New Frontiers in the Offshore Geotechnics and Design

Morteza Sheshpari *, Saman Khalilzad University of Ottawa, Department of , CBY Building, 161 Louis- Pasteur, Ottawa, ON, Canada, K1N 6N5. *e-mail: [email protected]

ABSTRACT and development of the offshore structures are growing because of the demand for the offshore oil and gas and developing interest in the offshore wind farms. The offshore is an applied science that covers related problems in the offshore environment. Generally, the offshore industry is very innovative with -funded research that its results are rapidly and cautiously used in the practice. Different site investigation and design methods are required for each sea; for example highly overconsolidated clays and compacted in the North Sea, carbonate sediments in Australian the offshore, and in Brazil and India; ultra-high plasticity soft clays in West Africa, each requires individual approach for field investigation and foundation design. Geophysical explorations for the offshore projects can obtain some essential data for site investigations, foundation design, and construction process. Therefore, some geophysical methods in the offshore site investigations are discussed within this paper. Two modes for the geotechnical investigations in the offshore environments are predictable, including the seabed mode, and drilling mode, which more details about them are presented in the paper. There are different types of offshore oil and gas, and wind turbine structures, which are used based on project type, depth, and economical factors. Different foundation types and design methods for the jackup, jacket, gravity base, and floated oil and gas structures, and the offshore wind turbines are discussed further. It was concluded that appropriate foundation from the available types considering the depth of water, site conditions, applied loads, purpose of the structure and the potential geohazards should be selected and designed for the offshore structures. It can be inferred that application of in-situ data and design principles of traditional offshore foundation should be adjusted for new types of foundation regarding the geometry, applied load, geohazards, and used cautiously. Different approach and criteria should be considered for the offshore wind turbine foundations than the oil and gas platform foundations due to the difference in the applied load and the foundation geometry. The application of numerical analyses combined with the analytical methods for better and accurate results and understanding in the foundation design and geohazard assessment should be considered as a necessary approach. KEYWORDS: Offshore, Geotechnics, Foundation Design

INTRODUCTION After second half of the 20th century, the oceans and their sediments are recognized to be a great source of mineral assets in the form of hard minerals, petroleum, and gas (Gerwick, 2007). The offshore structures construction and development are growing because of demand for the offshore oil

- 1 - Vol. 21 [2016], Bund. 1 2 and gas and developing interest in the offshore wind farms. All of the offshore structures have chance to experience geohazards. Hence, their foundations should tolerate structural weight and applied loads from the surrounding environment (Dean 2010). The need for more oil and gas increased hydrocarbon exploration, production, in new regions such as Texas and the Middle East shallow offshore , the North Sea, and the offshore of Australasia and the Far East, South America, India and newly West Africa (Randolph et al., 2005). The offshore oil and gas now provide approximately one third of the world’s energy demand (Gerwick, 2007). The types of the offshore structures changed from fixed steel or platforms, to floating structures. New developments have resulted in different innovative anchoring systems, each developing under the influence of variable loading regimes applied on them. The increase in the depth of waters where oil and gas platforms should be installed from under 200 m in the 1980s to 2000 m and more now, has triggered great investment deep water anchoring and mooring system for floated plat forms (Randolph et al., 2005). Designing the offshore geotechnical engineering structures evolved from onshore practice, with areas of the foundation scale and installation methods separated from the original onshore principles during the last 30 years (Randolph et al., 2005). Generally, the offshore industry is very innovative with well funded research that its results are rapidly and cautiously used in practice (Dean 2010). The offshore geotechnical engineering is an applied science that covers related problems in the offshore environment. The offshore industry progresses in the geotechnical aspects can be described as a few very large diameter piles groups are being used instead of many moderate-sized piles; deep skirts are used instead of the excavation of shallow soft sediments. They can change the effective foundation depth to the skirts tip levels penetrated several diameters into the seabed. Underwater embedment has increased the application of suction (or underpressure) skirted or caisson foundation installations. Special attention to the effects of cyclic loading on the offshore structures has been taken. Special design codes have been developed for proper design of the offshore structures (Randolph et al., 2005). The offshore wind energy is getting a more popular source of renewable energy, and it has been forecasted 16 GW of the offshore wind turbines to be installed by the end of 2014, and global total of 75 GW by 2020 (Madsen et al., 2012). The offshore wind turbines foundation approximately cost for 1/3 of the total cost of the offshore wind farm budget and in deep water (>25m), its cost accounts for 50% of the total project (Madsen et al., 2012). Almost, 75% of all wind parks today, use monopiles as their foundation. Monopiles are simply made structure with welded steel pile, and require no preparations in the seabed. Negative aspects of the monopiles are their requirement for heavy duty piling-drilling equipment, and not being appropriate for the seabeds with many large boulders (Madsen et al., 2012). Jackets and tripod which are appropriate for deeper waters, require minimum preparation in the installation site before embedment. Another type of the offshore wind turbines foundation is the bucket foundation made from thin shell structure. At waters with more depth, the diameter of the bucket should increase, but large aspect ratio between caisson diameter and wall thickness cause critical buckling instability in them during installation (Madsen et al., 2012). In this paper, site investigation methods for the offshore structures especially, oil and gas platforms, and wind turbines are reviewed and discussed. Then different foundation types, including traditional and new ones for different oil and gas platforms and wind turbines are presented and discussed in more detail. Foundation design methods for different types of foundation in the offshore structures are reviewed and discussed. At the end a general conclusion comparing new trends with traditional ones in the geotechnical engineering aspects of the offshore structures is presented.

Site investigation methods in the offshore environments and their new trends Development of oil platforms and wind turbines for energy needs in the offshore environment in recent years proved that different site investigation methods are required for each sea. For example, Vol. 21 [2016], Bund. 1 3 highly overconsolidated clays and compacted sands in the North Sea, carbonate sediments in Australian offshore, and in Brazil and India; ultra-high plasticity soft clays in West Africa, each requires individual approach for field investigation and design (Randolph et al. 2005). Costs of the site investigation for using special vessels which can be typically $250,000 to $500,000 per day, in addition to similar costs for investigation operations are most important factors in the offshore site investigation (Randolph et al. 2005). Site investigation procedure for the offshore environments, generally such as other projects include desk study, geophysical exploration, geotechnical investigation, taking samples to the laboratory or performing in-situ tests.

Geophysical surveys and explorations in the offshore environments Geophysical explorations for the offshore projects can obtain some essential data for the geotechnical investigations and foundation design and construction. Usually, the information is collected by the combination of different techniques. For example, the seabed bathymetry and topography can be detected by echosounding or swath echosounding (especially in locations with uneven seabed, outcrops, corals, pockmarks, waves, etc.), using acoustic (sound) energy and its reflection from the seabed or geological layer (Danson 2005; Dean 2010). Bathymetric surveys should be performed over an area of approximately 1 km square around the proposed structure location. Spacing of survey lines should not be usually more than 100 metres by 250 metres (Jardine 2009). The seabed features and obstructions can be studied by sidescan sonar (in traditional method) (Danson 2005; Jardine 2009; Dean 2010), or by modern mosaicing systems, which create an acoustic photograph of the seabed (Danson 2005). For detection of stratification and layering status of under the seabed, sub-bottom profilers such as, pingers, boomers are used (Danson 2005; Jardine 2009). Pingers are good in the detecting shallow geological features such as gas accumulations, relict channels and faults in the depths of less than 50 meters in the seabed while boomers with higher frequency can penetrate to a depth of 100 m in the seabed with resolution of 0.3 to 1 m. New technology, which is designed to replace pingers and boomers, is CHIRP that can operate around central frequency (3 kHz-40kHZ) and provide better resolution from sediments close to the seabed (Danson 2005). LiDAR which is an air-borne laser scanning system is a new technology which is used in shallow (Usually less than 30 m to 50 m deep) and clear waters (waters containing low concentrations of suspended particles) to measure the depth of water and location of the seabed. Two laser frequencies are used that their difference shows water depth. The first one is a red-light laser and is reflected from the sea surface, and second one is green light laser that is used for penetration to the seabed. For vast area, LiDAR provides measurements in the lower cost with same accuracy compared to the multi-beam (Danson 2005). The seabed surface surveys and mapping become out-of-date due to construction and drilling activity or sediments mobilization and should be redone after six months (Jardine 2009). For detection of metallic items, such as pipeline and cable crossings, metallic debris and ammunition located on the seabed or beneath it, magnetometers are used (Danson 2005; Jardine 2009). To find information about geological data, reflection seismic systems by digital data and penetration to 50 m for geotechnical determinations are used. In shallow layers (<10 m depth), and shallow-water depths (<15 m) seismic refraction methods are used, due to limitations that low frequency seismic reflection methods encounter at these depths (Danson 2005; Dean 2010). For the detection of hard layers relative to the soft layers (sand, , ) electrical resistivity surveys are used in the offshore geophysical explorations (Danson 2005). Traditional sledge-based seismic systems apply pressure waves (P-Waves) to the seabed, but the new system presented by Puech et al. (2004) which applies surface waves to the seabed, and another new system described at Vanneste et al. (2007) produce and detect the sledgemounted shear waves. Application of ground-penetrating radar survey (GPR) a radar system which sends electromagnetic waves into the seabed and measures reflection for creating a picture of sub-bottom layers has been described by Eyles and Meulendyk Vol. 21 [2016], Bund. 1 4 (2008) as a new technology in the offshore environments. Remotely Operated Vehicles (ROV), and a newer technology called Autonomous Underwater Vehicles (AUV) for sub-bottom investigation are geophysical devices or platforms (Fig. 1) that can be controlled remotely for exploration purposes (Danson 2005; Dean 2010).

