (USACE, 1984) defines “failure” as greater than 30% damage to the armour layer, which corresponds to exposure of the underlayer and the imminent “unravelling” of the structure. This guidance is primarily given for breakwater slopes, typically 1:1.5 to 1:2.5.

Damage Parameter S, Based on Measured Profile Erosion

Van der Meer (1988), as reported in USACE (2006) and CIRIA (2007), used a damage definition based on Broderick’s (1984) non‐dimensional description of the cross‐sectional area eroded by wave action (i.e., damage S = Ae/ Dn502; see Figure 4.4).

Protective Outer Layer

Underlayer

Figure 4.4 Schematic illustration of eroded cross‐sectional area, Ae

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A damage value of S = 2 provides a good estimate of the initiation of damage for a two‐layer armour at slope of 1:1.5 to 1:2 and that S = 8 is indicative of failure, while values of S = 3‐5 and 4‐6 are indicative of intermediate damage for 1:1.5 and 1:2 slopes respectively (USACE, 2006; CIRIA, 2007). For a breakwater trunk with a front slope of 1:1.5, Vidal et al., (1992) set minimum threshold values of S for the different definitions of damage; Table 4.2 presents these threshold values. PIANC (1992) indicates that ‘moderate’ damage, which can be repaired economically for a breakwater armoured with rock can be represented by a value of S = 6.

Table 4.2 Damage Level S for Two‐Layer Rock Front Breakwater Slope (1:1.5) (Vidal et al., 1992)

Damage Level Description S

Initiation of Damage: certain number of armour units are displaced a distance more 1 than Dn50 leaving holes larger than the average pore size in the armour layer.

Iribarren Damage: the number of units removed from the upper layer is large enough 2.5 that a unit in the lower layer can be dislodged.

Start of Destruction: initiation of damage to the lower armour layer such that a 4 number of units are displaced.

Destruction: material from the secondary (filter) layer is removed and armour units 9 leave the slope continuously.

Based on the above guidance, the following damage levels were considered during the design of the rock armouring:  ‘moderate’ damage level as S = 4 for a 1.5:1 slope and S = 5 for 2:1 slope  ‘tolerable’ damage as S not exceeding 7 for both 1.5:1 and 2:1 slopes.

The proposed design of the breakwater has 3 layers at the trunk and 4 layers at the roundhead.

4.1.5 Acceptable Level of Risk of Damage

Coastal structures are designed to resist a harsh marine environment governed largely by the passage of episodic storm events. The greatest storm to occur over the life of the structure can only be estimated at the time of design, therefore there is always some inherent risk associated with the design conditions and it is important that the owner has a clear understanding of these risks. The storm duration and frequency will also influence the design.

Here, risk is defined as the probability that a given design event (e.g., a specified combination of monthly mean water level, storm surge and wave height) will be reached or exceeded at least once during the project life. If the design event is reached or exceeded, there will be certain

Gibraltar Point Erosion Control Page 86 Final Design 11503.101 DRAFT Baird & Associates consequences that must be taken into consideration. For example, if the design conditions are exceeded, there may be damage to the breakwater or groyne and a subsequent possibility of environmental impacts and direct or indirect economic costs.

The acceptable level of risk of damage is established by the Owner based on considerations of project function, human health and safety implications, environmental consequences, capital budget, economic impacts resulting from damage to, or destruction of the structure and ability or willingness to carry out repairs and maintenance. There are no firm rules for selecting the level of acceptable risk. Based on accepted industry practice (e.g., ISO 21650:2007; PIANC, 2003; PIANC, 1992; Lamberti, 1992; and ASCE, 1994), the level of acceptable risk for damage to the rubble‐mound breakwater and groyne was set as follows:

 40% risk for ‘moderate’ damage to the structure, including minor armour unit displacement, tolerable settlement

 20% risk for ‘tolerable’ damage to the structure, including toe or armour failure or breach of a significant portion of structure.

PIANC (2003) characterizes these as “indicative estimates on reasonable acceptable failure probabilities, which ensure designs are not far from existing practice.” These risk values were selected based on the following factors:

 no loss of human life is expected in case of damage or failure of the structure

 economic repercussion of damage or failure of the structure is considered to be low (i.e., generally less than five times the cost of the structure)

 environmental consequences of damage or failure of the structure is considered to be low (e.g., material used in core is clean, inert material and breach of the core would not immediately follow initial damage to the rock armour and repairs could be initiated before further significant damage occurred)

 proposed structures are flexible structures (i.e., damage is progressive due to the reserve strength of the rubble‐mound structure) and generally repairable.

A risk of 40% is within the guidance provided by PIANC (1992) which suggests that a higher value of risk, of the order of 50% to 60%, for small to moderate damage within the service lifetime is acceptable, provided that the damage can be repaired economically and the resulting design has a low value of risk (say 10% to 20%) for major damage. The lower risk value results in a greater return period design condition (i.e., higher wave height, higher water level).

Maintenance of the armour stone protection in response to the deterioration or “wear and tear” will likely occur. An allowance for repairs to the structure which may be required as a result of storms exceeding the design event should therefore be considered. Maintenance is discussed in Section 5.

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4.1.6 Return Period of Design Event

Once the acceptable level of risk and the design life are established by the Owner, the resulting return period event can be determined for use in the design. The return period of the design event is a function of the specified level of acceptable risk and the project design life and the relationship is shown in Figure 4.5.

Figure 4.5 Graphical relationship of design event return period, risk and design life

Given an acceptable risk of 40% for moderate damage and a design life of 50 years, the design event has a 100‐year return period. For a 20% risk for tolerable damage, the design event has a 225‐year return period. The design return periods are summarized in Table 4.3.

Table 4.3 Risk, Project Design Life and Design Event Return Period for Rock Armour Damage Level Approximate Level of Design Return Period of Design (slope) Acceptable Risk Life Event Moderate Damage S = 4 (1.5:1) 40% 50 years 100‐year S = 5 (2:1) Tolerable Damage S = 7 20% 50 years 225‐year (1.5:1 & 2:1) Major Damage 10% 50 years 475‐year

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4.1.7 Design Water Levels and Wave Conditions

For design purposes, the waves and static water levels were assumed to be independent events. Surge and waves are not independent events (i.e., the wind creating the waves will also result in a surge). It was assumed for the purposes of this final design that the return periods for waves and surge were similar. Return periods for the static water levels, storm surge and wave conditions are provided in Sections 2.4.3 and 2.5.2 respectively. The wave conditions were considered using 0.9 m of lake bed lowering which will occur progressively over the design life (Section 2.9). Thus a 250‐ year return period design event could be comprised of a range of combined events, such as a 25‐ year static water level combined with a 10‐year wave and a 10‐year surge, or a 10‐year static water level combined with a 25‐year wave and 25‐year surge. The static water level and the surge values were summed to provide a total design water level.

The design water level and wave conditions by risk level are summarized in Table 4.4.

Table 4.4 Design Water Levels and Wave Heights

Combination of Wave Height Return Period of Wave, Surge and Total Water Level Design Event Static Water Level (Upper Return Periods Confidence)

20‐yr wave & surge / 3.3 m +1.37 m CD 100‐year Return 5‐yr static water level (3.8 m) ~40% Risk 5‐yr wave & surge / 2.8 m +1.55 m CD 20‐yr static water level (3.0 m) 25‐yr wave & surge / 3.4 m +1.51 m CD 250‐ year Return 10‐yr static water level (3.9 m)

~20% Risk 10‐yr wave & surge / 3.0 m +1.64 m CD 25‐yr static water level (3.4 m)

50‐yr wave & surge / 3.7 m +1.52 m CD 500‐year Return 10‐yr static water level (4.3 m)

~10% Risk 10‐yr wave & surge / 3.0 m +1.75 m CD 50‐yr static water level (3.4 m)

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4.1.8 Geotechnical

It is recommended to undertake geotechnical boreholes at the time of construction, when the offshore barge is mobilized, to confirm the lakebed conditions. The foundation of the rubble‐ mound structure should then be assessed for the following construction and long‐term conditions:

 bearing capacity

 slope stability (slip surface) for both static and seismic conditions

 total and differential settlement

 liquefaction (seismic activity).

Settlement of a breakwater can occur by settlement of the subgrade (lakebed) or by densification of the rockfill. It is expected that the breakwater will be constructed on a fine sand layer several metres thick (Section 2.3). Available evidence (Section 2.3) is that bedrock is relatively deep below the lakebed and will not influence the design.

It has been assumed that some initial settlement of the sand will occur during placement of the bedding, core and armour. Some settlement of the armour layer is expected as the material consolidates and nests. Additional material may be required to maintain the design crest elevation. The required “overheight” depends on the lakebed characteristics, the height of the structure and the construction method. The bedding layer volume has been increased by 20% as an allowance.