Figure 1: Sea Demon ROV (Left), and AUV on a survey vessel (Right) (Danson 2005)

GEOTECHNICAL INVESTIGATION IN THE OFFSHORE ENVIRONMENT

General information about the in-situ sampling and testing in the offshore environments Two modes for the geotechnical investigations in the offshore environments are predictable, including the seabed mode, and the drilling mode. In the seabed mode where the sampler or in-situ testing device is positioned directly from the seabed, penetration depths are usually between 20 m and 60 m. In the drilling mode where the device is launched from the bottom of a , penetration can be performed in greater depths using platforms or vessels (Danson 2005; Randolph et al. 2005). Piston samplers are used in the soft to stiff clays, and if impractical thin walled push-samplers should be used. In the dense sand, hammer samplers, and in the stiffer environments such as boulder clays and cemented soils, rock coring methods can be employed (Danson 2005). Most common in-situ testing techniques are piezo cone penetration tests P-CPT) with measurement of the cone and sleeve resistance and in addition to the methods such as field vane shear test (VST), T- bar and ball probe tests in soft clays (Fig. 2) (Danson 2005; Randolph et al. 2005; White et al. 2010). Various drillings, sampling and coring systems can reach to the various depths of water for sampling, and more information can be found in table 1,2, and table 3 showing sample quality and recovery in the seabed and down-hole sampling devices (Danson 2005). Table 1: Penetration depths of the drilling, sampling, and coring systems and water depths (Danson 2005) Grab Sampler Unlimited** 0.1 m to 0.5 m (mechanical) Equipment Maximum Penetration Description Water Depth (m)* (m)* Drill mode Unlimited** Unlimited** borings from Vol. 21 [2016], Bund. 1 5 vessels

Rock corer 200 m 2 m to 6 m (the seabed unit) PROD™ the 20 m to 2,000 2 m to 100 m seabed m drilling/coring Basic gravity Unlimited** 1 m to 8 m corer Piston corer Unlimited** 3 m to 30 m Vibrocorer 1000 m 3 m to 8 m Box Corer Unlimited** 0.3 m to 0.5 m the seabed Push- 250 m 1 m to 2 m in Sampler Grab Sampler Unlimited** 0.1 m to 0.5 m (mechanical) Grab Sampler 200 m 0.3 m to 0.5 m (hydraulic) These figures should be used for general guidance only. ** Water depth is limited by the deployment winch and handling depths of drilling, sampling, and coring systems capabilities and water depths (Danson 2005). Table 2: In-situ testing penetration depth in different equipment and associated with various types of equipment and water depth (Danson 2005). Equipment Description Water Depth (m)* Penetration (m)*

Deck- or frame-operated penetration 20 m 2 m to 60 m tests the seabed wheeldrive penetration 500 m to 3000 m 2 m to 60 m tests Drilling mode downhole penetration Unlimited** Unlimited** tests PROD™ the seabed penetration tests 20 m to 2,000 m 2 m to 100 m Light weight wheeldrive penetration 2000 m 2 m to 5 m tests ROV penetration tests 300 m to 2000 m 1 m to 2 m Minicone penetration tests 250 m to 2500 m 5 m to 6 m the seabed vane test 250 m to 2500 m 5 m to 25 m * These figures should be used for general guidance only. ** Water depth is limited by the deployment winch and handling capabilities.

Table 3: Sample quality and recovery in different equipment (Danson 2005). the seabed sampling equipment Type of Equipment Sample Quality Recovery (relative to length of sample tube) sand clay rock sand clay rock Gravity / Piston corer 2 3 1 1 3 to 4 1 Vibrocorer 2 to 3 2 to 3 1 3 to 4 2 to 3 1 Grab sampler 1 to 2 1 1 1 to 2 2 1 Vol. 21 [2016], Bund. 1 6 Box corer 1 to 2 5 1 1 5 1 Rotary corer 1 2 3 to 4 1 3 3 to 4 DOWN-HOLE SAMPLING EQUIPMENT Type of equipment Sample quality Recovery (relative to length of sample tube sand clay rock sand clay rock Hydraulic piston sampler 3 to 4 5 1 3 5 1 Hydraulic push sampler 3 to 4 4 to 5 1 3 5 1 Hammer sampler 2 to 3 2 to 3 1 3 to 4 3 to 4 1 Rotary coring 1 2 3 to 4 1 3 3 to 4 1: Poor or inappropriate 2: Acceptable for non- 3: Moderately good 4: Good 5: Very good critical analyses

In-situ testing in the offshore environment (VST), T-bar and ball probe tests in the soft clays are used in the offshore environments (Danson 2005; Dean 2010; As explained above common in-situ testing techniques such as Randolph et al. 2005; White et al. 2010). the piezo cone penetration tests (P-CPT), vane shear test

Figure 2: Different penetrometers at the top left (from Randolph et al. 2005); Wheel-drive penetration test unit at the top middle; ROV mounted CPT unit at the top right; SEASCOUT with minicone system at the bottom left; Stand-alone vane test system at the bottom middle; Halibut vane test rig during recovery at the bottom right (from Danson 2005) .

CPT: The most common test in marine engineering investigations is (CPT) with a history of 4 decades. Primary objectives are set to find soil type, layering, Vol. 21 [2016], Bund. 1 7 undrained in clayey soils, and relative density and internal angle of sandy layers, which are required parameters for foundation design (Danson 2005; Dean 2010; Jardine 2009; Lunne 2012; Randolph et al. 2005). CPT can be applied in many different offshore projects such as described by Danson (2005): a-Investigation for the offshore oil & gas pipeline route, submarine cable route studies and burial evaluation, trenching activity and the slope stability studies. b- Geotechnical research of the seabed structures and anchors, or nearshore structures, or inshore structures. Pre-dredge studies and finding ground truth in geophysical exploration and morphological mapping. Wheel-drive penetration test unit made from a the seabed reaction frame, the wheel-drive mechanism, electronic control and data acquisition systems are suitable for the seabed in-situ testing. When a CPT should be done at an accurate spot in deep water, or when a series of uninterrupted in-situ tests in the linear format are required, such as a cable route or pipeline investigations, the ROV (Remotely operated vehicle) penetration test unit should be employed (Danson 2005). Minicones were developed for better stratigraphic outlining and soil parameter detection from vessels. They were primarily used in pipeline and cable route investigations but have been developed to the small subsea structures, anchors and dredging investigations (Danson 2005). Geotechnical investigation is necessary for the foundation design of offshore wind turbines. According to Gavin et al. (2012), almost up to 50% of the cost for development of the offshore wind farm belongs to the foundation cost, and Madsen et al. (2012) estimated that 46% of the cost in deep water projects belongs to the foundations and installation. CPT test is used as an in-situ test for site investigation in the wind turbine pile foundations. There are different corellations between CPT data and design requirement for the offshore pile foundation design. The non-easy procedure for the determination of the operational earth pressure coefficient K, and natural change in the seabed sand deposits, which create a problem in the estimation of shaft resistance have led to the development and application of correlations between design and in-situ tests parameters. Similarity between the (CPT) penetrometer installation and pile installation increased the usage of correlated parameters between CPT testing and pile foundation design in the offshore wind turbines (Gavin et al. 2011). Other sensors that can be accompanied with CPT device and detect other parameters are thermal conductivity probe, Electrical conductivity cone, Seismic cone, natural gamma, and dilatometer.

VST The vane shear test can be implemented in the , from wheel-drive machines or standalone tested rigs. VST is a fast and precise device for detecting the in-situ undrained shear strength of cohesive soils, such as soft clays (Danson 2005; Lune et al. 2011; Randolph et al. 2005). Vol. 21 [2016], Bund. 1 8 T-bar and ball penetration test The T-bar and ball penetration tests can be compared with the CPT but they differ in giving more accurate evaluation of the shear strength in very soft soils. The T-bar should be used only by the seabed penetration system, while the ball penetrometer can be used from the seabed or borehole based systems. An empirical evaluation from the resistance of the seabed soils during the T-bar or ball probe penetration into the seabed by pushing can be obtained. The T-bar made from a short rod-shaped bar attached vertically to the penetrometer main rods. While it is pushed into the seabed, a load cell located behind the bar measures the resistance. It gives greater confidence in the determining shear strength compared to (P)CPTs. The ball penetration device has same mechanism of the T-bar, but a ball replaces the bar (Danson 2005; Lune et al. 2011; Randolph et al. 2005).

The cyclic test of penetration Usually, it is common to monitor the extraction resistance of the T-bar or ball penetrometers, which can be developed to standard cone penetration tests too. By measurement of the extraction resistance, after correction applied to the overburden stress, a net negative extraction resistance is calculated (Randolph et al. 2005). Figure 3, shows common ratios of Tbar extraction resistance relative to the penetration resistance. A cycle of single penetration and extraction reduces the strength considerably, and softening to a total remoulded state needs 5 to 10 more cycles.

Figure 3: Variation in ratio of extraction to the penetration resistance of T-bar (Randolph et al. 2005).

Vol. 21 [2016], Bund. 1 9 Discussion about in-situ sampling and testing in the offshore environment and new technologies for in-situ testing As it was presented, two main modern in-situ testing systems in the offshore geotechnical investigations are downhole system, and the seabed system. In the modern downhole testing systems, the instrument (CPT, VST) falls down freely along the drill pipe and reach to the base, and then is jacked into the soil hydraulically. Test data are saved in the vicinity and then recovered when the whole system is taken out by the wireline attached to the top of the device later (Randolph et al. 2005). Similar downhole system is used by Fugro, which is called XP system (Hawkins & Markus, 1998). The seabed systems are being used increasingly in the deep water hydrocarbon projects, where anchored-floating facilities are constructed and the seabed sediments are soft. Penetration into the sea bed, while penetrometer bars are submerged under the tension load above the seabed, is done by a continuous push through wheel drive unit in the modern seabed systems (Peuchen 2000). Comparison of the downhole systems and the seabed systems show that the seabed systems despite having lower depth penetration power of 30-40 m, which is enough for anchoring platforms, has better quality in acquired data compared to the downhole systems that can penetrate into deeper layers (150 m or more). The sampling depth of the seabed systems is limited to the first 2-3 m depth of the seabed due to limitation in the drilling power (Randolph et al. 2005). To overcome this obstacle, another device, Remotely Operated Drill (PROD) (Fig. 3), was developed by Benthic Geotech, which combines properties of the penetration and taking samples in deeper layers (Carter et al. 1999). It can operate up to 2000 m water depth, and can drill and take samples down to 125 m into the seabed, and its CPT device can penetrate to depth of 100 m in the seabed as a robotic system (Carter et al. 1999). Two other types of the remote seabed sampler e.g. (Jumbo Piston Corer or JPC), and (STACOR sampler) have been presented by Young et al. (2000), and Borel et al. (2002 ) respectively, which can produce 20-30 m long samples. A database from historical development of the seabed rigs, including the recent ones presented by Lunne.

Laboratory testing In the offshore projects, obtained soil samples can be tested in the laboratories on the vessel or on-shore. The laboratory tests can be just a simple grain size analysis to more complicated standard tests according to the project needs and design parameters. Explanation of laboratory tests is out of this paper length and purpose.