4.1.9 Armour, Filter, and Core Design

To simplify construction, the number of different stone types and gradations was limited:  armour stone o A1 – 4‐6 tonne for rear crest and rear slope of breakwater o A2 – 2‐6 tonne primary armour for breakwater and outer section of groyne o A3 – 1‐ 3 tonne for breakwater toe berm and inner section of groyne  “core” stone (300 – 600 mm diameter rip rap)  bedding stone (2 – 250 kg).

A multi‐layer primary armour approach was selected to limit the individual size of the armour stone required and to provide a flexible structure that will well accommodate potential settlement and scour. Larger armour stone units are increasingly difficult to source and quality control for the large stone blocks is a concern. During the design and tender process for the Western Beaches breakwater, alternative designs were issued for smaller multi‐layer armour and larger double layer armour and it was determined that the smaller multi‐layer approach was more cost effective.

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The rip rap stone was selected as the “core” material in order to function as both core and filter (underlayer beneath armour layers). Due to the geometry of the cross‐sections, if a separate filter layer and core (using traditional quarry run material) were used, the volume of core required would be minimal. It would not be efficient to introduce a small volume of another type of material.

It is anticipated that the breakwater will be constructed by marine plant. Staging and loading of the stone materials would be in Toronto Harbour.

The groyne would be constructed by land‐based equipment from the shore. Stone material would be brought by to the Island and transported across the Island to the site. The crest level of the core of the inner portion of the groyne is +1.3 m CD and the width is 3.5 m to permit construction access.

4.1.9.1 Armour Stability

The hydraulic stability design of the armour stone was based on previous Baird experience with physical modelling of rubble‐mound breakwaters, particularly the Western Beaches breakwater (Appendix E), and standard empirical methods of Hudson and van der Meer to determine stone mass, W, for sloped structures, including shallow water adjustments where the depth of water is less than three times the wave height (d < 3Hs), as presented in USACE (2006) and CIRIA (2007). Test conditions for the Western Beaches physical model were similar to the conditions for Gibraltar Point. The test condition presented in Appendix E is more severe than the 250‐yr return period event for Gibraltar Point. The resulting damage level is “tolerable”. Physical model testing of the Gibraltar Point design was not within the scope of the present work and was not undertaken.

The stability coefficient, Kd, in the Hudson formula accounts for a number of factors such as type of waves (breaking or non‐breaking), location on structure (trunk or head), stone shape (irregular, cubic or rounded) and placement (number of layers). The selection of an appropriate Kd value is very important, and must consider many factors. Baird’s physical modelling experience provided key guidance for selection of the Kd values. Design wave and water level conditions for various risk and damage conditions are presented in Section 4.1.7. The effect of oblique wave attack was neglected to be conservative for armour sizing.

Accepted practice for exposed shorelines of the Great Lakes is to size the primary armour layer to resist wave forces and accept some level of risk that ice damage could occur and that repairs may be required. The performance of the existing breakwater at Toronto Island since 1930 suggests that armour stone can reasonably be expected to perform adequately under ice action.

4.1.9.2 Armour Stone Placement

Tight armour stone placement is an important factor in the performance of a rubble‐mound structure, including multi‐layer armouring. The specifications require that armour stones be placed such that they are “carefully keyed in and interlocked to provide a compact and integrated surface

Gibraltar Point Erosion Control Page 91 Final Design 11503.101 DRAFT Baird & Associates course”. A loosely placed armour layer is more susceptible to stone motion during wave action, resulting in an increased potential for displacement of armour stones and/or armour stone deterioration, both of which may lead to increased damage and maintenance/repair requirements. Tight stone placement will also limit the maximum acceptable “gap” between adjacent armour stones, with the objective of minimizing the potential for the loss of smaller underlying materials (i.e. filter and core) through the armour layer.

4.1.9.3 Armour Stone Quality

Stone quality is an important factor in the performance of rubble‐mound structures. The stone specifications require the armour stone material to meet specific quality requirements to minimize long‐term maintenance of the structure and provide stability during the design conditions.

It is important to emphasize the possible negative performance and maintenance outcomes from utilizing rock that is of questionable durability in the marine environment. Substandard materials will be susceptible to deterioration/degradation (such as abrasion and freeze‐thaw‐related fracture and break up) which will lead to reduced stone sizes, decreased stability, and an increased risk of damage to the structure (and associated requirements for maintenance/repair).

As noted earlier, the stability of the armour layer of a conventional breakwater is described by Hudson’s equation (USACE, 1984). The stable stone weight increases with the cube of the wave height. Other important parameters that affect the stability are the stone density, stone shape and placement (incorporated in the stability coefficient, Kd) and the structure slope. Clearly, stone degradation (i.e. abrasion, fracture, break up, etc.) may reduce the stone size and result in a reduction in the stability of the armour layer, with a consequential increase in the risk of damage to the structure.

4.1.9.4 Filter/Underlayer and Core

As noted, rip rap stone was selected as the “core” material in order to function as both the core and filter (underlayer beneath armour layers). The design of the filter stone followed standard practice provided in USACE (2006) and CIRIA (2007). The 300‐600 mm diameter rip rap is a common size produced by quarries servicing the GTA.

4.1.10 Crest Height and Crest and Rear Slope Stability

Table 4.5 presents a summary of the criteria considered in determining the crest height of the breakwater. A crest height of +2.5 m CD for the breakwater was selected. The armour crest design was based on the standard minimum width of 3Dn50.

The size of armour required for the crest and rear slope of the breakwater is typically larger than the front slope when the relative freeboard (freeboard/wave height, F/Hs) is less than 0.75. Under design conditions, the relative freeboard for Gibraltar Point is approximately 0.25 to 0.3. Stone

Gibraltar Point Erosion Control Page 92 Final Design 11503.101 DRAFT Baird & Associates sizing and configuration for stability of the crest and rear slope was based on Baird’s physical modelling experience, particularly for the Western Beaches breakwater (Appendix E).

Table 4.5 Offshore Breakwater Crest Height Considerations Criteria Lower Crest Height Higher Crest Height

Visibility (safety) for Worse Better boaters

Transmission of wave energy and resultant More Less erosion at shore Visual seascape from Less impaired More impaired shore

Width of crest Increases Decreases Size of crest and rear Increases Decreases slope armouring Size of front slope Decreases Increases armouring Capital cost Decreases Increases

4.1.11 Bedding Layer (Filter)

A 0.5 m thick bedding layer of 2‐250 kg stone is used under the core and the toe berm core to limit the loss of the finer underlying lakebed material (fine sand) through the coarser toe stone and core stone. The bedding layer is the same material as the scour apron material.

4.1.12 Toe Berm

A stone toe berm or bund is used at the toe of the primary armour layer in accordance with empirical guidelines (e.g., BS6349‐7:1991; USACE, 2006; and CIRIA, 2007). There are no definitive methods for designing toe protection and professional judgment is required in evaluating the various guidelines. The stability of the toe berm stones (i.e., size of stone) was estimated using the methodology in CIRIA (2007, Equation 5.188) and a value for Nod of 0.5 representing ‘start of damage’, which is considered safe for design. Initially, the toe of the berm is at ‐3.5 m CD (neglecting the thickness of the bedding layer). The stone size was checked for low water and high water conditions:  Low water: The sum of the 50‐year static low water level (‐ 0.2 m CD) and a 10‐year storm setup (surge) (0.22 m) is approximately +0.0 m CD. The wave condition was considered as depth‐limited (~0.65 x depth).  High water: 255‐year return period design event (Table 4.4).

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The toe berm consists of 2 layers of 1‐3 tonne armour on a bedding layer.

The toe berm width is 4.5 m (4 stones wide). This is consistent with empirical guidelines: BS6349‐ 7:1991 suggests a minimum width of 4 stones, while USACE (2006) provides for a typical minimum width of 3 to 5 stones.

The groyne toe is an extension of the primary armour and is two stones wide.

4.1.13 Scour Apron

A scour apron is provided beyond the toe berm of the breakwater. As with the toe stability analysis, it should be noted that there are no definitive methods for designing scour protection at this time and considerable professional judgment is required. The width of the scour protection apron was evaluated based on consideration of several empirical guidelines (e.g., Sumer and Fredsøe, 2002; Delos – Deliverable 43, 2002; and USACE, 1984). These methods provide estimates of the required width of the scour protection apron. The breakwater scour apron is 5 m wide and is comprised of the bedding material (2‐250 kg stone).

The groyne scour apron is 3 m wide.

4.2 Navigation Aids

The crest elevation was selected based on requirements for erosion protection (Section 3.1.2) and to permit the breakwater to be visible to boaters, as discussed in Section 4.1.10). Navigation aids, similar to those used at Western Beaches breakwater (Figure 4.6) will be placed at each end of breakwater.

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Figure 4.6 Typical navigation aid

4.3 Aquatic Habitat Enhancements

Aquatic habitat enhancements were determined through a series of meetings with TRCA biologists and Aquatic Habitat Toronto (AHT). The proposed aquatic habitat enhancements are shown in Figure 4.7 and consist of a series of stone shoals (Areas “A” to “G”) located at the roundheads of the breakwater, in the lee of the breakwater and at the lee side of the groyne. Random, isolated armour stone are also included to create additional relief. TRCA and AHT staff confirmed that stone shoals could be comprised of rip rap material similar to that being used for the breakwater filter/underlayer. The rip rap for Areas A and F also provide added scour protection at the breakwater roundheads.