FOUNDATION TYPES FOR THE OFFSHORE OIL & GAS PLATFORMS AND WIND TURBINES AND THEIR NEW TRENDS There are different types of offshore structure for oil and gas, and wind turbine projects, which are used based on the project type, water depth, and economic factors. Figure 5 and 6 shows different oil and gas offshore structures, and figure 7 shows different types of wind turbine structures. Vol. 21 [2016], Bund. 1 10

Figure 4: Development of the seabed rigs including the recent ones (Lunne 2012).

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Figure 5: Different types of oil and gas offshore structures (Image from Oil and Gas Journal).

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Figure 6: The offshore oil and gas structure types including: 1, 2) conventional fixed platforms; 3) compliant tower; 4, 5) vertically moored tension leg and mini-tension leg platform; 6) spar; 7,8) semi-submersibles; 9) floating production, storage, and off-loading facility; 10) sub-sea completion and tie-back to host facility (NOAA: Ocean Explorer, 2010).

Figure 7: Foundation types used in the offshore to support wind turbine structures (Gavin et al. 2011).

As it can be seen from Figures of 5, 6, and 7, some of the offshore structures in less deep waters are fixed to the seabed with foundation structure, while other ones float over sea and are fixed to the seabed by anchors. In this paper due to size limitation and purpose of the paper, fixed offshore structures with foundation are discussed with details while short descriptions are provided for anchored structures. Vol. 21 [2016], Bund. 1 13

JACKUP PLATFORMS A jackup is a mobile offshore platform that moves to the required location and made from a hull that holds drilling and other equipment on the top section, and three or more telescopic legs passing through the hull (McClelland et al., 1982; Young et al., 1984; Vazquez et al., 2005). A unit moves onto the location, sets its legs onto the seabed, and raises its hull out of the water. A large jackup (Figure 8) can be used in up to about 150m sea water depth. This unit has legs composed of a frame structure 10m2 or similar in plan view, installed on independent foundations, which are called spudcans, with the possible 20m diameter (Dean 2010). When some soft soils are present and large foundation is required, a mat foundation which also can be used on sandy seabed is employed (Turner et al. 1987; Murff and Young 2008; Templeton 2008). Different types of foundation that can be used with the independent legged jackups are shown in figure 9.

Figure 8: Large jackup platform and its footing (Dean 2010).

Light jackups may have legs made from simple tubular H section steel, which is installed into the seabed by penetration as long as the required is obtained. It is also possible to use a flat bearing plate (Dean 2010). In new and modern types of large jackups, usually a double- cone arrangement, consisting of a smaller central cone to assist the unit installation in the seabed is used. The double-cone arrangement has hexagonal or octagonal plans, and can be considered circular for the geotechnical analysis. Recently, skirted spudcans as a new footing type are used (Svanø and Tjelta 1993; Eide et al. 1996; Jostad and Andersen 2006; Andersen et al. 2008). Vol. 21 [2016], Bund. 1 14

Figure 9: Different types of foundations used for different jackups (Image from Dean 2010, in which modified and collected from Young et al., 1984; McNeilan and Bugno, 1985; Hambly et al., 1990; Hayward et al., 2003).

Jackup is a mobile structure and before establishment in the new location, following site assessment should be performed according to SNAME (2002): a) a geohazard assessment b) assessment of the foundation for installation, e.g. a preload check. c) assessment of the foundation for remote operations, e.g. a sliding check, and an overturning check d) assesmemt of jackup effects on nearby structures e) assessment of leg extraction when the jackup is moved to the another location. SNAME recommends that if the preloading a jackup to two times of the working vertical load is safe; it highly would be safe to use the jackup at that location. If it does not show a safe jackup preloading step or if there is no ballast tank capability for performing the test, following tests should be performed: a) bearing capacity and sliding check b) displacements check (Dean 2010). In-situ geotechnical investigation at the location of jackup platform should consist of drilling one borehole below the seafloor to a depth of 30 m, or to 1.5 spudcan diameters below the bottom line of spudcan, wichever is the greater. Also it should include one cone penetration test (CPT) with the distance of few metres from the main borehole to a depth of at least 20m. More boreholes would be necessary if some potential problems found in the main borehole or some information is missing. Sea floor hazard assessment is necessary to find problems in the sea floor (fig. 10) and prevents local overstress in the steel, and excessive bending moments in the legs. Vol. 21 [2016], Bund. 1 15

Figure 10: Types of seafloor hazards in spudcan installation location. (a) Rock outcrop or stiff seafloor sediments. (b) Footprints, or softened remoulded soils, from former jackup installation. (c) Punch-through failure: a stiff soil layer covers a soft layer. (d) Sloping hard layer or the seabed strata (Dean 2010).

JACKET PLATFORMS Jacket platforms are most common platforms in the offshore industry. The jacket made from an open-framed steel edifice built from tube-shaped leg chords, horizontal bracing, and diagonal bracing (Figure 11). It tolerates loads of a deck and other structures on the top such as helideck, drilling rig, and or office and accommodation set up (Dean 2010). Jackets are used in twine or triple format in some sites for drilling, production and accommodation. Jackets are stabilized against lateral loading by the seabed driven piles which may pass through leg chords, or inside of peripheral pile sleeves (Dean 2010). Suction caissons are employed as a replacement for the piles in the Europipe 16/11-E jacket, as an exception (Tjelta, 1994, 1995). The jackets design life is usually 20 years, but can be shorter or longer depending on the project. The main environmental forces during work life of jackets originate from wind, waves, and currents, and sometimes ice forces (Dean 2010). Loads that can act on the jacket foundations are shown schematically in figure 11.

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Figure 11: Types of load on 4-legged jacket foundation (Dean 2010)

The net horizontal load (Hnet) is from wind, wave, and current loads, and changes with time. The net vertical load (Wnet) is sum of the buoyant weight of the edifice, the weight topside modules, and vertical loads of the wind, wave, and current. Reaction forces of foundation at the left and right side of jacket can be shown by VA or B for vertical reaction, HA or B for vertical reaction MA or B for moments. Vertical and horizontal loads can be written by the following equation in an equilibrium status.

Wnet = 2VA+2VB (1)

Hnet = 2HA+2HB (2)

Calculating moments regarding to the intersection of the centreline with the seafloor resulted in:

Wnet yw+HnetL = 2MA+2MB+(VB-VA)S (3)

S is the space between legs at the seabed. If it is assumed that horizontal components of A and B are equal, and moment of A and B points are equal to M in simplified form; vertical components of the points A and B can be calculated from the following equations (Dean 2010):

W24 y HL− M V =net (1 −−w ) ( net ) (4) A 42SS

Wnet 24 yw HL net − M VB =(1 −− ) ( ) (5) 42SS Vol. 21 [2016], Bund. 1 17

HA= HB = (Hnet)/4. It can be seen that the horizontal loads impact the vertical, and the lateral loads on the piles. Moment forces are relatively small compared to HnetL. The conventions that HA= HB , and MA = MB are not usually accurate due to the change of the foundation stiffness with different values of the A and B loads. Nevertheless, the above equations are rational for early estimates (Dean 2010). There are different types of hazards (fig. 12) arising from the seabed for jacket platforms. Scours that form as a result of soft sediments that washed away in the sea floor by stormy or cyclic waves. Lateral stress conditions can be changed by the installation of jackup footing close to the jacket structure and destabilize the jacket. Also, seafloor slope failures can happen as a consequence of the liquefaction, flowslide, or turbidity at or close to the jacket foundation location (Dean 2010). The hydrocarbon , usually below some level are uncased and extreme fluid pressures applied for well drilling can create a hydraulic fracture. These fractures can decrease the mechanical integrity of the foundation soil mass (Schotman and Hospers 1992; Andersen and Lunne 1994). Drilling operations can cause other hazards such as cyclic settlements, predominantly in due to vibrations. Shallow gas and gas hydrates are other hazard potentials that can be collected in large spaces close to the foundation, or piles and decrease the lateral and vertical pile capacity (Dean 2010).

Figure 12: The seabed hazards for jackets. (a) Main and local scour by (Souls 1998; Whitehouse 1998). (b) Soil pressed toward the piles in neighbour jackup installation. (c) The seabed instability, from slope failure or flowslide or debris flow or turbidity current. (d) Shallow gas and hydrocarbon accumulation from fracturing or gashydrates and their problems due to drilling (Dean 2010). Vol. 21 [2016], Bund. 1 18

GRAVITY PLATFORMS Gravity platform that is also called gravity-base structure (GBS), (fig. 13) employs its weight to keep its stability against the environmental forces, and has been built for up to 300m water depth (Dean 2010). In the Condeep and Seatank forms of platform, one or several concrete legs are used to support the deck and topsides, transmitting loads to a cellular caisson base (Young et al. 1975; Mo 1976; Andersen 1991). A typical platform of this form, mounted in 100m water depth, usually includes a caisson (100m width by 20-40 m height). Usually, concrete legs in the hollow form can have 20m diameter, such a way that oil wells and gas wells pass through them to the base of the caisson and reach into the seabed.

Figure 13: Different forms of gravity structure (GBS) (Dean, 2010) (a) Condeep and Seatank form (adapted from Poulos 1988). (b) Concept of the Maureen platform built over a preinstalled template (adapted from Berthin et al. 1985). (c) GBS form for wind turbine structure (adapted from Staf, 2003). (d) GBS form for LNG facility (adapted from Raine et al. 2007).

The caisson that can store oil and provide weight for stability may be equipped with dowels, which are short vertical steel piles or concrete piles at the base. Also, vertical walls termed skirted can be used by penetration to the soft seabed and resting in the stronger layers transmitting load below. Skirts can stop ground loss in the platform peripheral by acting as shear keys (Dean, 2010). Vol. 21 [2016], Bund. 1 19

One of the world’s tallest concrete structures is Troll East Gas Platform, mounted in over 300m of water, using skirts penetrated 36m into the soft clay seabed. (Andenæs et al. 1996; Huslid 2001). The geotechnical engineering problems that should be addressed in designing GBS foundation are: Required foundation width; required skirt depth and spacing; installation step; scour potential; effects of accumulative cyclic loading; dynamics of water and earth i.e seismic events; bearing capacity and sliding capacity in dangerous loading events such as the seabed slope failure; liquefaction potential; consolidation potential; immediate and long-term settlements rates; potential (Young et al. 1975; Eide and Andersen 1984; Andersen 1991; Dean 2010). Site investigation procedure for the GBS platforms includes failure mechanisms examination in the seabed to a depth equal to the caisson width, and laterally in area about one caisson diameter distance from the edge of the caisson in clayey soil and several diameters of the caisson if the soil is mainly in sand size. For a small GBS being installed in the shallow water, one borehole for taking sample with a good geophysical survey may be sufficient. A larger GBS, one deep borehole, 100m below the seabed, 4-8 shallow borings down to 30m below the seabed, and 13 cone penetration tests (CPTs) in the upper few metres of the seabed are required (Hitchings et al. 1976). The impact of consolidation of clays, in settlement analysis, and in the approximation of pore water pressures and the development of strengths necessitate the understanding of the compressibility properties, permeability, and consolidation parameters in the clay layers in the GBS geotechnical investigation. The required knowledge can be obtained by extensive oedometer testing, by applying comparatively high values of the vertical , in-situ CPT dissipation tests, laboratory tests, including triaxial extension, compression tests, direct shear in normal and cyclic mode, and possibly simple shear, hollow cylinder, and even true triaxial tests (Dean 2010).

THE OFFSHORE WIND TURBINES FOUNDATION Types of foundations that mostly are used in the offshore wind farms are gravity bases, monopiles, jacket structures built on the piles or caissons, and floating turbines tied to the seafloor by tension anchors as a newer type of foundation (Fig. 7) (Gavin et al. 2011). Where the soil near the seabed level shows suitable bearing capacity, and water is shallow (<25 m) gravity based structures are employed. If near surface soil is not strong enough and water depth is less than 35m, monopiles are used. Monopiles are formed from large-diameter steel tubes that have very high moment resistance, and experience has shown that monopiles are good for the offshore wind turbine farms. Most of the projects done till now are located in the relative shallow water depths (<30m), (20%) employing gravity base foundation, and (75%) monopile foundation (Gavin et al. 2011; Madsen et al. 2012). Most of the developments that are planned for the next decades will be implemented in the water depths of 30 to 70 m. Great amount of research has been carried out on the gravity base structures, and research on monopiles (Fig. 14), and suction caisson-type foundations is in the progress. It is predicted that many of the new, deeper water structures will be established on the jacket structures (Fig. 15) (Gavin et al. 2011). Also, jacket and tripod structures are suitable for deeper waters and require minimum preparation for the installation at the site (Madsen et al. 2012). Since the critical loading phenomena in a single pile under tension for the foundation of the offshore wind turbine differs greatly from current developed and calibrated design methods, Gavin et al. (2011) investigated the accuracy of estimation of the shaft tension capacity for the offshore piles by the current design methods. Madsen et al. (2012) discussed strong and weak point of bucket foundation for the offshore wind turbines. The bucket foundation (Fig. 16) consists of a thin shell structure, that with an increase in depth of water, the diameter of the bucket increases, Vol. 21 [2016], Bund. 1 20 and it has a large aspect ratio between caisson diameter and the wall thickness. Instability in the form of buckling is an important issue for the bucket at installation time (Madsen et al. 2012).

Figure 14: Monopile foundations for the offshore wind turbines (Madsen et al. 2012).

Figure 15: Jacket and tripod foundation structure for the offshore wind turbines (Madsen et al. 2012). Vol. 21 [2016], Bund. 1 21

Figure 16: Bucket foundation structures for the offshore wind turbines (Madsen et al. 2012).

FOUNDATION ANALYSIS AND DESIGN METHODS IN THE OFFSHORE ENVIRONMENTS DESIGN METHOD FOR THE JACKUP PLATFORMS Preloading calculations for the jackup platforms are necessary to find spudcan penetration relative to the leg load (consisting of the static jackup weight and added water ballast), and for the safety checkup especially for punch-through phenomena. The required information for this calculation is the spudcan geometry, jackup properties, and soil layers type, unit weights, and strength properties. In practice, common bearing capacity relations are used to find spudcan tip penetration in different scenarios. It is supposed that the spudcan reaches the seafloor without disturbing it, and then some failure modes happen at the correspond penetration level and the most critical failure is selected as the limiting value (Dean 2010). For the offshore jackup foundations, the situation is different. In the offshore projects the spudcan displaces some soil when it is pressed into the seafloor. Hossain et al. (2003, 2005a, 2006) performed some experimental and finite element modeling about the spudcan penetration in the clay seabed (Fig. 17). By advancing the penetration, an open hole is created, and the soil particles are moved around from the beneath of spudcan. After that, more soil particles move and then flow onto the top of the spudcan; this prevents the wall collapse. Figure 17c depicts the wall failure development in an open hole as a result of the seabed softening resulting in infill material (Dean 2010). Backflow and infill loads on the bearing section, decreases total value that the bearing area can tolerate and can be calculated from the following relation, which is used by SNAME (2002).

' Wmax ≈− qAγ V (6) where Wmax is the maximum buoyant weight of the material on the top of a spudcan. q is the in-situ vertical effective stress at the surface of the bearing location area (A). ᵧ’ is the immersed unit weight of the soil, and V is the volume of soil moved by the spudcan (Dean 2010). Vol. 21 [2016], Bund. 1 22

Figure 17: Steps of spudcan penetration, failure types, backflow, and infill mode. (a) Failure in surface in early penetration into the soft clay (b) Flow failure resulting in backflow in the following more penetration into the soft clay (c) Wall failure producing the infill in next steps of operations (a,b,c: after Hossain et al., 2006). (d) Failure modes in clay with uniform shear strength are compared (Dean 2010).

Detection of the occurrence of backflow can be done using Meyerhof ’s (1972) method for the stability analysis of an unsupported slurry-field trench in clay as recommended by SNAME (2002). As it can be seen from Fig. 17-c, wall failure will occur if the hole depth, D, satisfies the equation (7):

D S > N u (7) γ ' BB

In which su is the mean undrained shear strength over the depth D, and ᵧ’ is the immersed unit weight of the soil averaged, and N is a stability number plotted in SNAME’s (2002) in Fig. 17d as a function of D/B. Hossain et al. (2006) after experimental and modeling research concluded that flow failure happens if the equation (8) get satisfied and can be used for different modes of the backflow in the preloading time:

D SS0.55 >−(uD ) 0.25(uD ) (8) BBγγ'' B

In which suD is the undrained shear strength of the clay located at the depth of spudcan bearing area plan, D is the penetration depth, and B is the diameter of bearing area in the spudcan. Fig 17-d, comparably shows two discussed mechanisms for the wall failure (Dean 2010). If a sand layer is present, the penetration will not cause backflow or infill, unless in the loose sands that backflow or infill may happen in unusual cases, in which spudcan is emplaced in a hollow inside the moving sandbanks. In layered soils, the backflow and inflow may happen if the sand layers collapse, while dragging clay layers (Dean 2010). Vol. 21 [2016], Bund. 1 23

Leg penetration design Leg penetration designs are done by plotting a graph of the leg load versus spudcan tip penetration. Leg load can be calculated as the soil resistance minus the weight of the backflow (Figure 18). Figure 18.a. indicates that the vertical load on the spudcan rises with penetration, and no failure is possible during loading upto the design preload. Also, the ultimate capacity of the spudcan will rise if an extra load is used. Figure 18.b shows a punch-through at point A, before the design achieves. By an increase in the load slightly above A, the foundation cannot tolerate the load, and a quick penetration happens until point B, where the bearing pressure equals to the bearing capacity. This danger should be considered by taking into account the values of the leg load and the leg penetration at the point, and the punch-through distance from A to B. If the distance between A and B is small, the rig movers may control the ballast systems to perfom the penetration without danger or damage (Dean 2010). Figure 18.c shows a small punch-through between C and D indicating a quick penetration due to small changes in soil characteristics. Figure 18d shows no punch-through till design preload, but indicates its occurrence at E if the spudcan was slightly overloaded. Also a potential will exist due to detection of engineering parameters or soil layers with error. Consider a dashed graph with the mentioned conditions, which for that a punch-through will happen at F. To determine factor of safety following relation can be used: Vol. 21 [2016], Bund. 1 24

Figure 18: The leg preloading curves and their interpretation. (a) Curves showing no problems, except if a significant lateral variation exists. (b) Curve showing punch-through in the preloading. (c) Curve showing small controllable punchthrough (d) & (e) The curve showing low factor of safety against punch-through in and after preloading period (f) The curve showing fast penetration and possible P−∆ failure during preloading (Dean 2010). FS=(Leg load at punch-through (E)) / (Maximum planned leg load) Usually, the factor of safety for punch-through failure is low at the maximum preload and should be considered by caution (Dean 2010). Figure 18-e shows a punch-through after the planned preload at the point G. Furthermore, at point H the leg load is lower than design preload, and possibly lower than the working load of the jackup. The graph pattern up to the G is possibly due to the presence of a strong soil layer. Fig 8-f does not show a punch-through whereas the rate of the rise in the ultimate leg load by penetration is minor from I to J hinting three potential problems. The first problem is from possible difference of actual soil characteristics or layering from those applied to the calculations. Hence, the real penetration in the preloading will differ from the predicted values. Vol. 21 [2016], Bund. 1 25

In the second case, a quick penetration is not possible if the rig movers cannot control the rate of ballasting accurately. In third problem, there is a potential for a P failure (Dean 2010).