The numerical modelling (Section 3.1) provides estimates of the expected conditions in the lee of the proposed breakwater between the beach and the structure. Wave heights may range from about 0.1 m in the middle area immediately behind the breakwater to about 0.3 m near the beach and to about 1 m towards both ends of the breakwater. Nearshore currents would generally be less than 0.3 m/s.

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Figure 4.7 Aquatic habitat enhancements

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4.4 Additional Sand Placement

Discussions with staff at Ministry of Natural Resources identified the potential to place additional sand, in the lee of the breakwater structure. The intent is to attempt to create additional beach area and partially recreate a natural dune system. The additional sand would be deposited by barge in the nearshore. Waves would then naturally carry the sand behind the breakwater. It is recognized that the area behind the proposed breakwater will be more sheltered than an open shoreline and the beach‐dune system will not act in the same manner as natural system. This component was not within the original scope of the project.

Nearshore sand placement was considered as an effective way to nourish the beach behind the proposed offshore breakwater at Gibraltar Point. It would allow for flexible delivery from alternate sources. In this approach, the placed sand will be sorted out by natural forces of waves and nearshore currents such that the ʺbeach sandʺ portion reaches the beach. Figure 4.8 presents a plan for the nearshore placement of the additional sand.

Definition of the appropriate placement location and depth is critical for successful nourishment. It is also necessary to estimate the rate of transport towards the beach to define the frequency and volume of each placement. The methodology used was as follows:

1. Using previous model results, an approximate placement area was identified. Sand should be placed between the 1 m and 2 m depth contours inshore of the west end of the proposed breakwater. 2. Given operational limitations of the placement barge, top of the placed sand should be approximately at ‐0.7 m CD (i.e., approximately 1.5 m depth at average peak summer lake level to accommodate the barge, based on experience at Tommy Thompson Park (pers. comm., TRCA). 3. 2DH numerical modelling of waves, nearshore currents and sediment transport was completed using HYDROSED for two scenarios: 1) design conditions with the breakwater and groyne with a placed mound of sand representing the conditions immediately after placement; and 2) design conditions with the breakwater and groyne without the placed sand representing the conditions when all sand has been carried away by wave action. As the placed sand moves towards the beach, interim transport rates were estimated by interpolation between the above two scenarios. 4. Using the model results, the rate of transport towards the beach and the frequency of the placement were determined.

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Figure 4.8 Additional sand placement plan

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There is a potential capacity to accommodate approximately 50,000 m3 of additional sand, resulting in a beach area of about 24,000 m2 at elevation +1 m CD. The preference is to undertake the placement in manageable stages of about 10,000 m3 at a time. It was estimated that it would take approximately 3 to 6 years for natural waves and currents to move the placed sand from the nearshore to the shoreline (depending on annual wave conditions). Figure 4.9 shows a schematic of the predicted movement of sand placed in the lee of breakwater following a one‐time placement.

The placement would be carried out 5 times to achieve the target sand volume on the beach. Figure 4.10 presents a schematic view of the projected final beach configuration following multiple placements.

Wave‐induced currents

Initial underwater placement area Wave‐induced sediment movement

Figure 4.9 Predicted movement of sand placed in lee of breakwater following one‐time placement

Consideration was given to the placement of 50,000 m3 of additional sand all at once versus 5 times 10,000 m3 over a period of 12 to 24 years. The preference is to take an adaptive management approach by placing an initial volume of 10,000 m3, monitor the performance of the placement and then make adjustments based on the observed outcome. The difference in the costs of the two approaches was evaluated using net present value (NPV). The NPV (3% discount rate) of placing 10,000 m3 every 5 years, ending in year 20 is $1 million (at unit rate of $25/m3 and including mobilization costs and monitoring costs for each occurrence). If the unit rate increases to $40/m3 the NPV increases to $1.4 million. The NPV of placing the additional sand all at once is $1.4 million, at $25/m3 and $2.1 million at $40/m3. The staged approach is less costly based on net present value.

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Toe of beach slope

Total Additional Sand Placement ~50,000 m3 (~24,000 m2 beach area at +1 m CD)

@ +1 m CD

*Following multiple underwater nearshore placements and redistribution by natural wave currents and possible beach grading

Figure 4.10 Additional beach sand showing projected final beach configuration following multiple placements

The option of excavating a wide trench to lower the base of the breakwater to match the anticipated lake bed lowering (i.e., 0.9 m) and use the excavated sand material for placement at the shore was also evaluated. This option is not cost effective because an additional volume of stone, equivalent o the volume of sand removed, would be required to maintain the crest height of the breakwater at the prescribed elevation. The additional stone is more expensive to supply and place (nominally $100/m3) than the value of the dredge sand (nominally $25‐40 /m3).

4.5 Estimated Quantities

The estimated quantities to construct the breakwater, groyne, aquatic habitat enhancements and the initial “additional sand” placement are provided in Table 4.6.

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Table 4.6 Estimated Quantities of Final Design

Item Item Description Quantity Units No. 1 Mobilization, Demobilization, Construction Facilities lump sum and Control and Site Preparation, Demolition, 1 Removals and Restoration 2 Stone Storage Area and/or Staging Area Rental Fee 1 lump sum 3a A1 stone (4 to 6 tonne armour stone) 24,000 tonnes 3b A2 stone (2-6 tonne armour stone) tonnes 36,000

3c A2 stone (2-6 tonne armour stone) to be used for Fish tonnes Habitat 200

3d 13,000 tonnes A3 stone (1-3 tonne armour stone)

4a Core Stone/Fish Habitat Stone 32,200 tonnes 4b Bedding Stone 25,200 tonnes 5 Sand Supply and Nourishment 10,000 cubic metres 6 Excavation 1,300 cubic metres 7 Private Aids to Navigation (including foundation & each 2 all fittings)

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5.0 FUTURE BREAKWATER MONITORING AND MAINTENANCE

5.1 Breakwater Monitoring

5.1.1 Introduction

Rubble‐mound structures respond to wave and ice action, water level changes and cycles of freeze‐ thaw and wetting and drying by changes in the profile of the structure and to the size and shape of its component parts. The armour stone may be displaced, cracked/fractured and/or abraded. Major failures by storm action are easily identified (e.g., displaced stone, slumping slopes, reduced crest width). Gradual degradation caused by settlement, fracture or abrasion may be much less apparent.

Post construction monitoring of the completed breakwater is strongly recommended. The monitoring program should include the establishment of an as‐built database of cross‐sections, stone gradations and photographs (general slope conditions as well as close‐ups of specific areas), all documented by location to allow duplication in the future. The roundheads and corners breakwaters are typically where damage, if any, might be expected first and warrant particular attention during inspection. A qualified coastal engineer with experience in breakwater design and performance should complete the monitoring. Subsequently, these cross‐sections, stone gradations and photographs should be repeated and compared to the as‐built information, thereby allowing quantification of changes and identifying the requirement, if any, for maintenance and repair work.

5.1.2 Frequency of Monitoring

The frequency of monitoring is presented in Table 5.1.

Table 5.1 Breakwater Monitoring Frequency Timing Monitoring Action Follow-up Action as Comments Required at Areas of Interest After Significant Routine Visual -Profile survey Hs>2.5 m (waves Storms Inspection -Diving inspection at breakwater) Annual (late Routine Visual -Profile survey spring) Inspection -Diving inspection

5.1.3 Routine Visual Inspection

A routine visual inspection of the breakwater should be carried out every year in the late spring and following any major storm. The routine inspection would consist of a visual reconnaissance along the breakwater crest and the above water portions of the front and rear slopes. A small boat is required to access the breakwater during the inspection. The station numbers along the crest

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 Slumping or settlement of slope or crest

 Reduction in crest width

 Large gaps between armour stones

 Dislodged or displaced armour stones

 Stone degradation (deterioration, cracking, fracture, spalling, rounding of corners).

A photographic log along the length of breakwater should be prepared. The annual survey should be compiled in digital format for future reference.

5.1.4 Detailed Inspections and Surveys

If the routine inspection uncovers significant deficiencies in the breakwater, additional detailed inspections should be undertaken as required, including profile surveys and diving surveys.

Profile surveys are recommended every 5 to 10 years.

5.1.5 Safety Considerations

Rubble‐mound structures are potentially hazardous areas on which to operate. The stone is uneven with gaps and the sloping front and rear faces pose an increased risk. Individual stones can be smooth and can be particularly slippery when wet. Near the waterline, water level variation exposes weed and/or algal growth making the surface slippery and movement can be dangerous and requires great care. Monitoring should not be undertaken when ice is present on the breakwater.

A team of at least two people should carry out the monitoring. All personnel should wear personal floatation devices. The monitoring team should be familiar with the nature of the lake including storm waves and should check the marine weather forecasts on a regular basis. Special care shall be exercised on the slope of the structure, near the water’s edge and on wet or slippery surfaces. When surveying in the water from a boat, all applicable boating safety regulations shall be observed. Diving should only be carried out qualified personnel and in strict accordance with all applicable worker safety rules and regulations.

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5.1.6 Profile Survey Procedures

5.1.6.1 Conventional Survey

A conventional survey is conducted using a rod and total station at regular intervals along the cross‐section profiles. Recognized survey procedures are also presented in EM 1110‐2‐1003 Engineering & Design ‐ Hydrographic Surveying (U.S. Army Corps of Engineers 2002).

Conventional surveys are limited in providing an accurate picture of the underwater portion of the breakwater. The natural/inherent variability in the armour layer is such that changes are difficult to detect using conventional survey techniques. Conventional surveys at spaced locations do not provide coverage between the profile lines. Newer, more comprehensive technologies are introduced in the following section.

5.1.6.2 Multibeam and Other Technologies

As an alternative to the conventional surveying technique, multibeam and sidescan sonar surveys provide detailed, high‐resolution data below the water, while aerial oblique photos/video, photogrammetric mapping and laser scanning can provide this information above the water. The collected multibeam data is used to create a complete three‐dimensional, digital terrain model (DTM) of the breakwater surface.

Newer technologies, including CODA Echoscope and UAV’s, have been used effectively by Baird.

5.1.6.3 Diving Surveys

If the routine visual inspection or the profile surveys indicate that further investigation is required, a diving survey may be necessary. Underwater inspections shall be conducted under the direction of a qualified engineer. The dive team should be equipped with voice communication systems and video equipment. A surface monitor connected to the video cameras carried by divers shall also be provided. The engineer shall be present on site to view the underwater conditions from the surface monitor for the duration of the underwater inspection and shall direct the divers with respect to the information and/or measurements that should be obtained. The engineer responsible for the field inspection shall prepare the underwater inspection reports.

5.1.7 Monitoring Armour Stone Degradation

Monitoring procedures are intended to identify armour layer damage as given by: 1) cavities; 2) unstable armour; and 3) stone deterioration. Stone deterioration may arise from abrasion, breakage due to movement or impact, spalling and facture due to weathering (feeze‐thaw; wetting and drying).

Monitoring armour degradation should be included in the routine visual inspection.

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Armour degradation can also be monitored in a more detailed, systematic manner at sample locations. At each sample location an area containing at least 100 armour stone is marked out. Each survey area should run from as low down the front slope as possible, up the slope, across the crest and down the rear slope as low as possible. It would be preferable to complete the monitor survey at a time when the water levels are low but prior to or after potential ice conditions. Typically, the survey area should be at least 5 m wide.

The first step is to count the total number of armour stones falling within the survey area. Generally it is convenient only to count those in the uppermost layer. Then the condition of each of the armour stones is recorded. A cavity is recorded where a void could be filled by an armour stone of the design size. Factured armour includes all examples where stones have broken in place. Sub‐size armour is all armour smaller than the specified lower limit. The state of interlock of the armour layer may be assessed by two methods. Unstable armour may be defined where the restraint given by adjacent units is reduced.

5.2 Maintenance and Repairs

The shoreline of Lake Ontario is a harsh environment (e.g., wave action, abrasion by suspended sediment, ice forces, freezing and thawing) and maintenance of the armour stone rubble‐mound protection structure in response to the deterioration or “wear and tear” of the structure can be expected. In addition to the replacement costs due to long‐term deterioration of the stone, there should be an allowance for repairs to the structure required as a result of storms exceeding the design event. As noted in Section 4.1, risk levels have been assumed for damage and failure of the breakwater.

Repairs might include replacing displaced or cracked armour stones and backfilling slumped core material at the toe of the back slope. Available data on maintenance costs of marine structures is limited. However, an estimate for an annual maintenance and repair budget of between 0.5% and 1.0% of the initial capital cost would be reasonable. It is unlikely that actual maintenance and repair work would be necessary every year, but the annual allowance should be budgeted on an ongoing basis to have sufficient funds accumulated when repair work is required.

Marine plant will be required to carry out repairs. The plant will include a crane and/or backhoe mounted on spud barge, material delivery barges and tugs to move the barges to and from the site and to position the barges around the site. Figure 5.1 shows typical marine plant required for breakwater repair work.

Work in the water may require requires permits from Toronto Region Conservation Authority, Department of Fisheries and Oceans and Ministry of Natural Resources. Stone material should be sound, durable and meeting the original specifications. The material should be clean, inert rock or concrete rubble meeting the lakefilling requirements of the Ministry of Environment.

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The navigational lights should be inspected and maintained on an annual basis. The lamp, battery and solar panel could be removed for the winter and reinstalled each spring prior to the boating season.

Figure 5.1 Typical marine plant required for repair work (top ‐barge mounted crane; bottom –backhoe on barge and tugs)

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6.0 REFERENCES

Angel J.R., and Kunkel, K.E., 2010. The response of Great Lakes water levels to future climate scenarios with an emphasis on Lake Michigan‐Huron. Journal of Great Lakes Research, 36 (Supplement 2), P.51, January.

ASCE, 1994. Planning and Design Guidelines for Small Craft Harbors, Revised Ed., ASCE Manuals and Reports on Engineering Practice No. 50. American Society of Civil Engineers.

Baird & Associates, 2003. Lake Ontario WAVAD Hindcast for IJC Study. Report prepared for U.S. Army Corps of Engineers and the International Joint Commission by W.F. Baird & Associates, October.

Baird & Associates, 1994a. Toronto Island Shoreline Management Study Report on Coastal Processes. A report prepared for Metro Toronto and Region Conservation Authority, February.

Baird & Associates, 1994b. Toronto Island Nature School Study Final Report. A report prepared for Metro Toronto and Region Conservation Authority, October.

British Standards Institution, BS 6349‐7: 1991. Maritime Structures – Part 7: Guide to the Design and Construction of Breakwaters.

Broderick, L.L., 1984. Riprap Stability Versus Monochromatic and Irregular Waves, M. thesis, George Washington University, USA.

CIRIA, C. CETMEF, 2007. The Rock Manual. The use of rock in hydraulic engineering. Report C683. CIRIA, London.

Coakley, J.P., and Poulton, D.J., 1991. Tracers for fine sediment transport in Humber Bay, Lake Ontario. Journal of Great Lakes Research 17(3):289‐303.

Delos – Deliverable 43, 2002. Environmental Design of Low Crested Coastal Defence Structures, Structural Design Report for LCS, EU Fifth Framework Programme 1998‐2002.

Dibajnia, M., 1995. Sheet flow transport formula extended and applied to horizontal plane problems. Coastal Engineering in Japan, Vol. 38, No. 2, pp. 179‐194.

Dibajnia, M., Moriya, T. and Watanabe, A., 2001. A representative wave model for estimation of nearshore local transport rate. Coastal Engineering Journal, Vol. 43, No.1, pp. 1‐38.

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Dibajnia, M. and Watanabe, A., 1992. Sheet flow under nonlinear waves and currents. Proceedings of 23rd International Conference on Coastal Engineering, Venice, ASCE, pp. 2015‐2028.

Environment Canada, 2004. Threats to Water Availability in Canada. National Water Research Institute, Burlington, Ontario. NWRI Scientific Assessment Report Series No. 3 and ACSD Science Assessment Series No. 1. 128 p. (http://www.ec.gc.ca/inre‐ nwri/default.asp?lang=En&n=0CD66675‐1&offset=17&toc=show accessed July 2 2014).

Gronewold, Andrew D., Clites, Anne H., Smith, Joeseph P., and Hunter, Timothy S., 2013. A dynamic graphical interface for visualizing projected, measured, and reconstructed surface water elevations on the earthʹs largest lakes. Environmental Modelling & Software, Volume 49, Pages 34‐39, November.

Hayhoe, K., VanDorn, J., Croley, T., Schlegal, N. and Wuebbles, D., 2010. Regional climate change projections for Chicago and the US Great Lakes. Journal of Great Lakes Research, Volume 36.

ISO, 2007. Actions from waves and currents on coastal structures. International Organisation for Standardisation. ISO Standard 21650:2007. 124 p.

Isobe, M., 1987. A parabolic equation model for transformation of irregular waves due to refraction, diffraction and breaking. Coastal Engineering in Japan, Vol. 30, No.1, pp. 33‐47.

Karlsson, T., 1969. Refraction of continuous ocean wave spectra, Journal of Waterways and Harbors Division. Proceedings of ASCE, Vol. 95, No. WW4, pp. 437‐448.

Lamberti, A., 1992. Example application of reliability assessment of coastal structures. Proc. Short Course on Design and Reliability of Coastal Structures, 23rd Int. Conf. on Coastal Engineering, Venice

Lewis, C.F.M., and Sly, P.G., 1971. Seismic profiling and geology of the Toronto waterfront area of Lake Ontario. Proc. 14th Conf. Great Lakes Research.