BEARING CAPACITY CALCULATION Bearing capacity calculation in the jackup platforms considers different types of failure for a spudcan penetrating into the layers of soil profiles. Soil can consist of clay, siliceous sand, and/or siliceous , calcareous and carbonate layers (Dean 2010 ). In the design procedure sands and are considered in a drained status. The effective internal friction angle, obtained from triaxial or or predicted by in-situ test data are used for the soil strength detection. The clays are considered in the undrained status, and soil strength is detected from the undrained shear strength su, obtained from triaxial or vane tests. In silts, calculation should be performed in both undrained and drained status, and conservative results should be used. The mentioned procedure usually assumes one or two soil layers. If more than two soil layers exist, calculation is done from the deepest penetration level. Top first two-layers are combined to an equivalent single layer with same bearing capacity such a way that spudcan is penetrating these two layers and resting on third layer. This technique is repeated as if it is required. Figure 19 adapted from Dean (2010) and changed, presents different bearing capacity failures from penetration of the spudcan into clay using conventional plane strain failure mechanism. In a status where a spudcan placed on a clay underlaid by a softer soil layer, a punching failure can develop. Failure mode observed by Hossain et al. (2005b) using centrifuge test models is shown in Fig. 19-b. SNAME (2002) employes Brown and Meyerhof (1969) method based on a simple model of a vertical soil column under the footing which is punched down toward the weaker soil below. A revised equation by Dean (2008) was introduced as follows:

qu,net= qu,net,b+4αSu H (9) B where qu,net,b is the net total bearing capacity that can be assumed if the spudcan placed on the surface of the lower layer, and su is the mean undrained shearing strength over the height of H. The next part of the equation is related to the shearing stress on the curved exterior of a vertical column which is pushed down under spudcan assumedly. The α parameter, is 1 if the full undrained shear strength of the top clay layer is organised on the surface, but usually for uniformity with SNAME (2002), it will be considered as 3/4. Vol. 21 [2016], Bund. 1 26

Figure 19: The bearing capacity failure modes. (a) Uniform clay layer failure as it was assumed in the common bearing capacity formula. Its mechanism changes from the surface failure to flow failure. (b) Failure mode for a clay over weaker clay: this mode experimentally and numerically confirmed by Hossain et al. (2005a). (c) Failure mode for clay over stronger clay; used solution borrowed from Meyerhof and Chaplin (1953). (d) Failure mode in uniform sand and the general bearing capacity formula was used. (e) Failure mode in sand over clay; as it was reported by Teh et al. (2008) in a model spudcan which penetrates a dense sand layer at the top and then penetrates an underlying soft clay layer (Figures modified after Dean 2010).

In status where a spudcan is placed on a clay layer underlaid with a harder soil layer, the stiffer layer stops the general shear mechanism to penetrate into it and a squeezing mode will initiate. SNAME (2002) employs the method of Meyerhof and Chaplin (1953) that a revised form of its calculation was proposed by Dean (2008) as follows:

(10) qu,net= ( NcFcsFcd +(B/3H) -1)Su (10)

Factors can be found in table 4. Vol. 21 [2016], Bund. 1 27

In the status where a spudcan is placed over a uniform sand layer, the conical tip of a spudcan creates pre-shearing while the spudcan penetrates into the soil, which impacts the bearing capacity greatly as experimented by White et al. (2008). SNAME (2002) employs the common bearing capacity formula, taking into account the failure type shown in Fig. 19d, as written below:

1 ' qu, net =γ BNγγ Fs F γ d +− q( N q F qs F qd 1) (11) 2 Factors can be found in Table 4.

Table 4: Bearing capacity factors (Dean 2010)

SNAME (2002) employs both relations for Nᵧ in the flat footings which should be used with caution due to overpredictions for bearing capacity in the previous projects. In calculation of large spudcans, the friction angle should be considered 5 degrees less than the value obtained in triaxial testing. Cassidy and Houlsby (2002) suggested some bearing capacity factors for the conical footings, though, White et al. (2008) by experimental tests showed that Nᵧ is about 1/2 of the factor for the flat footings, because of pre-shearing effect. In status where a spudcan is placed on the sand layer overlying a clay layer as it was observed in centrifuge model test by Teh et al. (2008) in figure 19-e. For calculation of bearing capacity in this mode the SNAME (2002) equation in a revised form can be written as: Vol. 21 [2016], Bund. 1 28

H '' q=++ q2 (γφ Hq 2 ) K tan (12) u, net u ,, net b B s where qu,net,b is the net ultimate bearing capacity at the surface of the lower layer. In the second factor, a coefficient of punching shear, Ks, is used and can be obtained from a graph in the original reference. It should be noted that the graphs did not give the full range of relevant friction angles. SNAME (2002) recommended Ks.tanØ 3su as lower bound value when punch-through initiates. su is the undrained shear strength value for the lower layer. SNAME (2002) recommends application of the load-spreading method as a replacement for the status of sand layer over clay layer. A supposed foundation at the depth H under the spudcan bearing surface, with the diameter of (B+2H/n), is reflected to tolerate the leg load and the weight of soil above the supposed foundation. The formula for calculating bearing capacity in this mode is as follows:

2H qq=(1 + ) 2 (13) u, net nB u,, net b

SNAME (2002) supposed n=3 to 5, and if n=5 is considered, a lower-bound approximation of the foundation load at the failure is required.

Design method for the Jacket platforms Mudmat design factors At first step, geotechnical calculation for mudmat in the jacket platform is required. Mudmat is a flat hardened plate that rests straight on the seabed and holds the jacket during piling (Fig. 20) (Dean 2010).

Figure 20: Elevated and plan views of different mudmats and jacket leg extensions. (a) Triangular mudmats on a two-bay Jacket at the lowest braces. (b) Square mudmats on a single-bay jacket installed below the lowest bracing (Dean 2010).

For mudmats, it is required to obtain curves of ultimate mudmat capacity versus mudmat dimension, considering different penetration levels into the seabed. More than one graph may be required for various mudmat configurations and likely lateral loading during the piling. The mudmat dimension design is done using the curves and other factors (Stockard, 1981). If the seabed contains uniform soil strata to a depth that is equal or larger than the longest dimension of proposed Vol. 21 [2016], Bund. 1 29 mudmat, just shallow foundation bearing capacity is required. Then the mudmat capacity Q that is tolerating the pure perpendicular load can be obtained from well-known equation of (Dean 2010): = +−γγ'' Q AsNFF{ u c cs cd z z} For clays (14)

1 ''' =γγγ +− (15) QA BNFFγγs γ d zNFFq qs qd  z For sands 2

In which, A is the mudmat bearing area, su is the undrained shear strength of clays, z is the mudmat penetration height beneath the mudline, and the other are general bearing capacity factors. The term - ᵞ’Z shows the effect of soil flow to the top of the mudmat. Helfrich et al. (1980) consider this term related to the Nq component for granular soils calculations. It will be taken into account if backflow happens during the period of temporary support. Where there is right angle triangular mudmats shape factors are obtained for the following corresponding foundations (Fig. 21) (Helfrich et al 1980). In clay (undrained) soils, calculate L=B by applying L as the length of the hypotenuse and B as the smallest height, e.g. B/L=0.5 in a 90-45-90 degrees triangle. In granular (drained) soils, suppose the triangle equal to a circle with same area, so B/L=1. In the layered soil profiles, the bearing capacity calculations of jackups with shape factors of the mudmat can be employed.

Figure 21: Calculations for mudmats. (a) L and B for the shape factors of mudmat in the plan view (Helfrich et al., 1980). (b) A typical size selection from site-specific design chart (after Helfrich et al., 1980). (c) Soil reaction mechanism for the jacket leg extension showing Soil displacement when the leg pushed down vertically into the seabed. (d) Components of the soil reaction (Dean 2010). (Dean 2010).

The jacket leg extension factors The jacket leg extension is another part that requires geotechnical calculation. The soil displacement and resulting resistance around a jacket leg extension in the pushed status into the sea floor can be seen in Fig. 21. Some soil in the passive wedge moves upward, and some soil falls creating an active wedge as a result of vertical penetration of the leg extension. When the extension penetrates the soil vertically, some soil in the deeper soil movements due to penetration create a complicated group of reaction forces that can be determined by ultimate axial pile capacity calculations. These calculations are applied to the shaft friction resistance, QS which is active on the outside of the leg extension, and is applied to the end bearing pressure of QP. Also, ultimate lateral pile capacity calculation is employed for the lateral resistance QL. Vol. 21 [2016], Bund. 1 30

In the practice, the real ultimate shaft resistances are under influence of the lateral load, and the ultimate lateral load is under influence of the shaft load. The QS and QL both have an impact on the QP in short piles and extensions. It is better to calculate each resistance force separately and combine with engineering judgement (Dean 2010).

Pile design factors Pipe pile (Fig. 22a), driven into the seabed by a hammer, is used frequently in the offshore projects. Its dimensions include D or D0, or the outside diameter, and a pipe wall thickness, t or w. The internal diameter is determined by Di =D - 2t, and pipes diameters are multiples of 6 inches, with wall thickness of 1/4 inch intervals. Common D/t ratios are found between 20 and 60. Low ultimate capacity and low friction between piles and carbonate sands or rocky seabed exist. These factors and driving difficulty in the rocks make the drilled and grouted piles (Fig. 22b) better options in these seabed materials (Dean 2010). Full displacement piles, piles, and belled piles are other options (Fig. 22c).

Figure 22: Different types of the offshore piles: (a) Pipe pile, (b) Drilled and grouted pile, (c) Belled pipe (Dean 2010).

A pile installed in some level below the seabed has a buoyant weight W’ which is the weight of the pile under influence of buoyancy in water for section above the seabed and section in soil below it. A vertical load V at the head of the pile is applied to drive pile into the seabed. The total buoyant load from the equation: Q = V+W’ if is plotted against pile settlement, s, produces a peak in the graph which represents ultimate axial pile capacity. Ultimate axial capacity and soil reaction schematically are shown in (Fig. 23) for coring and plugged situations, and related calculation can be found in the figure title. Vol. 21 [2016], Bund. 1 31

Figure 23: Presentation of the ultimate axial capacity and soil reaction mechanism in the coring and plugged states. (a) Pile status as installed before loading. (b) Coring in which pile cuts by penetrating into the soil and Total resistance = Qsx+Qsi+Qa. (c) Plugged status in which soil plug is pulled down by the pile, and Total resistance = Qsx+Qe+Qa. (d) Failure mechanisms in the soil (Dean 2010).