Lofgren, B.M., Quinn, F.H., Clites, A.H., Assel, R.A., Eberhardt, A.J., and Luukkonen, C.L., 2002. Evaluation of potential impacts on Great Lakes water resources based on climate scenarios of two GCMs.Journal of Great Lakes Research 28(4):537‐554.

Nishimura, H., 1988. Computation of nearshore current. In: Horikawa, K. (Editor), Nearshore Dynamics and Coastal Processes, University of Tokyo Press, Tokyo, pp. 271‐291.

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Permanent International Association of Navigation Congresess (PIANC), 1992. Analysis of Rubble Mound Breakwaters, Report of Working Group No. 12 of the Permanent Technical Committee II, Brussels.

PIANC, 2003. Breakwaters with Vertical and Inclined Concrete Walls. Report of Working Group 28 of the Maritime Navigation Commission.

Pullen, T., Allsop, N. W. H., Bruce, T., Kortenhaus, A., Schüttrumpf, H., & Van der Meer, J. W., 2007. EurOtop wave overtopping of sea defences and related structures: assessment manual.

ROM, 2002. ROM 0.2‐90 General procedure and requirements in the design of harbor and maritime structures. Spanish Ministry of Public Works and Urban Development.

Scott, D., Schwab, D., Zuzek, P., and Padala, C., 2004. Hindcasting wave conditions on the North American Great Lakes. 8th International Workshop on Wave Hindcasting and Forecasting, November

Shoreplan, 2007. Gibraltar Point Erosion Control Project Coastal Engineering Component, report prepared for Toronto and Region Conservation Authority, December.

Sumer, M. and Fredsøe,J., 2002. The Mechanics of Scour in the Marine Environment, World Scientific, New Jersey.

Toronto and Region Conservation Authority (TRCA), 2008. Environmental Study Report, Gibraltar Point Erosion Control Project, City of Toronto. February 15, 2008.

Toronto Harbour Commission (THC), 1977. Alternate Channel Study, Toronto Island Airport Study Program, March.

Toronto Harbour Commission (THC), 1983. Erosion Control Study, Toronto Island Beaches. Report prepared for Metropolitan Toronto and Region Conservation Authority.

U.S. Army Corps of Engineers (USACE), 1984. Shore Protection Manual (SPM), U.S. Army Coastal Engineering Research Center. Department of the Army.

U.S. Army Corps of Engineers (USACE), 2002. Engineering & Design ‐ Hydrographic Surveying EM 1110‐2‐1003.

U.S. Army Corps of Engineers (USACE). 2006. Coastal Engineering Manual. Engineer Manual 1110‐ 2‐1100, U.S. Army Corps of Engineers, Washington, D.C. (in 6 volumes).

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U.S. Army Corps of Engineers (USACE), 2010. Wave Information Studies Project Documentation. Coastal and Hydraulics Laboratory Engineer Research and Development Center. December, 2010. (http://wis.usace.army.mil/WIS_Documentation.shtml#po, accessed 4/2/2015).

Van der Meer, J. W., 1988. Rock slopes and gravel beaches under wave attack (Doctoral dissertation, Technische Universiteit Delft).

Vidal, C., Losada, M.A., Medina, R., Mansard, E.P.D., Gomez‐Pina, G., 1992. A universal analysis for the stability of both low‐crested and submerged breakwaters. Proc. 23rd Int. Conf. on Coastal Engineering, ASCE, Venice, Italy, pp1679‐1692.

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APPENDIX A

SITE VISIT REPORT (December, 2009)

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Gibraltar Point Site Visit Report‐ December 2009

A site visit of Gibraltar Point area on Toronto Island was completed on December 18, 2009. Figure 1 shows a recent satellite image of the area together with the points that track the route of site reconnaissance along the shoreline of the study area. Gibraltar Point has been known to be eroding for many years. Most notably to the west of the site, approaching Hanlan’s Point, erosion has historically resulted in considerable shoreline retreat of about 4 m/year. The shoreline to the east of Gibraltar Point, however, has become relatively stable in recent years, partly due to construction of groins and other structures. Erosion of Gibraltar Point itself has also slowed down in recent years after placement of some ad‐hoc shore protection.

Gibraltar Point is exposed to southwest waves but is partially sheltered by Toronto Island against east waves, despite the fact that east waves are the predominant waves in this part of Lake Ontario. It is believed that Toronto Islands were initially a series of continuously moving littoral drift deposits originating from the Scarborough Bluffs, to the east, and moving westward by wave‐ induced nearshore currents. Human interventions over the past century have resulted in loss of the sediment supply and stabilization of portions of the island.

Presently, Tommy Thompson Park (Leslie Street Spit) provides a complete barrier to transport of sediment eroded from Scarborough Bluffs. This has led to erosion of lakebed and shoreline around Gibraltar Point. Sediment provided to the littoral system by erosion of Gibraltar Point is mostly absorbed in the Hanlan’s Beach. The Western Gap jetty is located at the north end of this beach and the beach at this location has not reached its capacity yet. It is therefore, expected that Hanlan’s Beach will continue to trap the sediment eroded sediment from Gibraltar Point (if any) for many years in the future. Even when this beach reaches its capacity, the sediment is expected to be completely trapped by the Western Gap canal after bypassing the jetty. The design depth for this canal is 8.2 m. Any possible existing or future dredging needs for this canal would be an indication of sediment bypassing. Note that the Island Airport (Billy Bishop City Centre Airport) was created mostly by artificial fill and is not entirely a result of littoral transport.

The present section includes our initial observations about the site and the erosion issues. Below are descriptions for sections of the shoreline marked as Points 2 to 11 on Figure 1. Note that the tracklog (orange points) also shows the shoreline at the time of visit where it was possible to walk along the shoreline.

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Figure 1 Track logs and waypoints of the site visit

Gibraltar Point Erosion Control Appendix A Final Design 11503.101 DRAFT Baird & Associates

Points 2 and 3 There is a sandy beach behind an offshore breakwater on the west side of the Island Pier. It is called Manitou Beach (Figure 2). The breakwater was constructed in 1929 and is about 520 m long. Additional observations included: - Fine sand covers the beach with lots of pebbles - H=0.3 m, T= 5 to 6 s (visual estimation); easterly winds - A small pit 1 ft deep showed uniform mix of coarse sand and gravel underneath (Figure 3). Likely nourished material. No particular layering was observed.

- Starting where the offshore breakwater ends, there is an erosion zone downdrift (west) of Manitou Beach (Figure 4). This zone has been initially protected by gabions. The protection has been augmented by pieces of construction concrete forming an ad hoc revetment. The protected shoreline is about 200 m long. The existence of erosion zone indicates that net longshore transport at this location is towards the west. - There was a 0.5 m high scarp on top of the revetment indicating that storm waves can reach above the revetment height.

Figure 2 Manitou Beach and breakwater looking east towards island pier

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Figure 3 View of the pit dug on Manitou Beach

Figure 4 Erosion zone downdrift of Manitou Beach protected by gabions and ad hoc concrete

Gibraltar Point Erosion Control Appendix A Final Design 11503.101 DRAFT Baird & Associates

Points 4 and 5

Point 4 was near the eastward end of the fillet beach that has been formed east side of the intake structure (Figure 5). The beach was frozen at the time of visit. Wave height was about 0.6 m and period about 5 s based on visual observation. A shallow (20 cm deep) pit could be dug at this point showing thin layers of sediment (Figure 6), an indication of gradual deposition. Scattered pebbles were observed on the beach. Trace of wind‐blown sand was also observed (and wind‐blown sand traps/blocks). Fine sand is winnowed by wind and pebbles are remained.

Figure 5 View of the fillet beach east of intake structure

Point 5 was taken on the intake structure (groyne). There is also a (fillet) beach on the west side of the intake structure (Figure 6). This indicates that sediment transport at this location is bi‐modal (i.e., it has eastward and westward components). It is seen that the shoreline east of the groin at the time of visit (tracklog orange points in Figure 1) was inland of the shoreline observed in the satellite image. On the other hand, the shoreline east of Point 4 was seaward of corresponding shoreline in the satellite image. It may therefore be concluded that the shoreline in this area is relatively dynamic, moving back and forth as a result of sediment transport towards east and west under westerly and easterly storm events, respectively.

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Figure 5 View of the pit on the fillet beach with layers of sediment

Figure 6 View of the fillet beach west of intake structure

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Point 6

The beach consisted of fine sand with scattered pebbles and few fallen tree trunks (Figure 7). A small pit (1 ft deep) dug at this point showed rather uniform sand through the pit (Figure 8). The median grain size of sand based on hand lens estimate was 0.3 mm. The beach at this location thus seems to have been gradually eroding.