CPT-BASED DESIGN Increasing demand for large-diameter open-ended driven piles in the offshore projects fueled some research projects to investigate the relationship between qc and the pile base and shaft with differentiation between open and closed ended piles. Methods such as Fugro (Fugro-05), Imperial College (ICP-05), and Norwegian Geotechnical Institute (NGI-05), the University of Western Australia (UWA-05), were developed that have been described in Kolk et al. (2005), Jardine et al. (2005), and Clausen et al. (2005), Xu et al. (2005; 2008), respectively. The ultimate base resistance, qb0.1, is determined as the resistant in the pile head which displaced almost 10% of the pile diameter (D) in above mentioned methods. CPT-based design, methods are obligatory in loose sands, because API method for loose sand is not conservative according to ISO 19902. The four CPT-based methods as mentioned above in the ISO 19902 report are modified to equivalent full methods which can be found in the literature. Comparisons between the API and simplified methods in ISO 19902 for a 60 by 2 inches pile driven in the medium-dense siliceous sand are shown in Fig. 24a. The first graph on the left depicts the supposed cone penetration resistance, obtained from equation A.17.4-21 of ISO 19902, originally derived from Ticino sand data (Baldi et al. 1986; Jamiolkowski et al. 1988). The second graph depicts the unit shaft friction. The installed length of the pile does not influence the unit skin friction value in the API calculation while they have an impact in the CPT-based methods. Stress history impact was considered to plot values of 100m length of the installed pile. Unit end bearing values were compared at the third graph. In the ICP-05 method, the unit end- bearing only was considered over the pile end. The variations between the ultimate axial pile capacity and the length of installed pile are shown in the fourth graph. The API method yields the highest value for 100m installed length, and the UWA-05 (plugged) method produce a capacity almost one-half of the API value. Vol. 21 [2016], Bund. 1 32

Comparisons for the same pile in the dense sand, attributing a relative density of 80% was presented in Fig.24b. Unit end bearings conform well in all methods except for the ICP-05 method. There is a good agreement between methods for the ultimate axial pile capacities. At the 100m pile penetration, the values obtained from the NGI-05 method are about 50% larger than the API method.

PRACTICAL P-Y CURVES The relation between lateral load and displacement of the pile head is an important factor in designing piles. API RP2A and ISO 19902 provide shapes for the p—y curves which can be employed for the piles driven into layered soils composed of the soft clays or siliceous sand in the design step. The symbol ‘p’ used by API to show a stress in the recommendations for clay, but employed the same symbol to show the force per unit length in the sand. Also in ISO 19902, ‘p’ is used as the force per unit length for both sand and clay. The recommendations for the soft clays have been shown in Fig.25a. The ratio of the lateral soil resistance to the ultimate lateral resistance pult is plotted on vertical axes. The ratio of the lateral deflection y is divided by a reference deflection yc = 2.5 cD, and is shown on the horizontal axis. D represents the pile diameter and c represents the axial strain obtained in the undrained triaxial compression test. D is obtained at the time that the deviator stress during the test has achieved 50% of the ultimate deviator stress. Values listed in Fig. 25e were recommended by Rees and Van Impe (2001) for normally consolidated clays, considering shear strength. API RP2A and ISO 19902 made several p—y curves available. In short-term static loading, the full reaction changes at the greater lateral deflections. In equilibrium conditions of cyclic loading, the form of curve changes regarding whether the clay is inside the zone of decreased resistance, or not (Dean 2010). Typical shapes suggested by Reese and Van Impe (2001) for the soft clay after cyclic loading is shown in Fig. 25b. Zero resistance is considered until a lateral deflection is achieved during cyclic loading, and after that the curve proceeds to the API curve for cyclic loading. In the stiff clay, which is a clay showing undrained shear strength more than 96 kPa, it can be observed that stiff clays may be more brittle than soft ones, and may show greater cyclic degradation. The recommendations for the siliceous sands are shown in Fig. 25c which illustrates curves based on a hyperbolic tangent function. The constant k represents the rate of an increase in the beginning reaction modulus versus depth in the sand. The constant k changes with the angle of internal friction of the soil (see Fig. 25e). The parameter A can be obtained from Ax=max{ 3 − 0.8( / D),0.9} for static conditions, or as A=0.9 in an equilibrium cyclic conditions, in which x is the depth below the seafloor. The p—y curves suggested by Williams et al. (1988) and Wesselink et al. (1988) for calcareous and carbonate sands are shown in Fig. 25d.

ULTIMATE LATERAL PILE CAPACITY The maximum horizontal load that a pile can tolerate at a specific point is called ultimate lateral pile capacity. In the jacket platforms, the specific point can be the level of connection between the pile and the jacket, or the connection of the lowest horizontal bracing level on the jacket. Plastic failure mechanism in the soil and the pile are taken into account in the calculation of ultimate lateral pile capacity. A shallow failure mechanism in the soil just below the seafloor as can be seen from Fig. 26a happens when the pile moves from left to right, and pushes a passive wedge of soil upward. Ground may heave close to the laterally loaded pile up to several times of its diameter (Reese and Van Impe, 2001). A gap or an active failure zone or a gap builds up behind the pile. Different failure mechanism may happen at the deeper level where the prevents vertical movement of the pile. At this situation, the soil moves around the pile when the pile moves, however remains in the same horizontal level during the movement. The ultimate lateral Vol. 21 [2016], Bund. 1 33

resistances pult determination for the two failure mechanisms has been explained in API RP2A and ISO 19902, and is summarised below. A representative plot from the results of uniform soil profile is shown in Fig. 26c. In the shallow failure mechanism, reduced resistance, deep failure mechanism happens. In the layered soils, the graph shows several intersections, in which ISO 19902 recommends the first intersection should be the depth of the reduced resistance. As mentioned the term ‘p’ is used for clays and sands in API, and ISO 19902 methods. Calculation of the ultimate force per unit length, pult, at a depth x in soft clay soils under static loading for shallow and deep failure can be written by equations of (16) and (17), respectively:

pult = 3suD + ’vD+Jxsu (16)

pult = 9sσ uD (17) where su is the undrained shear strength at depth x, D is the pile diameter, σ’v is the vertical effective stress at depth x, and J is an experimental parameter varying between 0.25-0.5. J is typically considered 0.5 if no representative data exist (Dean 2010).

For underconsolidated soils, σ’v derivation should consider actual pore pressures in the soil. usually, the depth of reduced resistance (xR) can be obtained from:

(18) 6suD = σ’vD+JxR su

In normally consolidated soils, su is roughly 0.2 σ’v (Semple and Gemeinhardt, 1981), so the reduced resistance zone depth would be about 6D=J, or almost 12 times of pile diameters if J= 0.5. In over-consolidated soils, su is greater, producing smaller depths xR (Dean 2010). Vol. 21 [2016], Bund. 1 34

b

Vol. 21 [2016], Bund. 1 35

Figure 25: Different practical p—y curves. (a) For soft clay (API RP2A and ISO 19902). (b) For soft clay after cyclic loading (Matlock, 1970; Reese and Van Impe, 2001). (c) For siliceous sand (API RP2A and ISO 19902). (d) For carbonate sands, static loading, shown for pult/pref= 0.9, obtained by Novello’s (2000) equation. (Image from Dean 2010). In the siliceous sands under static loading, the API and ISO formulas for determining the ultimate force per unit length, pult, at a depth x for shallow and deep failure mechanisms can be written as equations (19) and (20), respectively:

pult = C1x + C2D ’v (19)

pult = C3D ’v σ (20)

In which C coefficients can be obtained from a σgraph, presented in the first, two graphs of Fig. 26c. They are the C coefficients, changing with the angle of internal friction ’ of the soil. The depth of the reduced resistance zone can be obtained from the equation 21: φ

xR = [(C3-C2)/C1] D (21)

∙ Vol. 21 [2016], Bund. 1 36

The third diagram in the Fig. 26c illustrates the inferred depth ratios xR=D. The depth of reduced resistance changes from about 10 times of the pile diameters if ’=20°, to almost 22 times of the pile diameters, if ’=40° (Dean 2010). φ φ

Figure 26: Different soil failure mechanisms in the lateral loading. (a) Shallow soil failure mode shown in elevation (top) and plan view (bottom). (b) Deep soil failure mode shown in elevation (top) and plan view (bottom) (after Fleming et al., 1992). (c) Calculation graph for the depth of the reduced resistance zone. (d) Coefficients for siliceous sand (after API RP2A and ISO 19902) and H=D ratios for the depth of the reduced resistance zone (Dean 2010). Vol. 21 [2016], Bund. 1 37

DESIGN METHOD FOR THE GRAVITY PLATFORMS The gravity platforms are usually retained in level status when they are lowered to the seabed. By this way, the dowels and the skirts can start to penetrate the seabed almost concurrently. Similarly, Fig. 27 shows the major design issues for parallel mode.

Figure 27: The major design aspects of the parallel descent installation. (a) Dowel penetration. (b) Penetrated dowel and skirt with dome contact stress in the grouted mode and protected scour (Dean 2010).

The dowels can be defined as the vertical steel piles with the diameter up to about 2m (Gerwick 2007). Vertical resistance calculation for these dowels can be found precisely in the Section 6 of DNV (1992). Two types of calculations are suggested, one for the most common soil resistance, the other one for the highest probable resistance. The dowels follow pile design procedure, with special considerations if reducer for the friction is applied. Skirts and ribs also follow the pile design procedure. A general equation for these three elements can be written as:

R= Rs + Rp (22) where R is the net soil resistance against the penetration, Rs is the skin friction at the either side of a dowel, or the wall of a skirt, and Rp is the toe bearing pressure of the dowel or skirt. The lateral resistance calculations for the dowels are similar to the lateral pile capacity (Dean 2010). Skirted foundations and anchors are appropriate solutions for the different types of fixed offshore platforms, and possible alternative for the offshore wind turbines (Ibsen and Thilsted, 2011). An important consideration during the design of these structures is the design for cyclic and dynamic loading. The cyclic loading induces excess pore pressures in clays, silts, and even sandy soils under a wide caisson. The pore pressure along with strain—history effects, can affect the stiffness and strength of a soil. In less permeable soils, these effects can last for several weeks or months, and add up with each successive storm. In stress path analysis for the above problem, a number of key locations based on calculation in the soil are selected at the first step. Fig. 28a illustrates six locations A-F that can be a potential failure surface in the soil. Each key point should show the initial soil state prior to the platform installation plotted on both diagrams, and will be updated with the design effects. An average effective stress versus a shear or deviator stress can be seen in the first diagram. Here, in Fig. 28, the triaxial parameters p0, q are used, but it can be more suitable to use the vertical effective stress versus the shear stress on a horizontal plane, or a Mohr’s circle diagram for some key points. The Fig. 28c shows second diagram at a key point that represents the volume value change. The mean normal effective stress is plotted versus the . In clays, (NCL) the onedimensional asymptotic compression line and elastic swelling or Vol. 21 [2016], Bund. 1 38 recompression lines (EL) can be plotted by the data. The critical states line (CSL) can be plotted as a line or curve geometry by mimicing the shape of the onedimensional compression curve on the diagram. Its points can be obtained from the critical states of the triaxial test data.

Figure 28: The stress-path approach (a) Finding key points. (b) Stress space and path plot (NTL, no tension line; FL, failure line; CSL, critical state line; YL, yield locus; V, in-situ state; P, pre-consolidated state). (c) Stress versus volume plot, volume is same as the void ratio here (NCL, one-dimensional compression line; EL, elastic line). (d) Interactive Stress—path analysis and laboratory testing (Dean 2010).

A major design challenge for the skirted foundations in sand is to push the skirted in required depth for desired capacity. To ease the high penetration resistance in the sands, suction assisted method is selected. Suction installation method may create the piping channels that break the hydraulic seal and halt further penetration. A numerical modeling by (Ibsen and Thilsted 2011) investigated the failure limits during suction installation in both homogenous and layered soil profile. They performed a numerical flow analysis to obtain the hydraulic gradients produced in the reaction to the suction applied, and showed the results in a closed form solutions. These solutions compared with the large scale tests, implemented in a natural seabed at a test site in Frederikshavn, Denmark, can be used to evaluate the suction thresholds against piping. These closed form solutions are valid for other the offshore skirted foundations and anchors in homogeneous or layered sand (Ibsen and Thilsted 2011). Figs. 29 and 30 illustrate applied suction, and the results of numerical modeling respectively. As it can be seen from Fig. 29 the suction required for the installation of bucket 5 is higher than the suction resulting in the piping.

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Figure 29: The applied suction “p” that was required for the installation of the buckets.

Figure 30: The results of the numerical modeling (FLAC) plotted as the normalized seepage length in exit gradient versus relative penetration. a) Installation condition in homogenous sand. b) Installation condition in sand over a flow boundary (Ibsen and Thilsted 2011).

Ibsen and Thilsted (2011) compared the numerical studies with the installation tests and concluded that, the exit gradient beside the skirt control time of the piping occurrence. In the installation mode for the homogeneous sand, the internal hydraulic gradients numerical studies by different researchers have produced different formulations. Nevertheless, the empirical relations give similar critical suctions for the skirt penetrations in a practical depth (Ibsen and Thilsted 2011). Andersen et al., (2008) suggested a method for calculating the penetration resistance and required Vol. 21 [2016], Bund. 1 40 underpressure for skirts getting penetrated in the dense sand with or without overlaying or underlying clay layers. These suggestions used data from prototypes, field tests, and laboratory modeling in the dense sand. The results and their comparison demonstrated that both inducing extra penetration force and reducing penetration resistance from the under-pressure (suction) eases the skirt penetration in the sand significantly. It was also observed that interbedded clay layers can stop water flows from the sand and eradicate the positive drop in the penetration resistance (Andersen et al., 2008). Shallow sliding failure modes (Fig. 31) are analysed considering that the skirts can tolerate the implied loads. Each mode will need to be tested in design procedure, typically by using a sliding block analysis, plasticity method, or using limit equilibrium (Chen and Liu, 1990). In all deep seated cases (Fig. 32), the three-dimensional nature of the failure surface should be considered. These failure modes can be analysed by the plasticity theory, or adapting the method of slices used in the slope stability problems depending on the soil properties. Variety of slip surface depth, centres and radii for the curved parts should be tried until the lowest factor of safety is determined. Additionally, a finite element analysis by an elasto-plastic constitutive model for the soil coupled with a realistic failure criterion can be employed (Dean 2010).

Figure 31: Different types of shallow sliding failure (adapted from Young et al., 1975). (a) Passive wedge failure. (b) Deep passive failure. (c) Sliding base failure. (d) Sliding failure occurred in a shallow weak layer along with largely spaced skirts, can be prevented by decreasing the skirt spacing adequately. (e) Sliding failure in a deeper weak zone that can be prevented by increasing the skirt length adequately (Dean 2010). Vol. 21 [2016], Bund. 1 41

Figure 32: Deep seated failure analysis. (a) Slip surface for the joint vertical, horizontal, and moment loading (adapted from Lauritzen and Schjetne, 1976). (b) CARL and CARV failure surfaces (after Andersen, 1991). (c) Generalised failure surface through a weak zone, analysed using the method of slices (adapted from Young et al. 1975). (d) Sliding block analysis (adapted from Georgiadis and Michalopoulos, 1985) (Dean 2010).

Mana et al., (2012) investigated the kinematic soil failure mechanisms close to the offshore skirted foundations, placed in a little over-consolidated clay, under undrained compression and tension. They conducted digital image examination of the drum centrifuge tests and compared to the results of a finiteelement analyses. It was observed through analysis of images taken in the centrifuge tests (Fig. 33) that different kinematic mechanisms control the soil failure in the tension and compression modes.

Figure 33: The particle image velocimetry (PIV) analysis example for d/D = 0.5 under compression loading in a centrifuge at 200g (Mana et al., 2012). Vol. 21 [2016], Bund. 1 42

During the tension mode, a reverse end bearing mechanism with a bulb of soil under the foundation with continuous mobilization in low ratio of 0.1 for the skirt depth to the foundation exists. Results of the full model centrifuge test showed that the bearing capacity factor is not influenced by the different types of failure mechanisms in the compression and uplift modes. Some noticeable difference beside good agreement (Fig. 34) was observed between the failure mechanisms happened in the centrifuge tests and those predicted by the finite-element analyses (Mana et al., 2012).

Settlement problem in the gravity structures can be an immediate settlement, a gradual settlements from the cyclic loading and variation of the load condition during the project, and long- term settlements due to the primary and secondary consolidation and regional subsidence (Dean 2010). Immediate settlements happen at the time of load application, and can be elastic or elasto- plastic. They can be found by a finite element program using an appropriate elasto-plastic constitutive model, or by settlement calculation for shallow foundation (e.g. Bowles, 1996; Das, 2005). For the primary consolidation, an example is shown in Fig. 35. A gravity base emplaced on a relatively thin layer of clay with height of H, which is overlying a sand layer acting as a drain path is shown. Due to the small H compared to the caisson breadth B, the water will flow vertically into the sand layer. Fig. 35.b. illustrates a different situation where the radial consolidation equation is more applicable. The gravity base is shown, emplaced on a thin compressible layer covering comparatively impermeable soil or rock. Water flows radially in this case as a primary option (Dean 2010).

Figure 34: The comparative evaluation of the velocity contours observed in PIV and predicted by FEA in (a) compression mode and (b) uplift mode (Mana et al. 2012). Vol. 21 [2016], Bund. 1 43

Figure 35: The basic models of consolidation under a gravity platform emplaced on a thin layer of relatively compressible clay. (a) Almost one-dimensional consolidation of a thin clay layer over a comparatively permeable layer. (b) Almost radial consolidation of a thin clay layer over a comparatively impermeable layer. (c) Initial vertical total stress under the caisson base. (d) The degree of consolidation Uv versus the time factor Tv (Dean 2010).

DESIGN METHOD FOR THE OFFSHORE WIND TURBINES FOUNDATION Monopiles are widely used for the offshore wind turbine foundation. Pile design methods used for the oil and gas rigs and evolved to API-RP2A code during time are adapted for this sector too. In the American Petroleum Institute (API) first edition (API 1969), the local shear resistance (τf) was approximated by a usual earth pressure approach:

' tσfV= K tan δ f (23) where σ’v is the vertical effective stress, K is an earth pressure, and δf the interface friction angle. K values of 0.7 and 0.5 related to the compression and tension loading correspondingly is considered empirically. δf changes with the soil type, evidently (Gavin et al. 2011). Lings (1985), discussed an important density bias using the API method. This method overpredicts the capacity of the piles in loose sand and considerably underpredicts the capacity of piles in dense sand. Van Weele (1989) which approved these findings mentioned the impact of the bias results on the economical design of foundations in the North Sea sand, which is heavily overconsolidated, and in very dense sand. The problematic estimation of the working earth pressure coefficient K, and variation in the sand deposits naturally, causes the techniques that use average of the shaft resistance or soil properties along the pile shaft seem questionable. This doubtful estimation increased the application of the correlations between δf and insitu test parameters, primarily obtained by CPT tests (Gavin et al. Vol. 21 [2016], Bund. 1 44

2011). The tensional loads applied vertically to the foundations of wind turbines are much more than those obtained during the calibration of the offshore design methods. When the design techniques optimized and based on CPT methods were used to determine the required pile length for supporting the usual wind turbine loads, contradictory results (Fig. 36) were observed by Gavin et al. (2011). Radial effective stresses detected in the instrumented open-ended piles show that (Fig. 37) the semi- empirical factors do not correctly represent the differences in the radial stress governing around closed- and open-ended piles (Gavin et al. 2011). These values exposed that the piles installed in loose sand tolerated much greater rates of radial stress degradation in the cyclic loading than the piles installed in dense sand. However, the piles installed in loose sand show much greater values of friction fatigue during the cyclic loading (Gavin et al. 2011).

. Figure 36: Approximation for a pile length necessary to support a 5 MW wind turbine with D = 1.5 m Gavin et al. 2011).

Seidel and Coronel (2011) developed a new method, combining accredited methods of static (ICP) and cyclic (RATZ software) performance to evaluate cyclic loading effect on the axially loaded piles of offshore wind turbines. Accurate predictions were obtained for the deflections, cyclic failure mechanisms and capacity reductions by this method. This method can predict the pile failure under cyclic loading with RATZ software if the input values for the peak skin friction are realistic like ICP method (Fig. 38). The methodology developed further by using the degraded profile from one test as input for the next (Fig. 39). The reduction based on the overall reduction was selected and peak shaft friction relative to the entire length of the pile was reduced (Seidel and Coronel, 2011). The major hypothesis for the defining pile capacity after the failure was the selection of the maximum reduction due to cyclic loading. This reduction is recognized by the post-peak residual value of skin friction, which was selected as the 70% of the peak value in research of Seidel and Coronel (2011).

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Figure 37: Assessment of the measured radial stress on the open and closed-ended piles: (a) closed end; (b) OE1; (c) OE2; (d) An incremental filling ratio (Gavin et al. 2011).

Figure 38: Evaluation of results from RATZ with the tests pile R3 (cyclic test 2.R3.CY2) in compared mode (Seidel and Coronel, 2011) Vol. 21 [2016], Bund. 1 46

Figure 39: Comparison of Skin friction profiles over depth after test 2.R3.CY2 (Seidel and Coronel, 2011).

Figure 40: The comparison of deflection lines of monopiles (D = 5.0m and 7.5m) calculated using p-y (Lpile software, 2000) and FEM (Achmus et al., 2009).