Figure 7 View of the beach looking towards west at Point 6

Figure 8 View of the pit dug at Point 6

Gibraltar Point Erosion Control Appendix A Final Design 11503.101 DRAFT Baird & Associates

Points 7 and 8

This beach is located east of Gibraltar Point and had a 0.6 m high scarp at the time of visit as seen in Figure 9. The scarp had a crenulate (J) shape in plan, typical of downdrift beach erosion plan shape indicating recent erosion of this beach due to a westerly event. In fact there was a sustained strong south to southwesterly event between Dec 9 and Dec 12, 2009, i.e., a week before the present site visit. Therefore the beach between Gibraltar Point (Point 9) and the intake structure (Point 5) is dynamic, moving back and forth as a result of sediment transport towards east and west under westerly and easterly storm events, respectively. The scarp featured numerous sand layers (Figure 10) indicating a historically depositional area currently being eroded. Although dynamic, the beach in this area is gradually eroding.

Figure 9 Crenulate beach scarp on the beach east side of Gibraltar Point (looking towards east)

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Figure 10 Layered sand in the beach scarp

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Point 9 – Gibraltar Point

Gibraltar Point is currently covered with a revetment. The east half (about 50 m) of the revetment protecting the washroom building seems to be well constructed. The west half of the revetment, where it gradually turns towards the north, is made of construction concrete and debris.

Figure 11 The revetment at Gibraltar Point protecting the washroom building

Figure 12 View of the west/north half of the revetment at Gibraltar Point

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Point 10

The area between Gibraltar Point (Point 9) and Hanlanʹs Point (Point 11) showed the most erosion with lots of fallen trees on intermittent beach segments. Part of the shoreline in this section is protected by a number of rectangular concrete blocks (Figure 13) likely to protect an existing infrastructure (water main etc.). There was a beach on the north side of this protection (Point 10). The beach was about 13 m wide, 70 m long, made of fine sand with scattered pebbles on top and fallen trees. The beach was frozen and was almost flat. The existence of this beach within the above erosion area is intriguing. It is possible that this beach was created by erosion of sand dunes on the back of the beach due to the strong westerly storm preceding our site visit. Figure 15 shows eroding dunes between Point 10 and Point 11.

Figure 13 Shoreline protection in the erosion area between Gibraltar and Hanlan’s points

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Figure 14 View of the beach at Point 10 (looking to the south)

Figure 15 Eroding dunes south of Hanlan’s Point

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Point 11‐ Hanlan’s Point

There is a curved wide beach at Hanlan’s Point. The area around the shoreline is covered by cobbles few cm in size (Figure 16). The beach is about 25 m wide, almost flat with lots of pebbles and backed by eroded sand dunes (Figure 17). Tree roots were visible in the dune scarp. It is likely that water levels are high during storms and reach dune toe. Hanlan’s Point marks the north end of the erosion zone. It is not clear why this point has been relatively stable. It is likely that as a result of local geology, the lakebed in this area is covered with cobbles that have protected this point from erosion.

Hanlan’s Beach is located north of Hanlanʹs Point. It is about 1.5 km long and becomes wider and wider moving towards the north (Figure 18).

Figure 16 Curved beach at Hanlan’s Point

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Figure 17 Small dunes in the back of the beach at Hanlan’s Point

Figure 18 View of Hanlan’s Beach taken from Hanlan’s Point

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APPENDIX B

LAKEBED VIDEO (TRCA, 2014)

Gibraltar Point Erosion Control Appendix B Final Design 11503.101 DRAFT Baird & Associates

LAKEBED VIDEO (TRCA, 2014)

TRCA undertook a video of the lakebed in the vicinity of the project in August 2014. The video transects are shown in Figure B‐1. Observations from the video are summarized in Table B‐1. The bottom is sandy along all the transects. A limited amount of stone was observed at various locations. A section of pipe was observed. Several typical images from the video are provided in Figure B‐2.

Figure B‐1 Transects of lakebed video (adapted from TRCA, 2014)

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Table B‐1 Summary of lakebed video observations

Transect (total video time in Observations minutes:seconds)

1 Sandy bottom, no obstructions visible. Includes close‐ups of bottom (e.g., 4:44) (11:25) Transect loop ~6:30‐7:30 min. 2 Sandy bottom. 1:37–2:00 min. loose stones visible (6:39) Remainder sand, no obstructions visible. 3 Sandy bottom, no obstructions visible. (2:39) 4 Sandy bottom, no obstructions visible. (3:25) 5 Sandy bottom ‐00:47 pipe visible (cast iron?) (3:22) ‐2.34‐3:02 armour stones/boulders, smaller stone 6a Sandy bottom, no obstructions visible ‐0:52 ? (9:06) ‐2:50 algae ‐7:34 metal beer bottle ‐Remainder sand, no obstructions visible. 6b Sandy bottom, no obstructions ‐2:15 large stone? (7:45) ‐3:41 small patch loose stone? ‐Remainder sand, no obstructions visible.

7 Sandy bottom ‐1:13‐1:28 loose stones (5:06) ‐1:38‐1:45 loose stones ‐2:18‐2:27 pipe; stones ‐2:38 stones ‐2:49 armour stone (grooved/drilled edge?) ‐3:21 stone ends ‐4:19‐4:32 stone ‐Sand bottom

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Figure B‐2 Images from video (Transect # noted in upper left corner; time noted in lower left corner; description provided in Table B‐1)

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APPENDIX C

NEARSHORE WAVES EXTREME VALUE ANALYSIS

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GIBRALTAR POINT – NEARSHORE WAVE EXTREME VALUE ANALYSIS

Data Source The wave data used in this analysis was taken from two different sources. Deepwater waves were obtained from Baird’s Lake Ontario hindcast at Station 2416. Nearshore wave conditions were extracted at six points from the Gibraltar Point HydroSed model (Table 1). HydroSed runs were used to transform the deepwater wave data to nearshore locations.

Table 1. Summary of Wave Data Extraction Points and Sources Point Water Depth [m] Source Deepwater (2416) 72 WAVAD Hindcast 38 7.0 HydroSed Model 35 6.5 HydroSed Model 33 6.5 HydroSed Model 32 6.0 HydroSed Model 16 5.0 HydroSed Model 14 5.0 HydroSed Model

The nearshore points from the HydroSed model are located adjacent to Gibraltar Point on the southwest side of the Toronto Islands (Figure 1), whereas the deepwater point is located approximately 4 km south of the Islands, in 72 m of water.

Figure 1. Location of Nearshore Points from HydroSed Model (contours shown are total depth of water with w.l. +0.6 m CD; blue=5 m and black=10 m)

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

The probability of exceedance for the deepwater point (2416) and a representative nearshore point (14) are depicted in Figure 2. The wave height at the deepwater point is substantially higher than in the nearshore zone, which can be attributed to refraction that occurs as the waves propagate to nearshore.

Figure 2. Probability of Exceedance for Point 2416 (Deepwater) and Point 14 (Nearshore)

The wave height intensity roses in Figure 3 illustrate a difference between the nearshore and deepwater points. Large waves from the east are reduced by refraction, resulting in decreased cumulative wave energy at the nearshore locations. For the nearshore zone, largest waves are from the southwest quadrant.

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Figure 3. Wave height roses for Point 2416 (Deepwater) and Point 14 (Nearshore)

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

A peak‐over‐threshold (POT) extreme value analysis was conducted to identify design wave events. Further information regarding the methodology used in this analysis can be found in Annex A.

Summary The wave height at each point for a range of return periods were then determined based on the preferred probability distributions (Table 3). Also note that the largest events from the 40 year hindcast for the deepwater point (January 10, 1977) and the nearshore points (March 5, 1964) have been included for comparison.

Table 3. Summary of Wave Heights for Given Return Periods.

Significant Wave Height [m, Hm0] Largest Point 5‐Year 10‐Year 25‐Year 50‐Year 100‐Year Event in Record Deepwater (2416) 4.66 4.98 5.39 5.68 5.98 5.58 38 (d=7.0 m) 2.90 3.19 3.62 3.97 4.34 4.51 35 (d=6.5 m) 2.86 3.13 3.53 3.86 4.20 4.31 33 (d=6.5 m) 2.83 3.09 3.46 3.77 4.10 4.18 32 (d=6.0 m) 2.79 3.03 3.39 3.67 3.98 4.01 16 (d=5.0 m) 2.61 2.76 2.95 3.10 3.25 3.23 14 (d=5.0 m) 2.56 2.68 2.83 2.95 3.06 3.10

There is a notable difference in the deepwater and nearshore wave condition. As expected, the wave height is higher in the deeper water (e.g., Pt. 38 versus Pt. 16) and the waves are slightly higher at the east end of breakwater location compared to the west end (e.g., Pt. 16 versus Pt. 14). The 100‐year wave height ranged from 3.06 to 4.34 m at the nearshore points. In some cases, the event on March 5th, 1964 met or exceeded the 100‐year event estimated by the statistical analysis.

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Annex A: Extreme Value Analysis Methodology

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

Wave data for the period from 1961‐2000 were processed in X‐Wave to conduct a peak‐over‐ threshold (POT) analysis. The parameters in Table 2 were used to define storm events used in the analysis. Table 2. Summary of Peak‐Over‐Threshold Analysis Parameters.

Min Storm Min Time Between Point Threshold [m] # Events Duration [h] Events [h] Deepwater (2416) 3.60 2 48 58 38 2.20 2 48 56 35 2.20 2 48 56 33 2.20 2 48 54 32 2.20 2 48 55 16 2.15 2 48 58 14 2.10 2 48 56

Probability of exceedance curves were then plotted to verify the suitability of the threshold values used (Annex B). The storm listing from X‐Wave was then processed in Excel where it was plotted against several different probability distributions. The Three‐Parameter Weibull distribution was then selected based on goodness of fit and visual comparison (Annex C).

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Annex B: Probability of Exceedance Curves

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Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

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Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

Annex C: Extreme Value Analysis Spreadsheet Printouts

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Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

Gibraltar Point Erosion Control Appendix C Final Design 11503.101 DRAFT Baird & Associates

APPENDIX D

EVALUATION OF POTENTIAL OFFSHORE SAND SOURCES

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EVALUATION OF POTENTIAL OFFSHORE SAND SOURCES

Gibraltar Point Erosion Control Study (Memorandum by Bair d & Associates, March 16, 2011)

Potential offshore sand sources were identified and evaluated for suitability as beach nourishment at Gibraltar Point. The screening criteria for determining suitability are summarized as follows:

Grain size – The gradation of the source sand should be similar to sand gradation at Gibraltar Point and Hanlan’s Beach, with a median grain size (D50) at least as large as the median grain size at Hanlan’s Beach (e.g., 0.2 mm to 0.4 mm). A smaller median grain size would be less stable and would increase the volume and frequency of nourishment.

Sand quality – The sand quality must meet the Ontario Ministry of Environment criteria for open water disposal (Fill Quality Guidelines for Lakefilling in Ontario).

Sufficient volume – The source must contain a sufficient volume of sand, including thickness of the sand layer, to allow practical, sustainable dredging on an annual basis with no adverse environmental impacts. The sand management option requires an average of approximately 30,000 m3 per year of clean sand.

Environmental impact – The removal of the sand must not cause adverse environmental impacts and it should be compatible with integrated shoreline management considerations.

The potential sources evaluated are as follows:

 Hanlan’s Beach  Western Gap  Centre Island East  Eastern Channel  Keating Channel  Humber Bay  Bluffer’s Park  Woodbine Beach  Ashbridge’s Bay  Offshore Eastern Beaches/Scarborough Bluffs.

The potential sources are shown in Figure 1.

Gibraltar Point Erosion Control Appendix D Final Design 11503.101 DRAFT Baird & Associates

Gibraltar Point Erosion Control Appendix D Final Design 11503.101 DRAFT Baird & Associates

Hanlan’s Beach

The sand in the nearshore of Hanlan’s Beach is limited in volume as it only extends approximately 200 m from the shoreline to a depth of about 3 m below chart datum. Evidence from aerial photographs and reports by TRCA fisheries staff indicate that beyond this depth is hard bottom comprised of till and bedrock; there is no “reservoir” of sand offshore. Further to the west, towards Ontario Place and beyond, bedrock outcrops in the nearshore and there is no accumulation of sand.

Removal of the sand from the nearshore at Hanlan’s Beach for beach nourishment at Gibraltar Point would be counterproductive for the stability of Hanlan’s Beach because the nearshore sand is part of the overall beach profile. Also, the beach profile extent is limited. Evidence from aerial photographs and reports by TRCA fisheries staff is that a very hard glacial till with bedrock exists in the nearshore. Therefore, the nearshore sand at Hanlan’s Beach is not a suitable source for beach nourishment.

“Recovery” of sand from the nearshore profile that had been originally placed through some initial nourishment (e.g., see Western Gap) would require skimming a thin layer annually from the nearshore area. Recycling the sand in this manner would reduce the present ongoing advance of the beach‐dune system.

Western Gap

The Western Gap, a navigation channel into Toronto Harbour, is located at the northwest end of Hanlan’s Beach. As previously discussed, the net sediment transport direction in this area is northwest along Hanlan’s Beach. Public Works Canada redredged the approach to the Western Gap in 1972 and at the same time a sink was created by dredging to ‐8.8 m chart datum to a line 22 metres south of the south wall of the channel produced (THC, 1983). By 1982, the sink had filled with 36,000 m3 of sediment and sedimentation was again progressing into the channel. Recently, the Toronto Port Authority (TPA) estimated that approximately 2,000 m3 of sediment deposits in the Gap each year (TRCA, 2008). This Baird report estimates the potential rate of sediment infilling into the Gap as 3,000 m3 per year on average. The TPA presently does not maintain the Western Gap at seaway navigation depth and dredging is infrequent. In 2004, TPA estimated that approximately 61,000 m3 of sediment needed to be dredged from the Western Gap (TRCA, 2008); only 3,360 m3 of sediment was dredged in 2006 and disposed of at Tommy Thompson Park (TRCA, 2008). Discussions between TRCA and TPA indicate a mutual preference for disposing of clean sand dredgeate from the Western Gap at Gibraltar Point (TRCA, 2008). It should be noted that TRCA has no jurisdiction to direct the disposal of sediments at Gibraltar Point or any other location.

The present accumulation within the channel at the Western Gap is insufficient to sustain ongoing sand management at Gibraltar Point; it would initially provide only about 2 years of supply for the

Gibraltar Point Erosion Control Appendix D Final Design 11503.101 DRAFT Baird & Associates sand management option and subsequently only about ten percent of the annual volume required. However, sand dredged from the Western Gap could be sufficient to supply the breakwater and focused sand management option at Gibraltar Point. It is expected that the sand would be of suitable size and quality. Testing would be required.

Centre Island East

There is an unknown volume of sand in the nearshore area off the easterly portion of Centre Island. However, it is not expected that a sufficient volume exists to provide the required annual quantity for the sand management option at Gibraltar Point without adversely impacting the shoreline in this area. Minimal additional sand is being transported into this area due to the sheltering effect of the Leslie Street Spit headland.

In 1974 approximately 90,000 m3 of dredged sand was placed to create a 75 m wide beach extending from the bridge at Algonquin Island to Ward’s Island beach (Baird, 1994). The subaerial part of the placed sand was quickly eroded and deposited against the west pier of the Eastern Channel which had been extended and realigned in 1974. The realigned beach reached equilibrium and any additional sediment moving into the area freely bypasses the west jetty and adds to the shoal accumulation in the Eastern Channel (see below). No dredging was undertaken by April 1983 (THC, 1983). According to the 1994 Toronto Islands Shoreline Management Study Report of Coastal Processes (Baird, 1994a), the Eastern Channel was dredged in 1983, which was the first time since 1965. The dredge volume in 1985 was 28,500 m3 and it was dumped offshore.

Eastern Channel

Sediment accumulation in the Eastern Channel decreased dramatically after construction of the Leslie Street Spit commenced in 1965. The Leslie Street Spit blocked the easterly waves as well as the sediment from the northeast. Baird (1994) reported that the Eastern Channel was last dredged in 1983, the first time since 1965, and 28,500 m3 was removed and dumped offshore. Since that time, a shoal off the south end of the west jetty extending into the Eastern Channel appears to be accumulating at a rate of 5,000 to 6,000 m3/yr. This volume from the Eastern Channel is insufficient to supply the sand management option at Gibraltar Point on a sustainable basis but would be sufficient to supply the breakwater and focused sand management option. It is expected that the sand would be of suitable size and quality; testing would be required.

Keating Channel

Currently, sediment is dredged from the Keating Channel at a rate of approximately 30,000 m3 per year. Based on samples from 2006, it was estimated that about 75% of the Keating Channel sediment was finer than 0.125 mm (including about 50% silt and clay). This sediment is too fine for beach nourishment at Gibraltar Point as the existing surficial sand at Gibraltar Point has only about 10% finer than 0.125 mm. Historically, the quality of the Keating Channel sediment does not meet

Gibraltar Point Erosion Control Appendix D Final Design 11503.101 DRAFT Baird & Associates

MOE open lakefill guidelines (i.e., exceeds lowest effect level threshold), although continued sampling indicates that the contaminant level sediment quality is improving. Technologies proposed for the Don Mouth Naturalization and Port Lands Flood Protection Project may allow for the reuse of some of the sediment in the future. If the sediment could be cleaned, it would also have to be processed to provide a suitable grain size distribution for beach nourishment.

Humber Bay

Sediment in Humber Bay consists mainly of silt and clay which are largely the result of deposition from the Humber River (Rukavina, 1969). This sediment is not suitable for beach nourishment at Gibraltar Point as it is too fine (i.e., grain size smaller than required).

Bluffer’s Park, Eastern Beaches and Woodbine Beach and Ashbridge’s Bay

The Scarborough Bluffs, Bluffer’s Park, Woodbine Beach and Ashbridge’s Bay in eastern Toronto are located within a littoral cell which stretches from East Point to the Leslie Street Spit. A littoral cell is a closed system (i.e., no sediment is transported into or out of the cell). It is important to understand that any actions which alter the sediment supply or transport processes within a littoral cell may affect downdrift shorelines within that cell.