Achmus et al. (2008) performed 3D finite element analysis to study the behavior of monotonic and cyclic loading considering the interaction between the pile and the . In their study, a particular numerical approach was used to consider the effect of cyclic loading. Achmus et al. (2008) noticed that the common p-y method according to API is not appropriate to determine the behavior of 5m and 7.5m- diameter piles (Fig. 40). The degradation stiffness model was developed by Achmus et al., (2009) in the form of a numerical modeling to investigate the cyclic lateral loading and pile deformations in a practical manner (Fig. 41). A Vol. 21 [2016], Bund. 1 47 comparison between the results from this model and the existing test results for piles in the sand presented good agreement. Achmus et al., (2009) observed that the displacement accumulation rate in the monopiles, considerably depends on the loading level. The loading level is the relative amount of actual load to ultimate load. The ultimate load of a horizontally loaded pile depends mainly on the installed pile length. Hence, the accumulation rate of any level, is principally controlled by the installed length of a monopile (Figs. 41 and 42). The pile diameter has less impact on the ultimate load, comparatively (Achmus et al., 2009). Using the results of the parametric study, Achmus et al., (2009) developed some design charts for both static and cyclic loading. Using these charts, displacement of a monopile in sand at the level of seafloor, and at the level where the load is applied, depending on load cycles can be determined in the static and cyclic loading modes.

Figure 41: Change in the stiffness in two pile–soil systems versus the number of load cycles (Achmus et al., 2009).

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Figure 42: Evaluation of the lateral deflection Figure 43: Design chart for the lateral of the piles using the degradation stiffness displacement of monopiles tolerating static model (Achmus et al., 2009). lateral load versus normalized load (Achmus et al., 2009).

Figure 44: Incremental rate of the displacement versus normalized load under cyclic loading (Achmus et al. 2009)

Bucket foundation is getting considerable attention in the offshore wind turbines industry. Ibsen et al. (2005) performed five years R & D project that showed bucket foundation (Fig. 45), is a suitable and feasible option for the wind turbines in an appropriate soils, and water depths approximately up to 40 meters. Its cost effectiveness because of the reduced steel weight by half compared to monopile, and easiness of the installation are main favourable aspects of the bucket foundations (Ibsen et al., 2005). A bucket foundation design method emphasizing the service performance was suggested in Ibsen et al. (2005) research. The bucket dimensions were determined using the load combination and Limit State (ULS, SLS, ALS or FLS) leading to the appropriate largest dimensions. In the offshore wind turbines, the Vol. 21 [2016], Bund. 1 49 design load value is usually emergency stopping of the wind turbine, ice load or fatigue (Ibsen et al. 2005). Investigation of the penetrability of the bucket to reach overcoming status of the driving force (combination of the load on the bucket and the applied suction, not more than the critical suction), against the resistance force of the soil is necessary (Ibsen et al., 2005). Houlsby et al., (2005) using data obtained from the pilot test in Frederikshavn, Denmark, investigated the application of the suction caisson foundation type to the offshore wind turbines. They concluded that: I) suction caissons in monopod or tripod or tetrapod forms can be used in the offshore wind turbines. II) The mixed effect of low vertical load and high horizontal load and the moment is the major point in designing these foundations for the wind turbines (Fig. 46). III) Ultimate capacity in the wind turbines are approached from the stiffness and fatigue perspectives. IV) Monopod foundation designs are mainly influenced by the moment loading. V) Tripod or tetrapod foundation designs are mainly influenced by the tensile loading. VI) The moment-rotation response of caissons in sand was studied by different methods, including the physical model tests and the numerical modeling tests. VII) By increasing the amplitude of the moment, the stiffness reduces and the hysteresis increases.

Figure 45: Embedding a prototype bucket foundation at the test site in Frederikshavn, Denmark : (a) during embedding, (b) at the end of embedment (Achmus et al., 2009).

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Figure 46: A caisson foundation shows loading, and (exaggerated) displacement conventions (Houlsby et al. 2005).

Anchored off shore structures The steady termination of the hydrocarbon reserves in shallow waters has caused the offshore oil and gas industry target water depths now close to 2000 m (Aubeny et al. 2001; Ehlers et al. 2004). The progress toward deep water and ultradeep water demands the substitution of fixed platforms by the floating facilities anchored to the seabed. They need an anchoring systems design to tolerate great vertical components of the load. The offshore floated structures are used in deep water installations. Vertically loaded plate anchors (VLAs) are used in the deepwater (up to about 1500m) temporary or permanent mooring systems for exploration drilling rigs (Colliat 2002). Application of the VLAs in up to 3000m of the water depth in ultra-deepwater moorings can be developed using current methods. Field load tests to enhance the applicability of the suction piles and VLAs in the permanent moorings in the soft deepwater clays were conducted by Colliat (2002). The freefalling torpedo anchor, according to the R&D activity was recognized as the most capable anchor concept, especially for application in the temporary Moored Mobile Offshore Drilling Units (MODUs) at more than 3000m water depth limit (Colliat 2002). Torpedo anchors (Fig. 47) follow a design scheme that after releasing from a particular height above the seabed, they are embedded well by penetration into the seabed. This penetration happens after getting velocities up to 100 km/h in free-fall drop inside the water. The anchors are usually made from metal with the dry weight of 50 - 100 tonnes and an elevation of 10-15 m (NGI website). Another anchoring system is called the Suction Embedded Plate Anchor (SEPLA) (Fig. 48). It usually provides the positive aspects of the suction caissons such as the penetration depth and the geographical location, along with the geotechnical effectiveness and cost-effectiveness of the vertically loaded plate anchors (Wilde et al., 2001). Vol. 21 [2016], Bund. 1 51

Figure 47: The deep penetrating anchor (DPA) anchor for the installation at the Gjøafeltet, Norway (left), and DPA installation set up (right) (Photo: Gemini 2009-04 , Figure: after Lieng et. al, 1999) adapted from NGI website.

Figure 48: a) A schematic SEPLA installation: 1) suction installation; 2) caisson retrieval; 3) anchor keying; 4) mobilised anchor (Gaudin et al., 2005) , b) A real SEPLA (Image credit intermoor.com) Vol. 21 [2016], Bund. 1 52 The ultimate capacity of plate anchors in vertical monotonic mode in the undrained clay, Fu, is usually written in terms of the projected area of anchor or A, undrained shear strength of the soil Su located at the mid-level of the anchor, and a dimensionless factor N, by following equation (Gaudin et al. 2006):

Fu=NASu (24)

A series of centrifuge tests was performed by Gaudin et al. (2006) to study how the suction installation of caisson, its retrieval and locking plate anchor, influence the performance of the suction installed plate anchors. They concluded that the installation of the combined caisson–anchor system can be calculated by former caisson embedment theory with considering the other bearing and frictional resistance from the existence of the plate anchor. Appropriate agreement (Fig. 49) with the experimental data occurred by applying a bearing capacity factor of 7.5 for the caisson and anchor edges, and similar equal internal and external frictional coefficients of 0.4 (Gaudin et al., 2006).

Figure 49: The comparison of the penetration and extraction resistance and related coefficients (Gaudin et al., 2006). However a great advancement has happened in the offshore geotechnical engineering, some failures happen in the offshore structures. More details about them can be found in the mentioned references, especially at Dean (2010).

CONCLUSION From the different topics covered in this review paper, following conclusive points are summarized. The demand for the offshore structures including the oil and gas platforms, wind turbines is increasing rapidly. This increase is due to need for more offshore energy reservoirs, even in the ultra-deep waters and more shares from renewable offshore wind energy. The geophysical techniques including the traditional seismic methods and new methods greatly increase the accuracy of site investigation and Vol. 21 [2016], Bund. 1 53 reduce its costs. New methods in the geophysical exploration including application of the surface waves and ground penetration radar (GPR) in the seabed exploration have become common, recently. CHIRP a new technology, designed to replace pingers and boomers, can operate around central frequency (3 kHz – 40 kHZ) and provide better resolution from sediments close to the seabed. LiDAR, an air-borne laser scanning system started to be used in shallow (Usually less than 30 m to 50 m deep) and clear waters to measure the depth of water and location of the seabed. The most common in-situ testing techniques used in the offshore environments are the piezo cone penetration tests (P-CPT), vane shear test (VST), T-bar and ball probe tests to find soil properties and design appropriate type of foundation. When a CPT should be done at an accurate spot in deep water, or when a series of uninterrupted in-situ tests in the linear format are required, the ROV penetration test unit should be employed. Minicones developed for better stratigraphic outlining and soil parameter detection from vessels, are used in the pipeline and cable route investigations, and small subsea structures, anchors and dredging investigations. Different types of offshore structure, and foundation for the oil and gas plat forms and wind turbines are selected depending on the depth of water, site properties and the purpose of the structure. For the Jackup platform following site assessment should be performed according to SNAME: a) a geohazard assessment b) assessment of the foundation for installation, e.g. a preload check. c) assessment of the foundation for remote operations, e.g. a sliding check, and an overturning check d) assesmemt of jackup effects on nearby structures e) assessment of leg extraction when the jackup is moved to the another location. Jackets, the most common offshore platforms, are used in twine or triple format in some sites for drilling, production and accommodation. The jackets are stabilized against the lateral loading by the seabed driven piles or by suction caissons. Different types of hazards from the seabed can affect the jacket platforms. Scours that form as a result of soft sediments washed away in the sea floor by stormy or cyclic waves, and seafloor slope failures as a consequence of the liquefaction, flowslide, or turbidity at or close to the jacket foundation location are some types of those hazard. Hydraulic fracture that can decrease the mechanical integrity of the foundation soil mass and shallow gas and gas hydrates are other hazard potentials. The gravity base platforms with typical concrete legs in the hollow form with 20m diameter are necessary to be checked for all types of settlement potential using laboratory or in-situ tests. Monopiles, gravity base, and bucket structures are most common type of foundation for offshore wind turbines, respectively, and currently. It can be inferred that application of in-situ data and design principles of traditional offshore foundation should be adjusted for new types of foundation regarding the geometry, applied load, geohazards, and used cautiously. Different approach and criteria should be considered for the offshore wind turbine foundations than the oil and gas platform foundations due to the difference in the applied load and the foundation geometry. The application of numerical analyses combined with the analytical methods for better and accurate results and understanding in the foundation design and geohazard assessment should be considered as a necessary approach.

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