The wave‐induced net longshore sediment transport direction is east from East Point to west at Ashbridge’s Bay. Historically, the source of the sediment has been erosion of the Scarborough Bluffs and the nearshore. Prior to the construction of the Leslie Street Spit the sediment moved along the shore to the Toronto Islands; construction of the Spit, stopped the movement of sediment to the Islands. Construction of the headlands at Ashbridge’s Bay resulted in the longshore sediment becoming trapped at Woodbine Beach.

Following construction of the Bluffer’s Park headlands in the mid 1970’s, a fillet beach accreted on the updrift (east) side of the Park; the fillet beach has evolved to the point that sediment from the east bypasses the headland. Shoreline protection of the bluffs updrift of Bluffer’s Park has reduced the sediment supply. The fillet beach at Bluffer’s Park provides erosion protection to the bluff and is also a valuable recreational resource. Therefore, the beach at Bluffer’s Park is not a suitable source for sand for the sand management option at Gibraltar Point.

West from Bluffer’s Park to the RC Harris Water Treatment Plant (WTP) the supply of sediment from erosion of the bluffs has stopped because the shoreline has been fully protected; erosion of the nearshore continues. A recent update (Shoreplan 2010) of earlier estimates (Atria, 1993; Sandwell, 1991; Philpott, 1988) indicates that the sediment supply reaching RC Harris WTP is approximately 15,000 to 20,000 m3 annually. The Eastern Beaches, extending from RC Harris WTP to Woodbine Beach, provide a valuable recreational resource and shoreline protection and are dependent on the sediment supply from updrift of RC Harris WTP. The diminishing sediment supply has put erosion stress on the Beaches requiring numerous intervention structures and management.

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Woodbine Beach is a fillet beach that has accumulated on the updrift side of the Ashbridge’s Bay headlands at the westerly end of the Eastern Beaches. The fillet beach has evolved to the point that sediment bypassing occurs at the Ashbridge’s Bay headland. Approximately 15,000 m3/year of sand can potentially bypass the Ashbridge’s Bay headland where it then deposits in Ashbridge’s Bay and Coatsworth Cut (Baird, 2011; Shoreplan, 2010). Bypassing of the Ashbridge’s Bay headland may increase over time as the shoal increases. However, this may be limited by the decreasing supply from updrfit. Negligible sediment is transported northwards along Leslie Street Spit since the Spit itself is armoured with concrete rubble and the shoreline profiles are relatively steep. Estimates are that approximately 4,000 m3 per year, on average, of sediment is dredged from Coatsworth Cut.

Removal of sand at Ashbridge’s Bay from the littoral cell to Gibraltar Point would be contrary to the recommended integrated shoreline management approach for the Eastern Beaches (e.g., Fenco MacLaren, 1996; Sandwell 1991) which advocates “recycling” sand from Ashbridge’s Bay back to the updrift end at RC Harris WTP. Also, the median grain size of the sediment accumulating in Ashbridge’s Bay is typically 0.12 mm to 0.18 mm (Baird, 1999) which is finer than the material required for Gibraltar Point.

Offshore Eastern Beaches and Scarborough Bluffs

A reserve of sand on the lakebed of Lake Ontario offshore of the Eastern Beaches and the Scarborough Bluffs has been identified (B.A.R., 1994; Atria, 1993). A commercial proposal was initiated to dredge sand from this reserve in an area of approximately 14 km2, 1.5 to 4 km offshore Ashbridge’s Bay to Bluffer’s Park, in depths ranging from 10 m to 40 m below chart datum. The substrate was estimated to be fine to coarse sand in a layer 1.5 m to 5 m thick (B.A.R. 1994). Based on sediments samples collected from within the proposed dredging area (B.A.R., 1994), the median grain sizes varied from 0.18 mm to 9.1 mm (Atria 1994). The larger values indicate a possible gravel lag deposit where the finer materials have been moved downdrift. The finer sediments with a median size of about 0.2 mm were typically distributed over the southwest portion of the site, while the coarser materials were located at the middle of the site. Due to the depth of the sediment, initial preliminary study (Atria 1994) indicated that it may be beyond the primary active nearshore zone, yet it may still serve as a source for the Eastern Beaches.

Offshore sand resources belong to the province. Any proposal for dredging would require detailed field investigations and analysis and would require an environmental assessment, including evaluation of the fisheries, the benthic community, shoreline processes and increased turbidity at the City of Toronto water intake pipes.

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Conclusion

Potential offshore sand sources were identified and evaluated for suitability as beach nourishment at Gibraltar Point. All the sources are inappropriate for use at Gibraltar Point due to physical limitations and incompatibilities with other shoreline management strategies. The present sand volume accumulated within the channel at the Western Gap is insufficient to sustain an ongoing sand management program at Gibraltar Point. Sources at Hanlan’s Beach, Centre Island East and the Eastern Channel are not suitable due to insufficient volumes and potential adverse impacts to the shoreline. Sediment dredged from the Keating Channel is not of suitable grain size and does not meet provincial lakefill quality guidelines. Sediment in Humber Bay is too fine for use at Gibraltar Point. Sources at Bluffer’s Park, Eastern Beaches and Woodbine Beach are not suitable because removal of sand from these areas would result in adverse impacts to these vital beaches. Sediment accumulating at Ashbridge’s Bay is typically too fine for use at Gibraltar Point. In addition, removal of sand from Ashbridge’s Bay to Gibraltar Point would be contrary to the recommended integrated shoreline management approach for the Eastern Beaches. A potential source in deeper water offshore of the Eastern Beaches and the Scarborough Bluffs would require detailed analysis and environmental assessment and is not presently available for aggregate extraction.

In summary, local nearshore or offshore sand sources do not have sufficient volume or suitable grain size and quality for ongoing, sustainable sand management at Gibraltar Point. Imported sand is required for the pure sand management option.

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APPENDIX E

WESTERN BEACHES BREAKWATER PHYSICAL MODEL

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Western Beaches Breakwater Physical Model (Baird & Associates)

A series of physical model tests was undertaken by Baird in 2005 to inform the design of the Western Beaches breakwater. The wave heights at the Western Beaches breakwater are estimated to be higher than the wave heights at the Gibraltar Point breakwater, in part due to the fact that Western Beaches breakwater is in deeper water (‐ 6 m CD) than the proposed Gibraltar Point breakwater (‐4.4 m CD, with allowance for lakebed lowering). The deeper water allows higher wave to reach the Western Beaches breakwater. The model scale was 1:20.

Two alternatives were tested as part of the final test series (Test Series C):  1a, consisting of 4 layers of 1‐6 tonne stone on the front slope  1b, consisting of 3 layers of 1‐6 tonne armour on the front slope.

Both alternatives used 2 layers of 4‐6 tonne on the crest and rear slope. The model was tested at an equivalent crest elevation of +2.5 m CD and a crest width of 7 m. The test section for 1a is shown in Figure 1. The model section prior to testing for 1b is shown on Figure 2. Test Series C was undertaken with stone with a specific gravity of 2.9.

The test section was subject to series of wave and water level conditions. The test condition similar to the design conditions for Gibraltar Point was as follows: H = 3.8 m, T = 10 s Water level = +1.7 m CD; Freeboard = 0.8 m.

The model section after these wave conditions is shown in Figure 3. Figure 4 is during the test.

Figure 1 Breakwater test section, Test Series C (Alternative 1a shown)

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Figure 2 Test Series C1b prior to testing with 3‐layers of 1‐6 tonne stone (Top view of breakwater section prior to testing. 3 layers, 1‐6 tonne armour stone on front slope, painted white; the blue represents an area of the front slope over a different gradation of filter material; 2 layers of 4‐6 tonne armour stone on the rear crest and back slope painted yellow; the brown area below yellow slope is the lower part of inner slope (core stone))

Figure 3 Test Series C1b after testing with 3‐layers of 1‐6 tonne stone (water level = +1.7 m; freeboard = 0.8 m; Hs = 3.8 m; T = 10s)

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Figure 4 Hydraulic model during Test C1b (water level = +1.7 m CD; freeboard = 0.8 m; Hs = 3.8 m; T = 10 s)

The model testing showed that stone movement occurs under the design storm conditions (see Figure 3). The front slope damage is characterized by some downslope displacement of armour (from upper to lower slope), as well as a few stones being thrown up onto the crest. The upper slope has been reduced to two layers in some areas. The damage level is “tolerable”. The filter layer was not exposed. Failure of the structure did not occur. Movement of the stone can cause breakage. Significant sliding of stones at the toe of the front slope was not observed. Rear slope damage is localized, with one area where several stones have been removed, leaving a single layer of armour. The other rear slope issue is erosion of the rear core bench, which has caused some slumping of the rear filter bench. However, this has not undermined the rear slope armour. Note that the Gibraltar Point breakwater rear slope armouring continues to the toe of the slope.

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